recent developments in design and construction techniques of brick masonry buildings

Recent Developments in Design and Construction

Techniques of Brick Masonry Buildings

3-4 March, 2012

Department of Civil Engineering

Institute of Technology

Banaras Hindu University

Varanasi-221005, India

Proceedings of the

Workshop

Editors

P. K. Singh P. R. Maiti

Recent Developments in Design and Construction

Techniques of Brick Masonry Buildings

3-4 March, 2012

Proceedings of the

Workshop

Organized by




Varanasi-221005, India

Sponsored by

University Grants Commission

New Delhi

(Under SAP Scheme)

Editors

P. K. Singh P. R. Maiti

© Department of Civil Engineering, IT- BHU

March-2012

ISBN: 978-81-921121-1-4

Published by




Varanasi, India

DISCLAIMER: Neither the editors nor Department of

Civil Engineering, IT-BHU is responsible for statements and opinions printed in this publication. Editors and publishers bear no responsibility with regard to accuracy or authenticity of the information contained in this proceedings and do not accept liability of any

kind for any error or omissions towards this publication.

Proceedings of the Workshop on Recent Developments in Design and Construction Techniques of Brick Masonry Buildings, 3-4 March 2012

ii

In Commemoration of 150th Birth Anniversary

Mahamana Pandit Madan Mohan Malaviya ji

(25.12.1861–12.11.1946)

Founder of Banaras Hindu University, Varanasi, India


iii

Preface

Masonry buildings are widely constructed for housing in

rural and urban areas. This type of buildings basically consists of

un-reinforced masonry wall panels with or without confining

element. Earthquake resistant buildings are required to withstand

the largest earthquake of a certain probability that is likely to occur

at their location, and loss of life should be minimized by preventing

their possible damage or collapse.

In the recent earthquakes of Bhuj 2001, Kashmir 2005 and

Sikkim 2011, several masonry houses collapsed, causing loss of life

and properties which occurred due to non-engineered buildings.

These earthquakes have exposed the seismic vulnerability of

construction practices being followed in the country. For centuries,

masonry construction has been used for buildings in the areas

where good quality bricks are economically produced. Confined

brick masonry, i.e. masonry with vertical tie columns and

horizontal bands, represents one of the most widely used

construction systems in India and other parts of the world.

Confinement of brick masonry prevents its brittle failure and

improves the ductility of the masonry when subjected to severe

seismic loading.

Numerical modeling of the seismic behavior of masonry

structures presents a complex problem due to the constitutive

characteristics of the structural materials. In India the seismic

design of the buildings is based on IS 1893-2002, IS 4326-1993

and National Building Code of India-2005. But these codes do not

fully cover this type of construction. However, Euro code covers

confined brick masonry construction in detail. The main objective of

this workshop is to disseminate design and construction practices

of earthquake resistant brick masonry buildings.

P. K. Singh & P. R. Maiti


iv

Acknowledgement

To give shape to the proceedings and the workshop in general a

large number of individuals and groups have contributed in many

ways and it is our pleasure to acknowledge their efforts. We are

extremely thankful to the speaker for their contribution. The

contributory authors deserve praise for their contribution and co-

operation, which is resulted in the timely publication of the

proceedings.

We are especially grateful to our colleagues namely Prof. V. Kumar,

Dr. S. Mandal and Dr. Rajesh Kumar of the Civil Engineering

Department for their support at different stages of the workshop.

We are thankful to University Grants Commission, New Delhi for

providing necessary funds for the workshop.

We wish to acknowledge the help we received from various

individuals and institutions in the preparation of the proceedings.

P. K. Singh & P. R. Maiti


v

Contents

Mahamana Pandit Madan Mohan Malaviya ji

ii

Preface

iii

Acknowledgement

iv

Contents

v-vi

Masonry Structures: Prospects, Problems and Tasks K. S. Jagadish

1-14

Failure and Behavior of Masonry Structures in Recent Sikkim Earthquake 2011

D. Bandyopadhyay and J. S. Ali

15-28

A Systematic Design Approach of Coupled Shear Wall Buildings during Earthquake Dipendu Bhunia

29-58

Effect of Constituent-Characteristics on Durability of Masonry and Concrete Structure

V. Kumar

59-66

Provisions of Different Codes in Brick Masonry Buildings: A Critical Review Rajesh Kumar

67-100


vi

Earthquake Resistant Confined Brick Masonry Buildings P. K. Singh

101-124

Analysis of Confined Brick Masonry Buildings

P. R. Maiti

125-152

A Study on Indian Codes and Performance Based Design

Dipendu Bhunia

153-170

Earthquake Scenario of India and Its Relation to Various

Rock Types Medha Jha

171-184

The Effect of Dynamic Loading on Structural Integrity

Assessment Debasish Khan

185-198


1

Masonry Structures: Prospects, Problems and Tasks

K. S. Jagadish

Formerly Professor, Department of Civil Engineering, Indian

Institute of Science, Bangalore, India

Currently Professor of PG Studies Department of Civil Engineering,

R V College of Engineering

1. INTRODUCTION

Masonry structures have fallen into disrepute in recent years in

India. The reinforced concrete framed structure is considered to be

superior even for two storeyed buildings. Part of the problem is the

dependence on the burnt brick, its energy intensity and the

resource depletion due to loss of top soil. It is however, necessary to

note that there has been a revival of masonry even for moderate

high rise structures in the West. Switzerland and Denmark, who

did not have a steel industry of their own, preferred to built 15 to

16 storey buildings out of high strength bricks which were locally

available. England and U.S also had high rise masonry going up to

17 storeys Figs (1, 2, 3and 4). In the US, the masonry is built out of

hollow concrete blocks which can accommodate vertical

reinforcement for earthquake resistance.


2


3

2. Why Masonry? It is now pertinent to ask why one can think of masonry in the

Indian context. Table-1 presents the energy content and carbon

emission of building blocks, cement and steel. It is seen that the

burnt brick, cement and steel require higher amount of energy than

the other. Their carbon emission is also high. The stabilized mud

block is made using 7% cement addition to sandy soil.


4

Figure 5: SMB being made in soil block press.


5

Fig 6 and 7 shows hollow concrete and hollow clay blocks.

TABLE-1: ENERGY AND EMISSION OF BUILDING MATERIALS

SL.NO

MATERIAL UNIT ENERGY/ UNIT MJ

CO2 / UNIT Kg

1 CEMENT Kg 3.60 0.80

2 STEEL Kg 28.10 2.2 - 2.8

3 BRICK ONE BRICK 3.75 - 4.5 0.33

4 STABILISED MUD BLOCK

BRICK EQUIVALENT

0.90 0.19

5 HOLLOW CONCRETE BLOCK

--DO-- 0.9 - 1.18 0.14 - 0.18

6 HOLLOW CLAY BLOCK

--DO-- 1.80 0.18

7 SANDSTONE BLOCK (BHUJ)

--DO-- 0.88 0.09

8 GRANITE (BANGALORE)

--DO-- 0.00 0.00

9 GEOPOLYMER + SOIL

--DO-- NA 0.06


6

From the table it is clear that use of cement, steel and burnt brick

are not desirable if energy consumption and CO2 emission are to be

reduced. Since buildings in India take up about 30% of the carbon

emission in the country, there is a great need to reduce their

emission. Table-2 describes the energy and emission due to

different building technologies.

Table 2: Energy and Emission due to building technologies

BUILDING TYPE

EMBODIED ENERGY GJ/M2

CARBON EMISSION

T/M2

OPERATIONAL ENERGY, 25

YEARS GJ/M2

CO2, 25 YEARS T/M2

8 STOREY RC FRAME + BRICK IN-

FILL

4.2 0.41 9.3 0.91

4 STOREY RC FRAME + BRICK IN-

FILL

2.7 0.25 9.3 0.91

4 STOREY SMB

MASONRY WITH RC FLOORS

1.33 0.13 9.3 0.91

2 STOREY SMB

MASONRY WITH SMB

FLOOR

0.62 0.06 9.3 0.91

It is clear that the RC frame construction with brick in-fill is the

worst for energy and emission. Masonry using stabilized mud block

(or Hollow Concrete block) leads to 50% less energy and carbon


7

emission. It is hence important to utilize masonry for as great

height as is feasible. With the example of hollow concrete block

(reinforced for earthquake resistance), one can easily think of

masonry buildings for 10 storeys. With stabilized mud blocks one

can construct upto 5 storeys. A large majority of the high rise

buildings in India range from 4 to 10 storeys and it is essential to

explore this option. Already, there are more than 300 buildings

using hollow concrete blocks for high rise housing. There is a hotel

in Nashik going upto 9 storeys built by Mr. Ganesh Kamat of

Ganaka Engineers.


8

Figure 8 & 9: 6 storeyed building, Mumbai and 9 storeyed building in

Nashik.

Figure 10: An SMB wall.


9

3. Barriers to Use of Masonry

It is now necessary to understand why India has missed this

opportunity of using low energy building technique.

a. Firstly, brick production is a time, energy and labour

consuming industry and grinds to a halt in the rainy season. It

also requires significant area of land for making and drying

bricks. Its cost is hence rising rapidly.

b. Engineers of today do not learn masonry design. Two storeyed

buildings are built on thumb rule by using brick of 3.5MPa

strength. For higher storeys, the requirements of brick/ block

strength, type of mortar to be used is not known.

c. There is hardly any research in our universities on masonry so

that recent innovations of hollow concrete blocks, reinforced

masonry and stabilized mud blocks are unknown. Only 4 or 5

reports/ papers have been published in India, between 1947 to

1990.

d. The quality of most of the concrete blocks is very poor and

they cannot be used for more than two storeys. There is a need

to set up quality hollow concrete block manufacturing units

like ‘Besser Co’.

The knowledge that load bearing masonry using hollow concrete

blocks/ stabilized blocks/ hollow burnt clay blocks is cost effective

and energy efficient is not known to the user public or the

professionals.


10

4. Recent Positive Developments

As a corrective to the above lacunae, the Dept of Civil Engineering,

Indian Institute of Science launched a detailed R & D programme

from 1990 onwards. About 6 Ph.D.’s and 3 M.Sc.’s have been

produced under the guidance of the author between 1990 and

2004. Currently, 5 more Ph.D. programmes and several M.Tech

dissertations are underway at the Visvesvaraya Technological

University. Electives on Masonry have been introduced in

Undergraduate and Post-graduate courses. The author is also

working on a Text book on Structural Masonry which is likely to be

published before the end of the year.

Two companies in Bangalore are manufacturing high quality hollow

concrete blocks with strength of 6.0 to 7.0 MPa. They can be

comfortably used upto 5 or 6 storeys. Machines to make stabilized

mud blocks are available in Bangalore, Auroville and New Delhi.

5. Tasks to be undertaken

Courses on structural masonry must be started in all leading

Engineering colleges. Short term courses are to be organised for

training teachers and practicing engineers in masonry. Periodic

conferences and workshops to be organized for wider dissemination

of ideas.


11

6. Outline of masonry research

Structural masonry has been extensively researched in the west.

One can refer to the books by HENDRY (1), SAHLIN (2), DRYSDALE

& HAMID (3), and NARENDRA TALY (4) to obtain comprehensive

information on western literature. This is however inadequate in the

Indian context since our bricks have low strength and lower elastic

modulus. The research thesis by MATTHANA (5), SARANGAPANI

(6), RAGHUNATH (7) AND GUMASTE (8) give comprehensive

information on brick masonry in India. The paper by GUMASTE et

al [9] is also useful.

Research in masonry is based on the strength of masonry unit

(brick or block), strength of mortar and strength of masonry

element like prisms and wallettes. In general the strength of

masonry element is less than the strength of masonry unit and the

ratio may be referred as masonry efficiency. Fig 11 shows the

sketch of typical masonry prisms, Fig 12 shows prisms after test.


12

The strength of masonry wall depends further on the slenderness

ratio and eccentricity of loading. There is hence a need to test storey

height walls. Fig 13 shows a storey height wall under test.


13

Such tests need a very tall loading frame which may not be

available in all colleges. Such frames have been set up at Indian

Institute of Science, B.M.S. College of Engineering and R.V. College

of Engineering.

More detailed research on the strength of walls using hollow

concrete blocks and hollow clay blocks is necessary if high masonry

has to become a reality in India.


14

REFRENCES

[1] A.W.HENDRY, STRUCTURAL MASONRY MACMILLAN PRESS,

LONDON, 1998.

[2] S. SAHLIN, STRUCTURAL MASONRY PRENTICE HALL, N.J., 1971.

[3] R.G. DRYSDALE AND A.A.HAMID, MASONRY STRUCTURES

BEHAVIOUR AND DESIGN, THE MASONRY SOCIETY, BOULDER,

COLORADO, 2008.

[4] NARENDRA TALY, DESIGN OF REINFORCED MASONRY

STRUCTURES MCGRAW HILL, 2001.

[5] M.H. MATTHANA, STRENGTH OF BRICK MASONRY AND

MASONRY WALLS WITH OPENINGS, PH.D. THESIS, DEPT OF CIVIL

ENGINEERING, INDIAN INSTITUTE OF SCIENCE, BANGALORE,

DEC 1996.

[6] G.SARANGAPANI, STUDIES ON THE STRENGTH OF BRICK

MASONRY

[7] PH.D. THESIS, DEPT OF CIVIL ENGINEERING, INDIAN INSTITUTE

OF SCIENCE, BANGALORE, MAY 1998.

[8] S. RAGHUNATH, STATIC & DYNAMIC BEHAVIOUR OF BRICK

MASONRY, PH.D. THESIS, DEPT OF CIVIL ENGINEERING, INDIAN

INSTITUTE OF SCIENCE, BANGALORE, JAN 2003.

[9] K.S. GUMASTE, STUDIES ON THE STRENGTH & ELASTICITY OF

BRICK MASONRY WALLS, PH.D. THESIS, DEPT OF CIVIL

ENGINEERING, INDIAN INSTITUTE OF SCIENCE, BANGALORE,

JAN 2004.

[10] GUMASTE.K.S, K.S.NANJUNDA RAO, B.V.V.REDDY AND

K.S.JAGADISH, STRENGTH & ELASTICITY OF BRICK MASONRY

PRISMS AND WALLETTES UNDER COMPRESSION, MATERIALS

AND STRUCTURES, 40, 241-253, 2007.


15

Failure and Behavior of Masonry Structures in Recent Sikkim Earthquake 2011

D. Bandyopadhyay 1 and J. S. Ali 2

1Associate Professor, Department of Construction Engineering,

Jadavpur University, Kolkata & NPEEE Fellow 2Assistant Professor, Department of Civil Engineering, Alliah

University, Kolkata

1. INTRODUCTION

Most of the structures in structures in Indian subcontinent are

built as unreinforced masonry structures built with bricks or

stones bonded with cement or lime-mud or simply mud mortar.

These structures are normally designed for vertical loads and they

do behave well under that considering the fact that bricks and

stones have a decent compressive strength. As soon as they are

subjected to lateral forces, typical in case of earthquakes, high

shear and flexural forces arise leading to the failures of these

structures. The strength of masonry under these conditions often

depends on the bond between brick and mortar (or stone and

mortar), which is quite poor. This bond is also often very poor when

lime mortars or mud mortars are used. This is quite evident in the

recent Sikkim earthquake 2011, in which large parts of India

including Sikkim, northern parts of West Bengal etc. were affected.

A masonry wall can also undergo in-plane shear stresses if the

inertial forces are in the plane of the wall. Shear failure in the form

of diagonal cracks is observed due to this. However, catastrophic

collapses take place when the wall experiences out-of-plane flexure.

This can bring down a roof and cause more damage. Masonry

buildings with light roofs such as tiled roofs are more vulnerable to

out-of-plane vibrations since the top edge can undergo large

deformations. The behaviour of masonry buildings after an

earthquake is significantly important and useful to identify any

inadequacies in earthquake resistant design. Studying types of

masonry construction, their performance and failure patterns helps

in improving the design and detailing aspects. After the Sikkim

earthquake on the 18th September 2011, causing severe damage in

masonry structures in the region of Sikkim and North Bengal the

authors have visited the affected areas thrice to study the damages

to buildings.

2. The Sikkim Earthquake 2011

The earthquake of magnitude M6.9 struck at 18:10:48 IST on

September 18, 2011 with its epicentre located near India-Nepal

border region, about 68 km NW of Gangtok, Sikkim as shown in

Fig. 1. It was a shallow focus event, which was felt in India, Nepal,

Bhutan, Bangladesh and China. The tremors lasted for about 30-40

seconds and felt in several Indian states such as Assam, parts of

West Bengal, Bihar, Uttar Pradesh, and Delhi. Three aftershocks

were also felt in Sikkim within 30 minutes of the initial earthquake.

About 100 deaths are reported in India including at least 60 in

Sikkim state though the affected area has low population density of

an average of 88 persons/sq. km. The state capital Gangtok is the

biggest city in the area and Chungthang, Lachung and Mangon in

North Sikkim are major towns which have suffered considerable

damage to structures. Kalimpong and Darjeeling towns in north

side of West Bengal have also suffered significant damages

particularly in masonry structures. The affected region lies in the

high risk seismic zones of IV of Indian seismic code IS: 1893, 2002


17

with the expected intensity VIII. This region has experienced

relatively moderate seismicity, with several earthquakes in the last

few decades prior to the recent event on the September 18, 2011.

The earthquake followed by heavy seasonal rains triggered more

than 300 landslides, rock/mudslide causing much devastation.

Landslides cut off the severely affected areas from the rest and

hampered the rescue and relief work in this difficult terrain.

General damage to buildings and other structures agreed well with

the intensity of ground shaking observed at various places.

Other major towns

Aftershock reported by IMD

Main Boundary Thrust

(MDT)

Field Trip on Road

Major towns damaged

Aftershock reported by USGS

Main Central Thrust (MCT)

River / Stream

However, unexpected severe damage in Gangtok and Kalimpong

were also observed.

3. General Observations

Extensive damage to masonry structures like school, Church and

hospital buildings was reported in the worst affected regions of

Sikkim and North Bengal. Many unique and inherently poor

architectural and construction features such as unsymmetrical,

weak partition walls in brick/block masonry or in lightly

reinforced/plain concrete, extended floor plans in upper stories

supported on cantilevered beams and slabs, construction on sloped

ground, unstable slopes, weak retaining walls, poor construction

material etc., significantly added to the seismic vulnerability of

structures. It was common practice in Sikkim to construct

residential buildings using bamboo/wood, prior to early nineties.

These traditional constructions (Shee-khim & Ikra) have better

earthquake resistance as observed in the present and past

earthquakes. Major RC-frame structures both governmental and

private buildings have seriously lacked earthquake-resistant

features compatible to the design level shaking. Most of the RC

buildings in Gangtok suffered varying degree of damage, from

moderate to collapse during this earthquake. The area has a

number of highway and pedestrian bridges on rivers, rivulets, and

gorges. Only minor damage to a few highway bridges was noticed..

The concrete gravity dams of National Hydroelectric Power

Corporation (NHPC) over Teesta River near Dikchu and Rangit River

near Rangit Nagar have not suffered significant damage due to

earthquake shaking or landslide. The poor earthquake performance

of cultural heritage such as monasteries, churches and old school

buildings is a source of concern as almost all historic structures

suffered varying degree of damages in this earthquake. The exterior

walls of these historical structures are constructed of stone

masonry mostly random rubble with low strength mortar. Heavy

damages have been observed to exterior walls at those historical old


19

structures. In Kalimpong, in West Bengal, front masonry elevation

of the historic Church has severly cracked and posed alarming

threat to its safety.

4. Case Studies & Discussion

Studies have been made to two numbers five storied residential

building at Gangtak, one at Temi and two at Mangan. Much of the

construction in Gangtak is of empirically constructed reinforced

concrete (RC) buildings of four to nine stories adjoining each other

on adjacent small plots, with buildings extending to the property A

majority of these buildings exhibited extensive damage to

unreinforced masonry (URM) infill panel walls due to weak masonry

and large unsupported length or heightto-thickness ratio. Most

buildings had a symmetric and uniform grid of beams and columns.

Some buildings that had open stories had severely damaged.

Likewise, buildings with asymmetry in placement of URM infill

walls, causing torsion, also were severely punished. Traditional

Ekra housing made of bamboo or wood framing with lightweight

infill panels of straw and plaster behaved exceptionally well like

past earthquakes. The inadequate stirrups in columns of a building

at Gangtak constructed with bad materials and poorly maintained

have suffered severe damage. The 250 mm square column size for

four story building at Mangan, North Sikkim with bad material have

cracked and damaged to an extent. Landslides have resulted in

differential settlement of column foundations and suffered damage

as observed in Temi, East Sikkim.

Figure 2: Severely Damaged Building under Demolition, Gangtak, Sikkim

Figure 3: Collapse of

Observatory Shed, Mangan

Figure 4: Wall crack

continued to Water Basin,

Mangan


21

A Stone Masonry Historical School Building constructed about 100

years ago at Darjeeling, North Bengal is surveyed in detail. The

building is a three storied combined type of structure, part of which

is constructed as simply unreinforced stone masonry and another

part is retrofitted with reinforced concrete structure with in-filled

stone masonry after severely damaged by the 1934 Bihar-Nepal

earthquake. The masonry portions are built using stones bonded

with cement-lime mortar. The structure is full of Gothic

architectural features which have been largely affected in the recent

earthquake. The ‘C’ type of unsymmetrical plan of the building

suffered significant damages during the earthquake. The legs of ‘C’

are unequal which has further aggravated the plan asymmetry

contrary to the earthquake resistance features. Asymmetric parts

have invited torsion in the structure resulting in out-of-plane

flexural failure. In addition, there was a large number. of non-

structural temporary sheds and other structures like masonry

chimneys and rooms made of wooden roof system used as

dormitory for students, above the second floor of the building.

These structures have undergone large deformations during the

earthquake and being made of brittle masonry materials, have even

collapsed. A bizarre structural arrangement was observed in the

second floor of the structure with floating columns. This floor was

probably constructed later and uniform structural arrangement was

not followed which resulted in vertical asymmetry. Evidently there

was no clear load path in this floor and this floor was worst affected

by the earthquake. Arrangement of staircase plays a very important

role in determining seismic performance of a structure. Since at the

location of staircase there occurs a discontinuity in the floor

diaphragm action and also the stiffness of the staircase region is

inconsistent with other portions of the structure, it is always much

vulnerable to seismic activity. Same has been observed in case of

this structure where vertical cracks along with settlement have

been observed in the region of staircase due to flexural failure.

Fig ure7: View of C shaped School Stone Masonry Building, Darjeeling

Figure 8: Crack above Arch, Ground Floor


23

Failures of several Stone Masonry Historical cottage buildings and

church constructed more than 100 years ago at Kalimpong, North

Bengal is also studied in detail. There are a common features of

vertical cracks initiated from centre of the arches over the ground

floor window continued to the corner of the window seal at the first

floor (Fig.13). In many occasions the key stones are separated and

dislodged. Out of plane failure of random rubble stone masonry

walls is another common failure symptom in these cottages. These

Masonry buildings with light slope roofs appears to be more

vulnerable and responsible for the out-of-plane vibrations since the

top edge can undergo large deformations. The weak bonds between

random sized stones with lime-mud mortar have contributed for the

failures. Uses of random stones in withes without through-stones

have further aggravated the problem.Separations of wall have been

observed at the corners of the outer walls. The long unsupported

length of the front wall of a historical church have separated from

the cross walls and severely cracked.

Figure 9: Out of Plane Failure of Walls above Lintel

Figure 10: Cracked Stone Masonry Wall

Figure 11: View of the

Church, Kalimpong

Figure 12: Cracked Arch Crown

and Masonry

Figure 13: Cracks from Arch Crown to Window


25

5. Concluding Remarks

The damage to built environment, economic loss and human

casualties caused by the Himalayan earthquakes are increasing

rather proportionally with the growth of population and subsequent

settlements in its upper reaches. The general pattern of damage to

structures, particularly of masonry buildings, landslides, etc. is

consistent with the recent M6.9 Sikkim 2011 earthquake, except a

few building collapses due to faulty design and or construction

practices. Monasteries being old and weak were deficient in

strength and needs to be retrofitted against future tremors. It is

unfortunate that society is not adequately prepared and therefore

the seismic risk in the region has risen to unacceptable levels which

may lead to a large-scale disaster. Based on the observations of the

damages caused to a variety of masonry structures during the

Sikkim earthquake 2011, the following conclusions could be drawn.

ü Majority of the multi-storied buildings exhibited extensive

damage to unreinforced masonry (URM) infill panel walls due

to weak masonry and large unsupported length or height to

thickness ratio.

ü Major RC frame structures both governmental and private

buildings have seriously lacked earthquake resistant features

compatible to the design level shaking. The earthquake

followed by heavy seasonal rains triggered many landslides,

rock/mudslide causing much devastation.

ü Unsymmetrical plan, uses of floating columns and aseismic

construction of chimney etc have suffered severe damage in a

three storied stone masonry school building about 100 years

old historical structure.

ü Masonry buildings in mud mortar or lime mortar are prone

to severe damage due to lack of bond strength.

ü Uses of random stones in withes without through-stones

have further aggravated the problem. The failures of such

structures are essentially due to out-of-plane flexure.

ü Masonry with cement mortar (which has higher bond

strength) has generally behaved better, but only good

masonry bonding is not sufficient for earthquake resistance.

ü Traditional constructions (Shee-khim & Ikra) have better

earthquake resistance as observed in the present and past

earthquakes.

ü Use of lintel band, as suggested by the Bureau of Indian

Standards (IS 13828:1993), with additional horizontal bands,

possibly at the seal level and at plinth level seems to be

required for better performance. The horizontal

reinforcement in the lintel band alone does not seem to

improve the ductility to the desired level for stone masonry

structure.

ü The provision of corner reinforcement in corners and

junctions, again as suggested by BIS, has to be properly

bonded with the surrounding masonry possibly with dowels

or keys to prevent separation.


27

ü Masonry buildings with light slope roofs appear to be more

vulnerable and responsible for the out-of-plane vibrations

since the top edge can undergo large deformations.

References

[1] Report on Evaluation of Sikkim 2011 Earthquake damaged

Structures, Jadavpur University, Kolkata.

[2] National Information Centre of Earthquake Engineering, IIT

Kanpur

[3] EERI News Paper, November, 2011

[4] Behaviour of Masonry Structure during Bhuj Earthquake,

2001, IISc, Bangalore.


29

A Systematic Design Approach of Coupled Shear Wall

Buildings during Earthquake

Dipendu Bhunia

Assistant Professor, Department of Civil Engineering, Birla Institute

of Technology & Science, Pilani, Rajasthan, India

1. INTRODUCTION

The growth of population density and shortage of land in urban

areas are two major problems for all developing countries including

India. In order to mitigate these two problems the designers resort

to high-rise buildings, which are rapidly increasing in number, with

various architectural configurations and ingenious use of structural

materials. However, earthquakes are the most critical loading

condition for all land based structures located in the seismically

active regions. The Indian subcontinent is divided into different

seismic zones as indicated by IS 1893 (Part 1) (2002), facilitating

the designer to provide adequate protection against earthquake. A

recent earthquake in India on 26th January, 2001 caused

considerable damage to a large number of RCC high-rise buildings

(number of storey varies from 4 to 15) and tremendous loss of life.

The reasons were: (a) most of the buildings had soft and weak

ground storey that provided open space for parking; (b) poor quality

of concrete in columns and (c) poor detailing of the structural

design (http://www.nicee.org/eqe-iitk/uploads/EQR_Bhuj.pdf).

Therefore, this particular incident has shown that designers and

structural engineers should ensure to offer adequate earthquake

resistant provisions with regard to planning, design and detailing in

high-rise buildings to withstand the effect of an earthquake to

minimize disaster.

As an earthquake resistant system, the use of coupled shear walls

is one of the potential options in comparison with moment resistant

frame (MRF) and shear wall frame combination systems in RCC

high-rise buildings. MRF system and shear wall frame combination

system are controlled by both shear behavior and flexural behavior;

whereas, the behavior of coupled shear walls system is governed by

flexural behavior. However, the behavior of the conventional beam

both in MRF and shear wall frame combination systems is governed

by flexural capacity and the behavior of the coupling beam in

coupled shear walls is governed by shear capacity. During

earthquake, infilled brick masonry cracks in a brittle manner

although earthquake energy dissipates through both inelastic

yielding in beams and columns for MRF and shear wall frame

combination systems; whereas, in coupled shear walls earthquake

energy dissipates through inelastic yielding in the coupling beams

and at the base of the shear walls. Hence, amount of dissipation of

earthquake energy and ductility obtained from both MRF and shear

wall frame combination systems are less than coupled shear walls

system in the high-rise buildings [Jain (1999), Englekirk (2003),

Park and Paulay (1975), Penelis and Kappos (1997), Smith and

Coull (1991), Naeim (2001) & Paulay and Priestley (1992)]. However,

the Indian codes of practice governing the earthquake resistant

design, such as IS 1893 (Part 1) (2002) and IS 4326 (1993) do not

provide specific guidelines with regard to earthquake resistant

design of coupled shear walls. On the other hand, IS 13920 (1993)

gives credence to the coupled shear walls as an earthquake

resistant option but it has incorporated very limited design

guidelines of coupling beams that are inadequate for practical

applications. It requires further investigations and elaborations

before practical use.

Further, it is reasonably well established that it is uneconomical to

design a structure considering its linear behavior during

earthquake as is recognized by the Bureau of Indian Standards [IS

4326 (1993), IS 13920 (1993) and IS 1893 (Part 1) (2002)] currently

in use. Hence an alternative design philosophy needs to be evolved

in the Indian context to consider the post-yield behavior wherein

the damage state is evaluated through deformation considerations.


31

In the present context therefore, performance-based seismic design

(PBSD) can be considered to offer significantly improved solutions

as compared to the conventional design based on linear response

spectrum analysis.

Performance-based seismic design (PBSD) implies design,

evaluation, and construction of engineered facilities whose

performance under common and extreme loads responds to the

diverse needs and objectives of owners, tenants and societies at

large. The objective of PBSD is to produce structures with

predictable seismic performance. In PBSD multiple levels of

earthquake and corresponding expected performance criteria are

specified [ATC 40 (1996)]. This aspect emphasizes nonlinear

analyses for seismic design verification of any structure. This

procedure gives some guidelines for estimating the possible local

and global damages of structures. A retrofitted structure can be

evaluated with the help of PBSD. Similarly, economics in the form

of life-cycle cost along with construction cost of the structure is

inherently included in PBSD [Prakash (2004)].

On the basis of the aforesaid discussion, an effort has been made in

this paper to develop a comprehensive procedure for the design of

coupled shear walls.

2. INVESTIGATION OF COUPLING BEAM

Coupled shear walls consist of two shear walls connected

intermittently by beams along the height. The behavior of coupled

shear walls is mainly governed by the coupling beams.

The coupling beams are designed for ductile inelastic behavior in

order to dissipate energy. The base of the shear walls may be

designed for elastic or ductile inelastic behavior.

The amount of energy dissipation depends on the yield moment

capacity and plastic rotation capacity of the coupling beams. If the

yield moment capacity is too high, then the coupling beams will

undergo only limited rotations and dissipate little energy. On the

other hand, if the yield moment capacity is too low, then the

coupling beams may undergo rotations much larger than their

plastic rotation capacities. Therefore, the coupling beams should be

provided with an optimum level of yield moment capacities. These

moment capacities depend on the plastic rotation capacity available

in beams. An analytical model of coupling beam has been developed

to calculate the rotations of coupling beam with diagonal

reinforcement and truss reinforcement.

2.1 Results & Discussion

The literatures [Paulay 2002; Hindi and Sexsmith 2001; FEMA356

2000; Xuan et al. 2008; ATC 40 1996; FEMA 273 1997; FEMA 356

2000; Munshi & Ghosh 2000; Galano & Vignoli 2000 and Englekirk

2003] and the results obtained from the ATENA2D (2006) software

package show the inconsistent modeling parameters and

inconsistent evaluative parameters. Therefore, a new model has

been created with some assumptions in the following manner to

carry out further study.


33

Assumptions:

• The effect of gravity loads on the coupling beams has been

neglected.

• Deflection of the coupling beam occurs due to lateral loading.

• Contra flexure occurs at the mid-span of the coupling beam.

• The confined concrete, due to the confining action is

provided by closely spaced transverse reinforcement in

concrete, is assumed to govern the strength.

Total elongation in the horizontal direction (Figure 1) due to lateral

loading can be written as:

bbb dL θ×=∆ (1)

and strain in the concrete, b

bc L

L∆=ε (2)

Lb

db

bbd θ×

2

bbd θ×

2

Figure 1: Schematic diagram of coupling beam

Hence, considering Equations (1) and (2) the following equation can

be written as:

coupling beam rotation, b

bcb d

L×= εθ (3)

The results, considering Equation (3) with maximum strain in

confined concrete ( cuε ) of 0.02 [Confining action is provided by

closely spaced transverse reinforcement in concrete as per ATC 40

(1996)], have been tabulated in Table 1.

Table 1: Maximum rotations in radians

Type of

Reinforce

ment b

b

d

L

Value as

per

Equation

(3)

Galano

and

Vignoli

(2000)

Englek

irk

(2003)

ATC40 (1996),

FEMA273

(1997) and

FEMA356

(2000)

Diagonal

<

1.5

< 0.03

0.062 0.04 0.03

Truss

1.5

to

4.0

0.03 to

0.08

0.084

0.06

-

It can be observed from Table 1 that the values obtained as per

Equation (3) have similar trend with the values specified by ATC 40

(1996), FEMA 273 (1997), FEMA 356 (2000), Galano and Vignoli

(2000) & Englekirk (2003).

Based on the above study, Table 2 has been prepared containing

modified parameters governing the coupling beam characteristics,

which are also considered for developments of the design technique

discussed below. As design technique is based on collapse

prevention (CP) level of structure, plastic rotation capacity given in

Table 2 is for CP level only.

Table 2: Modified Parameters governing the coupling beam

characteristics controlled by shear

Type of She b

d

L

Type of detailing Plastic Rotation


35

coupling

beam

ar

Spa

n to

Dept

h

Rati

o

Capacity (Radians)

'

cw fdb

Shear CP

Reinforc

ed

concrete

coupling

beam

2≤φ

No

limit

Conventional

longitudinal

reinforcement with

conforming

transverse

reinforcement

3≤ 0.015

6≥ 0.010

< 1.5

Diagonal

Reinforcement

(strength is an

overriding

consideration and

thickness of wall

should be greater

than 406.4 mm)

- < 0.03

1.5 to

4.0

Truss

Reinforcement(addi

tional

experimentation is

required)

- 0.03-

0.08

Steel

coupling

beam sp

p

V

Me

6.1≤

Shear dominant - bL

15.0

3. PROPOSED DESIGN TECHNIQUE

In this paper an attempt has been made to develop a technique to

design coupled shear walls considering its ideal seismic behavior

(stable hysteresis with high earthquake energy dissipation). For

preparing this design technique, symmetrical coupled shear walls

have been considered. Design/capacity curve of coupled shear walls

is obtained at the collapse mechanism of the structure based on

this technique. This technique is applied to both fixed base and

pinned base coupled shear walls. To start with, this technique is

useful in selecting the preliminary dimensions of symmetrical

coupled shear walls and subsequently arrives at a final design

stage. Further, this technique is particularly useful for designer,

consultant and practicing engineer who have no access to

sophisticated software packages. A case study has been done

implementing the technique with the help of Microsoft Excel

Spreadsheet and the results have also been validated.

3.1 Proposed Formulation

In Figure 2, the coupled shear walls are subjected to a triangular

variation of loading with amplitude F1 at the roof level. The value

of F1 is obtained corresponding to the CP level of structure.

Subsequently, the base shear and roof displacement can be

determined. The procedure involving Figure 2, the assumptions,

steps and mathematical calculation with initial value of F1 as unity

has been illustrated as follows.

Figure 2 (a): Coupled shear walls Figure 2 (b): Free body diagram

of coupled shear walls

3.2 Assumptions


37

The following assumptions are adopted for the design technique to

obtain the ideal seismic behavior of coupled shear walls.

1. The analytical model of coupled shear walls is taken as two-

dimensional entity.

2. Coupled shear walls exhibit flexural behavior.

3. Coupling beams carry axial forces, shear forces and moments.

4. The axial deformation of the coupling beam is neglected.

5. The effect of gravity loads on the coupling beams is neglected.

6. The horizontal displacement at each point of wall 1 is equal to

the horizontal displacement at each corresponding point of wall

2 due to the presence of coupling beam.

7. The curvatures of the two walls are same at any level.

8. The point of contra flexure occurs at midpoint of clear span of

the beam.

9. The seismic design philosophy requires formation of plastic

hinges at the ends of the coupling beams. All coupling beams

are typically designed identically with identical plastic moment

capacities. Being lightly loaded under gravity loads they will

carry equal shear forces before a collapse mechanism is formed.

All coupling beams are, therefore, assumed to carry equal shear

forces.

10. In the collapse mechanism for coupled shear walls, plastic

hinges are assumed to form at the base of the wall and at the

two ends of each coupling beam. In the wall the elastic

displacements shall be small in comparison to the

displacements due to rotation at the base of the wall. If the

elastic displacements in the wall are considered negligible then a

triangular displaced shape occurs. This is assumed to be the

distribution displacement/velocity/acceleration along the

height. The acceleration times the mass/weight at any floor level

gives the lateral load. Hence, the distribution of the lateral

loading is assumed as a triangular variation, which conforms to

the first mode shape pattern.

Steps

The following iterative steps are developed in this paper for the

design of coupled shear walls.

1) Selection of a particular type of coupling beam and determining

its shear capacity.

2) Determining the fractions of total lateral loading subjected on

wall 1 and wall 2.

3) Determining shear forces developed in coupling beams for

different base conditions.

4) Determining wall rotations in each storey.

5) Checking for occurrence of plastic hinges at the base of the

walls when base is fixed. For walls pinned at the base this check

is not required.

6) Calculating coupling beam rotation in each storey.

7) Checking whether coupling beam rotation lies at collapse

prevention level.

8) Calculating base shear and roof displacement.

9) Modifying the value of F1 for next iteration starting from Step (2)

if Step (7) is not satisfied.

3.3 Mathematical Calculation

The steps which are described in above have been illustrated in this

section as follows:

Step 1

The type of coupling beam can be determined as per Table 2 and

shear capacity can be calculated from Englekirk (2003).

Step 2

In Figure 2(b), free body diagram of coupled shear walls has been

shown; α and β are fractions of total lateral loading incident on wall

1 and wall 2, respectively, such that,

α+β =1.0 (4)

For symmetrical coupled shear walls, moments of inertias of two

walls are equal for equal depths and thicknesses at any level.


39

Further, curvatures of two walls are equal at any level. Hence based

on the Assumption (7), Equation (4) can be written as:

α = β = 0.5 (5)

Step 3

In this step, it is explained how to calculate the shear force

developed in the coupling beams for different types of boundary

conditions. CSA (1994) and Chaallal et al. (1996) defined the degree

of coupling which is written as,

otM

lTDC

×= (6)

where, bw LLl += ; T is the axial force due to lateral loading; Mot is

total overturning moment at the base of the wall produced due to

lateral loading. For fixed base condition DC varies from 0 to 1 and

Equation (6) can also be written as:

( )( ) ( )cb

b

w

a

b

LL

dkDC

×′= (6a)

The above Equation (6a) is proposed by Chaallal et al. (1996); N is

the total number of storeys, k′ is constant and a, b and c are

exponents which are given in Table 3.

Table 3: Values of constant k′ and exponents a, b and c

N k′ a b c

6 2.976 0.706 0.615 0.698

10 2.342 0.512 0.462 0.509

15 1.697 0.352 0.345 0.279

20 1.463 0.265 0.281 0.190

30 1.293 0.193 0.223 0.106

40 1.190 0.145 0.155 0.059

So based upon the above criteria and considering Equations (6) and

(6a), shear force developed in the coupling beam could be

determined as follows:

Fixed base condition:

For fixed base condition following equation can be written as:

( )( ) ( )cb

b

w

a

botN

i

iLL

dk

l

MVTC

×′×=== ∑

=1

(7)

where, Mot is total overturning moment at the base due to the lateral

loading.

Therefore, based on the Assumption (9) shear force in coupling

beam at each storey is,

N

V

V

N

i

i∑== 1 (8)

Pinned base condition:

In this study, pinned base condition has been introduced as one of

the possible boundary conditions for coupled shear walls. It can be

constructed by designing the foundation for axial load and shear

force without considering bending moment. It is expected that

stable hysteresis with high earthquake energy dissipation can be

obtained for considering this kind of base condition.

DC is 1 for pinned base condition from the equation (6). Hence, the

equation can be written as:

l

MVTC ot

N

i

i === ∑=1

(9)

Therefore, based on the Assumption (9) shear force in coupling

beam at each storey is,

N

V

V

N

i

i∑== 1 (10)

Step 4

After obtaining α, β and V at each storey for the particular value of

F1, bending moment values in each storey could be determined for

each wall. Subsequently, curvature diagram for each wall is

generated by using moment area method as adopted in the

Microsoft excel spreadsheet; which is required to determine the wall

rotation in each storey. The following equations are considered to

calculate the wall rotation.


41

Overturning moment at a distance ‘x’ from base with respect to

each wall can be written as:

( ) { ( )( ) }∑−

=

−−−×=iN

j

ssot jhxHjhHH

FxM

0

15.0 (11)

where, i is storey number and it is considered from the base as 0, 1,

2, 3… N.

Resisting moment in wall due to shear force in the coupling beam at

a distance ‘x’ from base can be written as:

( ) ∑=

+=

N

ij

jbw

wr VLL

xM22

(12)

where, net moment in the wall at a distance ‘x’ from base, generated

due to overturning moment and moment due to shear force in the

coupling beam, can be written as:

Mnet(x) = Mot(x) – ( )xM wr (13)

Wall rotation at i th storey for fixed base can be written as:

( )

IE

dxxM

c

ih

net

wi

s

∫= 0θ (14)

where, 12

3

ww LtI

×= (15)

For plastic hinge rotation at the fixed base of wall or rotation at the

pinned base of wall, Equation (18) could be written as:

( )0

0

w

c

ih

net

wiIE

dxxMs

θθ +=∫

(16)

where, 0wθ is the plastic hinge rotation at the fixed base of wall or

rotation at the pinned base of wall.

Step 5

i. Tensile forces at the base of wall 1 (T) as well as compressive

forces at the base of wall 2 (C) are calculated due to lateral

loading.

ii. Compressive loads at the bases of wall 1 and wall 2 are

calculated due to gravity loading.

iii. Net axial forces at the bases of wall 1 and wall 2 are calculated,

i.e.

Net axial force = Tensile or Compressive force due to lateral

loading (T or C)± Compressive load due to gravity loading.

iv. Then, according to these net axial forces for the particular

values of fck, bb, d and p, the yield moment values at the bases of

wall 1 and wall 2 can be determined from

P-M interaction curve [IS 456 (1978) & Jain (1999)]. Where fck,

bb, d and p are yield strength of concrete, breadth of a section,

depth of that section and percentage of minimum reinforcement

in that particular section, respectively; and P is the axial force

and M is the moment; here net axial force is considered as P in

the P-M interaction curve.

v. Therefore, if calculated bending moment value at any base of the

two walls is greater than yield moment value, plastic hinge at

that base would be formed, otherwise no plastic hinge would be

formed.

Step 6

The rotation of coupling beam in each storey is determined as

follows:

Rotation of coupling beam at i th storey for symmetrical walls

[Englekirk (2003)] as per Figure 3 is given by

+=

b

w

wibiL

L1θθ (17)


43

where, wiθ is rotation of wall at i th storey and can be calculated as

per Equation (14), wL = depth of wall, bL = length of coupling beam.

For plastic hinge rotation at the fixed base of wall or real hinge

rotation at the pinned base of wall, Equation (17) could be written

as:

{ }wiwbbi L θθ = (18)

where, wiθ can be calculated as per Equation (16) for fixed base of

all or for pinned base of wall and

wbL =

+

b

w

L

L1 (19)

Step 7

The rotational limit for collapse prevention level of different types of

RCC coupling beams and steel beams are given in Table 2. The task

Lw Lw Lb

2

bL

wiθ

wiθ

biθ

Figure 3: Deformed shape of a i th storey symmetrical coupled shear

was to check whether the rotations of beams attained their

rotational limit of CP level at the collapse mechanism of the

structure simultaneously.

Step 8

The roof displacement can be calculated as per the following

equations:

×= ∑

=

N

i

wisroof h0

θ∆ (20)

where, displacement at i th storey can be calculated as:

×= ∑

=

i

j

wjsi h0

θ∆ (21)

The base shear can be calculated as follows:

( )2

11 +×=

NFVB (22)

Step 9

The F1 is modified as follows when the condition of Step 7 is not

satisfied:

To obtain the collapse mechanism of the structure, it is required to

increase F1 with equal increment until all coupling beams attain

their rotation limit of CP level simultaneously.

3.4 Validation of the Proposed Design Technique

The following numerical example has been considered to validate

the propose design technique. In this study plan and elevation with

dimensions and material properties of the coupled shear walls have

been adopted as given in Chaallal et al. (1996).

3.5 Numerical Example

The coupled shear walls considered here is part of a 20-storey

office building (Figure 4). It is subjected to triangular variation of

lateral loading. The dimension and material properties are

tabulated in Table 4. Dead loads and live loads are discussed in

the following section. A comparison of the results regarding

design/capacity curve (Figure 6) and ductility [Equation (27)]


45

obtained from the proposed design technique with the results

obtained in SAP V 10.0.5 (2000) and DRAIN-3DX (1993) software

packages may, thus, be required. For obtaining more perfection

about the results, these two softwares have been considered here

simultaneously.

Table 4: Dimensions and material properties of coupled shear walls for

validation of proposed design technique

Depth of the wall

(Lw) 4 m

Width of coupling

beam (bb) 300 mm

Length of coupling

beam (Lb) 1.8 m Storey height (hs) 3.0 m

Depth of coupling

beam (db) 600 mm

Modulus of concrete

(Ec)

27.0

GPa

Modulus of steel (Es) 200.0

GPa

Number of storeys

(N) 20

Steel yield strength (fy) 415 MPa

Wall thickness (tw) 300 mm

Figure 5(a) and Figure 5(b) show the plan and sectional elevation of

the coupled shear wall building, respectively.

Figure 4(a): Plan view of building Figure 4(b): Coupled shear

walls at section ‘a-a’

3.5.1 Loading Consideration

Dead loads (DL) of 6.7 kN/m2 and live loads (LL) of 2.4 kN/m2 have

been considered as suggested in Chaallal et al. (1996). Total gravity

loading on coupled shear walls at section ‘a-a’ has been calculated

as the sum of dead load plus 25 % LL as per

IS 1893 (part 1) (2002) for floor; however, in case of roof only dead

load is considered.

3.5.2 Modeling of Coupled Shear Walls in Proposed Design

Technique

The modeling of coupled shear walls involving Figure 2,

assumptions and steps with mathematical calculation is already

described in Section “Proposed Formulation”.

3.5.3 Modeling of coupled shear walls in SAP and DRAIN-3DX

Wide column frame analogy [Mcleod (1966)] has been considered for

modeling of coupled shear walls in SAP V 10.0.5 (2000) and DRAIN-

3DX (1993) as given in Figure 5. In this analogy, shear walls are

represented as two line elements (centre line of shear wall) and

beams are represented as line elements (centre line of beam) by

joining with each other with rigid link. Beam column elastic

5m

5m

9m 9m 9m 9m

Lw

Lw

Lb

a

a

H

Lb

Wall 2

hs

I, A

I, A

Wall 1

db

LW LW


47

element (Type-17) and inelastic element (Type-15) are considered

for modeling.

Figure 5: Modeling in SAP V 10.0.5 (2000) and DRAIN-3DX (1993)

3.5.4 Calculation of Ductility

The obtained design/capacity curve from the proposed design

technique, SAP V 10.0.5 (2000) and DRAIN-3DX (1993) is

bilinearized. The bilinear representation is prepared in the following

manner based on the concepts given in ATC 40 (1996).

Coupling beam

Rigid link

0.5Lw Lb 0.5Lw

Capacity

Figure 6: Bilinear Representation for Capacity Curve

It can be seen from Figure 6 that bilinear representation can be due

to the basis of initial tangent stiffness and equal energies (Area a1 =

Area a2). Subsequently, ductility of the coupled shear walls has

been calculated as:

yieldroof

CProof

,

,

∆

∆µ∆ = (23)

where, CProof ,∆ and yieldroof ,∆ can be calculated from the Equation

(20); ∆µ is the ductility which represents how much earthquake

energy dissipates during an earthquake.

3.5.5 Results and Discussions

Coupled shear walls at section ‘a-a’ as shown in Figure 4 are

considered for conducting the study.

4. RCC coupling beam with Conventional longitudinal

reinforcement and conforming transverse reinforcement:

RCC coupling beam with Conventional longitudinal reinforcement

and conforming transverse reinforcement in each storey has been

selected as per Step 1 for the study. The results of this study for

fixed base as well as pinned base conditions have been shown in

the following manner.

15

10

SAP V 10.0.5

DRAIN-

3DX DESIGN

SAP V 10.0.5

DRAIN-

3DX DESIGN


49

Figure 7(a): Figure 7(b):

Capacity curve for fixed base condition Capacity curve for pinned

base condition

Table 5: Ductility of coupled shear walls considering different approaches

Method Ductility

Fixed base Pinned base

Proposed Design Technique 7 7.5

DRAIN-3DX (1993) 6.75 7.45

SAP V 10.0.5 (2000) 6.92 7.47

4.1 Discussions of numerical results

Figure 7(b) shows that the results obtained from proposed design

technique for pinned base conditions are almost similar with the

results obtained from DRAIN-3DX (1993) and SAP V 10.0.5 (2000).

Whereas, Figure 7(a) is showing a bit differences about the results

obtained from proposed design technique, DRAIN-3DX (1993) and

SAP V 10.0.5 (2000) although same dimensions, same material

properties and same loading were considered in all the three

techniques. However, the differences were not very high (5-10%).

Table 5 is showing the results about ductility obtained for fixed and

pinned base conditions with the help of the Figures 7(a) and 7(b)

and Section “Calculation of Ductility”. It is noticed that ductility for

pinned base condition is greater than fixed base conditions. It

means that stable hysteresis with high earthquake energy

dissipation can be obtained for coupled shear walls with pinned

base.

The results obtained from the proposed design technique are

satisfactory. However, it is necessary to find the limitations of the

proposed design technique. Therefore, in the following section,

parametric study is elaborately discussed to detect the limitations

of the proposed design technique.

4.2 PARAMETRIC STUDY

It has been observed from the CSA (1994) and Chaallal et al. (1996)

that the behavior of the ductile coupled shear walls depends on

degree of coupling, where degree of coupling depends upon depth

and length of the coupling beam as well as depth and height of the

coupled shear walls [Park and Paulay (1975) & Paulay and

Priestley (1992)].

Therefore, this study has been restricted on length of the coupling

beam and number of stories as basic variables and other

parameters are considered as constant. These parameters have

been considered in proposed method to make out effect on the

behavior of coupled shear walls. Further, modifications to achieve

ideal seismic behavior according to the proposed method have been

included in this study.

4.2.1 Model for Parametric Study

A typical building with symmetrical coupled shear walls is shown in

Figures 4(a) and (b). Coupled shear walls at section ‘a-a’ have been

considered to carry out the parametric study.

Loading Consideration


51

Loading has been considered as described as above.

4.2.2 Parameters

Table 6 mentions the different parameters with dimensions and

material properties which have been considered to carry out the

parametric study.

Table 6: Dimensions and material properties of coupled shear

walls for parametric study

Depth of the

wall (Lw) 4 m

Width of coupling

beam (bb)

300

mm

Length of

beam (Lb)

1 m, 1.5 m and 2

m Storey height (hs) 3.6 m

Depth of

beam (db) 800 mm

Modulus of concrete

(Ec)

22.4

GPa

Number of

stories (N) 10, 15 and 20

Yield strength of steel

(fy)

415

MPa Wall

thickness (tw) 300 mm

4.2.4 Discussions of the numerical results

From the above studies, the following discussions have been made

for the design of coupled shear walls under earthquake motion.

(i) Coupled shear walls with N ≥ 15 with equal storey

height m6.3h s = can be designed with an optimum ratio of

25.0L

L

w

b = for 25.1d

L

b

b = and 03

108−×=

I

I b to obtain

consistency between the behavior with respect to the wall

rotation and earthquake energy dissipation.

(ii) Pinned base condition can be provided at the base of the

shear wall as this type of base condition offers better

nonlinear behavior in compare to the fixed base condition.

5. ASSESSMENT OF PROPOSED DESIGN TECHNIQUE USING

NONLINEAR STATIC ANALYSIS

In this paper, nonlinear static analysis is carried out to determine

the response reduction factors of coupled shear walls at different

earthquake levels and through this analysis the proposed design

technique was more justified.

5.1 Design Example

The following design example is presented for carrying out the non

linear static analysis of coupled shear walls. These walls are

subjected to triangular variation of lateral loading. The base of the

walls is assumed as fixed. Table 7 mentions the different

parameters with dimensions and material properties which have

been considered to carry out the study.

Table 7: Dimensions and material properties of coupled shear walls for

nonlinear static analysis

Depth of the wall

(Lw) 4 m

Width of coupling

beam (bb) 300 mm

Length of beam (Lb) 1 m Storey height (hs) 3.6 m

Depth of beam (db) 800 mm

Modulus of concrete

(Ec)

22.4

GPa

Modulus of steel (Es) 200.0

GPa

Number of stories

(N) 20 and 15

Steel yield strength (fy) 415

MPa Wall thickness (tw) 300 mm


53

Figure 4(a) and Figure 4(b) show the plan and sectional elevation of

the coupled shear wall building, respectively. The place considered

for this study is Roorkee and soil type for this place is medium

(Type II); maximum considered earthquake (MCE) level and design

basis earthquake level (DBE) are considered for the study. Loading

has been considered as described as above.

5.2 Discussions of the numerical results

From the Table 8, response reduction factor of coupled shear walls

is varying between 05.2to22.1 for maximum considered earthquake

(MCE) level; which is almost same as the provision of CSA (1994) for

coupling beam with conventional reinforcement.

Table 8: Response Reduction Factors for DBE and MCE levels

Parameters 1e∆µ [Pore

(2007)]

2e∆µ [Pore

(2007)]

µξR [Pore

(2007)]

IDRSRµ

[First

Method of

Energy-

Ductility

Based

Response

Reduction]

[Pore

(2007)]

IDRSRµ

[Second

Method of

Energy-

Ductility

Based

Response

Reduction]

[Pore

(2007)]

dR as per

CSA (1994)

N=20

DBE 1.04 1.004 1.02 1.04 1.004

1.5 or 2 for

coupled

shear walls

with

conventional

reinforced

coupling

beam

MCE 2.05 1.2 1.58 2.05 1.34

N=15 DBE 1.01 1.00 1.002 1.01 1.00

MCE 1.87 1.13 1.39 1.87 1.22

6. CONCLUSIONS

From the above studies, the following conclusions have been made

for the design of coupled shear walls under earthquake motion.

(i) Design technique should be adopted for fixing the dimensions

of coupled shear walls.

(ii) Coupled shear walls with N ≥ 15 with equal storey

height m6.3h s = can be designed with an optimum ratio of

25.0L

L

w

b = for 25.1d

L

b

b = and 03

108−×=

I

I b to obtain

consistency between the behavior with respect to the wall

rotation and earthquake energy dissipation.

(iii) Pinned base condition can be provided at the base of the

shear wall as this type of base condition offers better

nonlinear behavior in compare to the fixed base condition.

(iv) The behavior of coupling beam should be governed by shear.

REFERENCES

[1] Applied Technology Council: ATC-40 Report (1996): Seismic

Evaluation and Retrofit of Concrete Buildings, Volume I, Redwood City,

California.

[2] ATENA2D: Version 3.3.0.3 (2006), Nonlinear Finite Element Integrated

Analysis, Cervenka Consulting, Praha, Czech Republic.

[3] Bureau of Indian Standards: IS-456 (2000), Plain and Reinforced

Concrete – Code of Practice, New Delhi, India.

[4] Bureau of Indian Standards: IS-4326 (1993), Earthquake Resistant

Design and Construction of Buildings - Code of Practice, New Delhi,

India.

[5] Bureau of Indian Standards: IS-13920 (1993), Ductile Detailing of

Reinforced Concrete Structures Subjected to Seismic Forces – Code of

Practice, New Delhi, India.

[6] Bureau of Indian Standards: IS-1893, part 1 (2002), “Criteria for

Earthquake Resistant Design of Structures: Part 1 General provisions

and Buildings”, New Delhi, India.


55

[7] Canadian Standards Association: CAN3-A23.3-M94, CSA (1994),

“Design of Concrete Structures for Buildings”, Rexdale, Ontario,

Canada.

[8] Chaallal, O., Gauthier, D., and Malenfant, P. (1996), “Classification

methodology for coupled shear walls”, Journal of Structural

Engineering, ASCE, 122(12), 1453-1458.

[9] Chao, S. -H., Khandelwal, K., and El-Tawil, S. (2006). “Ductile

fracture initiation in shear link webs”, Journal of Structural

Engineering, ASCE, 132(8), 1192–1200.

[10] El-Tawil, S., Harries, K.A., Fortney, P.J., Shahrooz, B. M. and

Kurama, Y. (2010), “Seismic Design of Hybrid Coupled Wall Systems:

State of the Art”, Journal of Structural Engineering, ASCE, 122(12),

1453-1458.

[11] Englekirk, R.E. (2003), Seismic Design of Reinforced and Precast

Concrete Buildings, John Wiley, NY.

[12] Federal Emergency Management Agency: FEMA-273 (1997),

NEHRP Guidelines for the Seismic Rehabilitation of Buildings,

Washington, DC, U.S.A.

[13] Federal Emergency Management Agency: FEMA-356 (2000),

Prestandard and Commentary for the Seismic Rehabilitation of

Buildings, Washington, DC, U.S.A.

[14] Fortney, P. J., and Shahrooz, B. M. (2009). “Boundary detailing of

coupled core wall system wall piers”, Journal in Advances in Structural

Engineering, 12(3), 299–310.

[15] Galano, L., and Vignoli, A. (2000), “Seismic Behavior of Short

Coupling Beams with Different Reinforcement Layouts”, ACI Structural

Journal, 97(6), 876-885.

[16] Harries, K.A., Mitchell, D., Cook, W.D., and Redwood, R.G. (1993),

“Seismic Response of Steel Beams Coupling Concrete Walls”, Journal

of Structural Engineering, ASCE, 119(12), 3611-3629.

[17] Harries, K. A., and McNeice, D. S. (2006), “Performance-based

design of high-rise coupled wall systems.” The Structural Design of Tall

and Special Structures, 15(3), 289–306.

[18] Hindi, A., and Sexsmith, R. (2001), “A Proposed Damage Model for

R/C Bridge Columns under Cyclic Loading”, Earthquake Spectra, 17

(2), 261–281.

[19] http://www.nicee.org/eqe-iitk/uploads/EQR_Bhuj.pdf

[20] Jain, A.K. (1999), Reinforced Concrete Limit State Design, Nem

Chand & Bros, Roorkee.

[21] Munshi, J.A., and Ghosh, S.K. (2000), “Displacement-Based Seismic

Design for Coupled Wall Systems”, Earthquake Spectra, 16(3), 621-642.

[22] New Zealand Standard: NZS 3101 (part 1) (1995). "The Design of

Concrete Structures" Wellington, NZ.

[23] Park, R., and Paulay, T. (1975), Reinforced Concrete Structures,

John Wiley & Sons, Inc., NY.

[24] Paulay, T. (1986), “The Design of Ductile Reinforced Concrete

Structural Walls for Earthquake Resistance”, Earthquake Spectra,

2(4).

[25] Paulay, T. and Priestley, M.J.N. (1992), Seismic Design of

Reinforced Concrete and Masonry Buildings, John Wiley & Sons, Inc.,

NY.

[26] Paulay, T. (2002), “A Displacement-Focused Seismic Design of

Mixed Building System”, Earthquake Spectra, 18 (4), 689–718.

[27] Paulay, T. (2002), “The displacement capacity of reinforced

concrete coupled walls”, Engineering Structures, 24, 1165–1175.

[28] Penelis, G.G., and Kappos, A.J. (1997), Earthquake-resistant

concrete structures, E&FN SPON, NY.

[29] Pore, S.M. (2007), Performance Based Seismic Design of Low to

Medium Rise RC Framed Buildings for India, Department of

Earthquake Engineering, IIT Roorkee.

[30] Prakash, V., Powell, G.H. and Campbell, S. (1993), “DRAIN-3DX

Base Program User Guide Version 1.10”, Structural Engineering,

Mechanics and Materials, Department of Civil Engineering UC,

Berkeley, California, USA.

[31] Prakash, V. (2004), “Whither Performance-Based Engineering in

India?”, ISET Journal, 41(1), 201-222.

[32] SAP2000: Advanced 10.0.5 (2006), Static and Dynamic Finite

Element Analysis of Structures, Computers and Structures Inc.,

Berkeley, CA.


59

Effect of Constituent-Characteristics on Durability of

Masonry and Concrete Structure

V. Kumar

Professor, Department of Civil Engineering, Institute of Technology,

Banaras Hindu University, Varanasi. India

1. INTRODUCTION

The most effective use of masonry construction is seen in load

bearing structures wherein it performs a variety of functions,

namely, supporting loads, subdividing space, providing thermal and

acoustic insulation, as well as fire and weather protection, which

normally in a framed building has to be accounted for separately. In

India, there has not been much progress in the construction of tall

load bearing masonry structures, mainly because of poor quality of

masonry workmanship and materials such as clay bricks that are

manufactured even today having nominal strength of only 7 to

10MPa. However, recently mechanized brick plants are producing

brick units of strength 17.5 to 25N/mm2 and therefore it is possible

to construct 5 to 6 storied load bearing structures at costs less than

those of RC framed structures.

The appearance of a finished confined masonry construction and a

RC frame construction with masonry in fills may look alike to lay-

man; however, these two construction systems are substantially

different. In confined masonry construction, confining elements are

not designed to act as a moment- resisting frame; as a result,

detailing of reinforcement is simple. In general, confining elements

have smaller cross sectional dimensions than the corresponding

beams and columns in a RC frame building. It should be noted that

the most important difference between the confined masonry walls

and infill walls is that infill walls are not load-bearing walls, while

the walls in a confined masonry building are load bearings.

There can be no denying the fact that concrete has virtually

dominated the field of construction in the 20th century. Due to its

versatile and numerous well known advantages, no cost effective

substitute has emerged for this 150 year old material. Therefore, it

is no wonder that most of the experts believe that concrete’s

distinction of being the largest man made material of construction

in the world is not likely to be challenged in the near future.

However, the concrete design and construction practices today are

essentially strength driven. Due to escalation in the repair and

replacement costs of structures and a growing concern about

sustainability of the concrete industry more attention is being paid

now to durability issues. In homogeneities in the micro-structure of

concrete are responsible for micro-cracks which grows into macro-

cracks due to weathering resulting into fast rate of transport of

water, carrying harmful ions and gases from the surface into the

interior of concrete and hence deterioration of concrete. The effect of

different ingredient on the durability of concrete has been covered

to show the changes in concrete technology needed for enhancing

the durability of structures.

The technology of concrete has traversed a long way from normal

strength concrete to high – strength, high performance to ultra high

performance concrete branching into a variety of innovative

developments in the fields of fiber reinforced concrete, polymer –

modified concrete, self compacting concrete, high volume fly ash

concrete etc.

Concrete is an environment-friendly as compared to the other major

materials of construction such as structural steel, aluminum etc.

The production of concrete involves least amount of energy

consumption, releases no byproducts and it is recyclable. Also it

can be made greener with the incorporation of vast proportion of

waste products from other industries such as fly ash from thermal

power stations, ground granulated blast-furnace slag from steel

industry, silica fume from silicon and ferro-silicon smelting plants

and rice husk ash improving the environmental profile and thus

sustainability of concrete on one hand and its durability on the

other hand.


61

During the past few decades, the phenomenon of premature

deterioration of concrete structure is being witnessed. This has

become a matter of concern in many countries bringing the issue of

durability of concrete in the fore front. Also the codes of practice of

many countries, including Indian code IS 456 -2000 have

undergone changes incorporating revised provisions pertaining to

durability of concrete. In this context an attempt has been made to

highlight the properties and sincere use of different constituent of

concrete on the durability of concrete.

Prof. P. K. Mehta, a concrete technologist has remarked, “In spite of

an accumulated knowledge base on how to build durable concrete

structures, there has been essentially no progress on the issue”. He

has also suggested that by a judicious selection of concrete making

materials and mix proportions, and by proper construction practice,

the micro-structural inhomogeneities in concrete can be

considerably reduced and the durability of structures can be

radically enhanced.

2. ROLE OF AGGREGATES IN CONCRETE

Aggregates play a crucial role in ensuring long-term durability of

concrete. This aspect is not adequately understood on many

occasions by persons involved in material selection and mix

proportioning, who tend to consider aggregate as an inert

component used essentially as economic filler and as a means to

render some volume stability to concrete.

Aggregate constitute nearly 70 to 80 percent of volume in concrete

and have profound impact on a variety of properties of concrete

both in the fresh and hardened states. It has a most visible and

significant influence on the properties of fresh concrete such as

workability, segregation, bleeding etc as well as on the hardened

properties, mainly the compressive, flexural, tensile, modulus of

elasticity, shrinkage, creep etc. Also the properties of aggregate

influences the durability properties of concrete such as resistance

to chloride and sulphate attack, carbonation, alkali-aggregate

reaction etc. Many of the properties are related mainly to the binder

phase but aggregates also play a key role in several ways.

3. ROLE OF CEMENT

Cement is one of the most important constituents of concrete and

its physical and chemical characteristics have profound effects on

the properties of concrete. Modern concrete also contains a variety

of admixtures – both mineral and chemical besides cement,

aggregate and water. Due to these additives, the nature of concrete

tends to be more complex and it becomes difficult to predict the

influence of a particular ingredient as the properties of fresh and

hardened concrete. However, Portland cement being one of the most

reactive materials in concrete, has a considerable influence on the

final properties of concrete. From the comparison of data on the

characteristics of commercially available Portland cements during

the last five decades, some interesting trends have been revealed. It

has been found that there has been increase in the C3S content and

decrease in C2S content, although the amount of total calcium

silicate had remained same. Also the increase in the fineness of

cement and SO3 as well as alkali contents have been observed. Due

to the forgoing, modern cements generally gain strength more

rapidly upto 7 days, due to higher C3S content and higher fineness

of the modern cement. Although these changes in the cement

composition and increase in fineness helped in obtaining high early

strength, but the percentage gain in moist cured concrete strength

between 28 days and 10 years has reduced.

Several structures constructed during last 20 years using high

strength concrete in USA are suffering from epidemic of durability

problems. Simultaneously more cases of serious and premature

deterioration of concrete infra-structure have been reported from

around the world. Some of the high strength concrete structures

cracked even before construction was completed. All these were due

to high thermal contraction and autogenous shrinkage resulting

from the use of high cement content and fast hydrating type

cement. Therefore, the concrete which gives high compressive

strength may not be durable, so the emphasis is also given to

permeability and hence durability of concrete. However, it has been

realized that it is difficult to produce durable concrete for


63

infrastructure work without the use of pozzolans. Concrete mixture

with cementitious material including 30 percent fly ash or 40

percent granulated slag has been found to be user friendly,

economical and crack resistant.

Significance of microcracks: continuous microcracks linking into

wider cracks originating from the concrete surface play a wider role

in increasing permeability and hence reducing durability of

concrete.

Hydration reaction of Portland cement minerals produce a

multiphase product that consists primarily of an adhesive poorly

crystalline, C-S-H (Calcium Silicate Hydrate) phase and some well

crystalline products including calcium hydroxide. In freshly mixed

and compacted concrete, water films forming around the coarse

aggregate particles raise the water cement ratio in close proximity

to these particles. In the interfacial transition zone between a

coarse aggregate particle and cement mortar, the space with high

water-cement ratio becomes filled with a porous framework of large

plate-like, oriented, and non-adhesive crystals of calcium

hydroxide. In conventional concrete, this is the weak area which is

highly vulnerable to microcracks. Therefore, reducing the area of

the interfacial transition zone in concrete, and elimination of the

defects and inhomogeneities within the hydrated cement paste

seem to be the proper tools to control microcracks.

In fully hydrated Portland cement pastes, approximately 24 percent

of the hydration product by mass consists of oriented

heterogeneously distributed and weakly bonded layers of calcium

hydroxide crystals serving as potential site for the formation of

micro-cracks. By transforming all or most of the calcium hydroxide

into the calcium silicate hydrate phase which is much more

homogenous hydration product, and stable, the problem of

microcracks in the concrete can be reduced. Therefore, concrete

mixtures with fewer microcracks can be produced by the use of

blended Portland cement containing large proportion of pozzolanic

cementitious materials.

4. ROLE OF WATER IN CONCRETE

Water is one of the most vital ingredients of concrete. Its use in

concrete can broadly be divided in four categories, viz., hydration of

the cementitious matrix, conferring workability to the mix, curing

during hardening process and washing of aggregates.

It has both beneficial and detrimental effects on concrete. It helps

in hydration process, and also in lubricating the concrete mix for

easy handling, transportation, placement and finishing. When used

for curing concrete, it is instrumental in improving the long term

strength gain, durability and many other properties of concrete

provided curing is done for adequate period. Besides this beneficial

role, water also happens to be the key element involved in a number

of deterioration phenomenon. It is a powerful solvent and it has

potential to carry aggressive chemical agents, which may prove

deleterious for hardened concrete. Carbonation of calcium

hydroxide formed in the hydration process cannot take place unless

carbon dioxide forms a weak acid by dissolving and associating with

water. The degradation like corrosion of embedded steel and alkali

aggregate reaction cannot proceed in absence of water or moisture.

Thus, water is both a ‘friend’ and ‘foe’ of concrete.

It is well known that by lowering the w/c, compressive strength and

other properties of concrete can be improved. However, water

demand increases with increase in workability, ambient

temperature and fines content of the cementitious powder. It is

easier to place concrete with higher water content but essential

properties of concrete are harmfully affected. Excess as well as lack

of bleeding of concrete is harmful as in case of higher evaporation

and no bleeding, risk of plastic shrinkage cracks increases and in

case of excess bleeding, the permeability of concrete increases due

to increase in microcracks. For this purpose, the concrete society

documents suggests, “it might be a good idea to consider the use of

an admixture that locks water on position in the fresh concrete

state”.

Also in hardened concrete, the presence of continuous microcracks,

facilitate the ingress of water from external sources, increases the

degree of saturation which is a pre-requisite for any damage to be


65

caused by frost action. Similarly, the penetration of salt water to the

surface of the reinforcement set the stage for corrosion. A

microcracked structure subjected to alternate wetting and drying

are prone to damage due to sulphate attack or due to alkali silica

reaction. Therefore, if the structure remains dry one can almost

have a trouble free reaction even if concrete contains reactive

aggregate and mobile alkalies.

5. CONCLUSIONS

To move towards the goal of sustainable construction industry, one

must achieve a radical enhancement in the durability of Portland

cement concrete which is the most widely used material of

construction in the world today.

The major root cause of concrete durability problems is the

presence of inhomogeneities in the hydrated cement paste. These

inhomogeneities serve as potential sites for microcracks and hence

increase in porosity which is the single parameter that has the

largest influence on durability.

Therefore, for achieving the desired durability, it is vital to select

and specify appropriate constituents of concrete in correct dosages

and giving due considerations to the exposure conditions. Hence, by

proper construction practice, the micro-structural in-

homogeneities in concrete can be considerably reduced and

durability of structures can be radically enhanced.

REFERENCES

[1] Guide to durable concrete, Editorial, ICJ, Vol. 79, no. 10,

Oct. 2005.

[2] Effect of cement characteristics on concrete properties,

Editorial, ICJ, Vol. 79, No. 4, April 2005.

[3] Role of water in concrete, Editorial, ICJ, Vol. 80, No. 3,

March 2005.

[4] How to specify concrete? Editorial, ICJ, Vol. 80, No. 6, June

2006.

[5] Mehta, P. K., Durability of concrete – the zigzag course of

progress, ICJ, Vol. 80, No. 8, August 2006.

[6] Nawy, E. G., Concrete construction engineering handbook,

CRC Press, New York, 1997.

[7] Mehta, P. K. and Monterio, P. J. M, “Concrete

microstructure,, properties and materials”, ICJ, Chennai,

1999.


67

Provisions of Different Codes in Brick Masonry

Buildings: A Critical Review

Rajesh Kumar

Associate Professor, Department of Civil Engineering, Institute of

Technology, Banaras Hindu University, Varanasi, India

1. INTRODUCTION

In India, most of the residential buildings up to four stories are

either constructed with load bearing brick masonry (BM) wall or

reinforced concrete (RC) moment resisting frame type with brick

masonry acting as partition/infill walls. In the current design

practice for reinforced concrete buildings, the infill walls are

considered to be non structural and analysis and design of

buildings are done neglecting the strength and stiffness

contributions of the infill. This leads to an incorrect idealization of

structure. According to Clause 4.4.3.1 of the National Building

Code of India, while designing the structure all the walls of the

structure should be planned to take load, so that it gives maximum

economy. Also according to Structural Masonry Designer’s Manual

for Design of Multi-Storey Structures, the frame in reinforced

concrete construction has to carry loads from the roof and floors,

and has to be strengthened to carry the weight of the walls. This

results in complete wastage of structural potential of the brick or

block masonry used in the walls.

In India, there has not been much progress in the construction of

tall load bearing masonry structures, mainly because of poor

quality of masonry workmanship and materials such as clay bricks

that are manufactured even today having nominal strength of only

7 to 10 MPa. However, recently mechanized brick plants are

producing brick units of strength 17.5 to 25 N/mm2 and therefore it

is possible to construct 5 to 6 storied load bearing structured at

costs less than those RCC framed structures. Use of confined brick

masonry can further improve its load carrying capacity and most

importantly its flexural and shear behavior under earthquake loads.

A construction system where RC members confine a plain masonry

walls on all four sides or reinforced masonry is called confined

masonry (1).

This paper deals with the study of different codes on the design of

masonry structures. The focus has been on the comparative study

of the codes and national building codes of India with respect to the

design philosophy, effect of reinforcement on masonry and design of

masonry under compression, flexure and shear.

2. Masonry Codes

The different codes have been reviewed and presented on the basis

of provisions related to design approach, member sizing and details.

2.1 Building Code Requirements for Masonry Structures (ACI 530-

02/ASCE 5-02/TMS 402-02)

These codes cover the design and construction of masonry

structures and are accompanied with a commentary on the building

code requirements. The code provides minimum requirements for

the structural design and construction of masonry units using both

allowable stress design as well as limit state design for unreinforced

as well as reinforced masonry. In limit state design, more emphasis

is laid on reinforced masonry than unreinforced masonry. A

empirical design method applicable to buildings meeting specific

location and construction criteria is also included.

2.2 Indian Standard Code of Practice for Structural Use of

Unreinforced Masonry (IS: 1905-1987(2))

The Indian Standard on masonry design was first published in

1960 and later on revised in 1969, 1980 and 1987. The current

third version, published in 1987, was reaffirmed in 1998. The

Provisions of this code is very similar to those of BS 5628: Part

1:1978. A separate handbook to this code, SP 20, 1991, is also

available. This Indian Standard provides recommendations for

structural design aspect of load bearing and non -load bearing walls

using unreinforced masonry only. Design procedure adopted

throughout the code is allowable stress design, along with several

empirical formulae. The code refers to IS: 4326 for strengthening


69

unreinforced masonry building for seismic resistance and does not

provide any calculation for the design of reinforcement.

2.3 Earthquake Resistant Design and Construction of Buildings- Code

of Practice (IS: 4326:1993(3))

This standard provides guidance in selection of materials, special

features of design and construction for earthquake resistant

buildings including masonry construction. The general principles to

be observed in the construction of such earthquake resistant

buildings as specified in this standard are lightness, continuity of

construction, avoiding/reinforcing projecting and suspended parts,

building configuration, strength in various directions, stable

foundations, ductility of structure, connection to non-uniform parts

and fire safety of structures. Special construction features like

separation of adjoining structures crumple section, foundation

design, roofs and floors and staircases have been elaborated in the

standard.

As per IS 4326: 1993 Clause 8.5, the load bearing walls can be

made thinner than 200 mm say 150 mm inclusive of plastering on

both sides. Reinforced concrete framing columns and collar beams

will be necessary to be constructed to have full bond with walls.

Columns are to be located at all corners and junction of walls and

spaced not more than 1.5 m but so located as to frame up the doors

and windows (Fig.1). The horizontal bands or ring beams are

located at all floor roof as well as lintel levels of openings. The

sequence of construction between walls and columns will be first to

build the wall up to 4 to 6 courses height leaving toothed gaps

(tooth projections being about 40 mm only) for the columns, and

second to pour M15 concrete to fill the columns against the walls

using wood forms on both sides. The column steel should be

accurately held in position all along. The band concrete should be

cast on the wall masonry directly so as to develop full bond with it

(Fig. 2). Such construction may be limited to only two storeys

maximum in view of its vertical load carrying capacity. The

horizontal length of walls between cross walls shall be restricted to

7 m and the storey height to 3 m.

Figure 1: Typical distribution of vertical confining elements in the plan

of the building


71

Figure 2: Framing of thin load bearing brick walls

.

As per Clause 8.4.2 of IS 4326: 1993 lintel band shall be provided

at lintel level on all load bearing internal, external longitudinal

cross walls. The band shall be made of reinforced concrete of grade

not leaner than M15 or reinforced brick work in cement mortar not

leaner than 1:3. The bands shall be of full width of the wall and not

less than 75 mm in depth and reinforced with steel (Fig. 3).

Figure 3: Reinforcement and Bending Details in reinforced concrete band

As per clause 8.4.1 of IS 4326: 1993, all masonry buildings shall be

strengthened by the methods, as specified for various categories of

buildings, as listed in Table 1.

Table 1: Strengthening methods for various categories of buildings listed

in IS: 4326-1993.

Building Category Number of Storeys Strengthening to be

provided on all Storeys

A ( αh < 0.05) i) 1 to 3 a

ii) 4 a, b, c

B ( 0.05≤ αh ≥ 0.06) i) 1 to 3 a, b, c, f, g

ii) 4 a, b, c, d, f, g

C (0.06 < αh < 0.08) i) 1 and 2 a, b, c, f, g

ii) 3 and 4 a to g

D ( 0.08 ≤ αh < 0.12) i) 1 and 2 a to g

ii) 3 and 4 a to h

E ( αh ≤ 0.12) i)1 to 3* a to h

Where, αh is Design Seismic Coefficient

a- Masonry mortar ( Cl. 8.1.2)

b- Lintel band

c- Roof band and gable band where necessary

d- Vertical steel at corners and junctions of walls

e- Vertical steel at jambs of openings

f- Bracing in plan at tie level of roofs

g- Plinth band where necessary

h- Dowel bars

* fourth storey is not allowed in category E

2.3.1 Shortcomings of IS: 4326-1993

IS: 4326-1993 provides for vertical column steels to tie the building

in the vertical direction. But by simply providing the steel in the

vertical direction confinement of masonry is not achieved, which is

the main reason by which horizontal force resisting capacity of

building increases many folds.

Further, if in place of vertical corner steel composite reinforced

concrete columns are provided then separation of walls at the

corner junction shall also be resisted by the flexural capacity of the

composite reinforced concrete columns.

Clause 8.5 of IS: 4326-1993 stipulate that the size of offsets should

be 40 mm which is too much and this will result in improper filling

of the concrete in offsets, leaving voids. Therefore the size of offset

should be kept as 5mm to 10mm for proper filling and interlocking

of aggregates of concrete.

2.4 Improving Earthquake Resistance of Low Strength Masonry

Buildings-Guidelines (IS: 13828:1993(4))


73

This standard covers the special features of design and

construction for improving earthquake resistance of buildings of

low strength masonry. The provisions of this standard are

applicable in all seismic zones. No special provisions are considered

necessary for buildings in seismic zone II if cement sand mortar not

leaner than 1:6 is used in the masonry. The various provisions of

IS: 4326:1993 regarding general principles, special construction

features, type of construction, categories of buildings and masonry

construction of low strength dealt with in this standard.

2.5 Ductile Detailing of Reinforced Concrete Structures Subjected to

Seismic Forces-Code of Practice (IS: 13920:1993(5))

This standard covers the requirements for designing and detailing

of monolithic reinforced concrete buildings so as to give them

adequate toughness and ductility to resist severe earthquake socks

without collapse.

2.6 Repairs and Seismic Strengthening of Buildings-Guidelines (IS:

13935:1993(6))

The code covers the selection of materials and techniques to be

used for repair and seismic strengthening of damaged buildings

during earthquakes and retrofitting for upgrading of seismic

resistance of existing buildings. The buildings affected by

earthquake may suffer both non-structural and structural

damages. This standard lays down guidelines for non-

structural/architectural as well as structural repairs, seismic

strengthening and seismic retrofitting of existing buildings.

Guidelines have been given for selection of materials for repair work

such as cement, steel, epoxy resins, epoxy mortar, quick setting

cement mortar and special techniques such as shotcrete,

mechanical anchorage etc.

2.7 National Building Code of India –Guidelines (2005(7))

Special features of design and construction for earthquake resistant

masonry buildings are given in National Building Code. For the

purpose of specifying the earthquake resistant features in masonry

buildings, the buildings have been categorized in five categories A to

E based on the seismic zone and the importance of building I (Table

2).

Table 2: Building Categories for Earthquake Resisting Features

Importance Factor Seismic Zone

II III IV V

1.0 A B C D

1.5 B C D E

Mortars, such as those in table 3 or of equivalent specification,

shall preferably be used for masonry construction for various

categories of buildings. Where steel reinforcing bars are provided in

masonry the bars shall be embedded with adequate cover in cement

sand mortar not leaner than 1:3 (minimum clear cover 10 mm) or in

cement concrete grade M15 (minimum clear cover 15 mm or bar

diameter whichever more), so as to acieve good bond and corrosion

resistance.

Table 3: Recommended Mortar mixes

Category of Construction Proportion of Cement-Sand

A M2 (Cement-Sand 1:6)

B, C M2 (Cement-Sand 1:6)

D, E H2 (Cement-Sand 1:4)

Masonry bearing walls built in mortar as specified in table 2 unless

rationally designed as reinforced masonry shall not be built of

greater height than 15 m subject to a maximum of four storeys

when measured from the mean ground level to the roof slab or ridge

level.

2.8 Euro Code (8,9,10)-Guidelines

Normally the tie-columns should fit into the thickness of masonry

wall and the minimum tie-column cross section is 150x150 mm.

The concrete for the confining members should be min grade M15.

According to EC, the contribution of the tie-columns and bond-

beams to the lateral resistance of the masonry house should not be

taken into account. Consequently specific design calculations for

confining elements are not required. The amount of reinforcement

in vertical and horizontal confining elements is determined on


75

empirical basis. The min steel tie-columns reinforcement for

construction in seismic zones is specified in EC. According to this

code the min reinforcement area for tie-columns is 240 mm2. For

tie-columns at the house corners and wall intersections, it is

recommended that, at least 4 x 12 mm dia mild steel bars are used

for reinforcement. In this case the total steel area is 314 mm2. Mild

steel stirrups of 10 mm dia are placed uniformly distributed at 200

mm offsets. Although the tie-columns and bond beams do not

provide frame system adequate splicing and anchoring of rebars is

required at all joints. Sixty rebar diameters splices are required

according to EC. In some resources tabulated data are provided,

where the area or rebars can be selected in dependence of

seismicity of the location and number of storeys in the house. Such

data is presented below in Table 4 for tie-columns.

Table 4: Recommended reinforcement for vertical confining elements

No of

storeys

Low:

ag <

0.1g

Moderate:

0.1g < ag <

0.2g

High:

0.2g < ag < 0.4g

2 1-2 4#8 4#10 4#12

4 1-2 4#8 4#10 4#12

4 2-4 4#8 4#10 4#12

6 1-2 4#10 4#12 4#14

6 3-4 4#8 4#10 4#12

6 5-6 4#8 4#10 4#12

To enforce the confinement of plane masonry by the confining

members EC 8 requires connecting the masonry and tie-columns by

means of rebar diameter #6 min at max 600 mm apart. These links

should be anchored at least 250 mm into the mortar joints. Brick

masonry should be constructed on the basis of following simple

instructions for quality workmanship:

• In dry and hot climate, masonry units should be soaked in

water before the construction in order to prevent quick

drying and shrinkage of cement based mortars.

• Masonry units should be assembled together in overlapped

fashion (Figure 4 and Figure 4a) so that the vertical joints are

staggered from course to course. To ensure adequate

bonding the units should overlap by a lenght equal to 0.4

times the height of unit or 40 mm, whichever is the greater.

• Same type of masonry units and mortar should be used for

structural walls in the same storey.

• Bracing walls should be constructed in the same time as the

load-bearing walls

• The thickness of individual walls is kept constant from storey

to storey.

In cases where general purpose mortar is going to be used, the

mortar joints thickness should be between 8 and 15 mm. EC 8

specifies that, in seismic zones, the load-bearing masonry wall

thickness should be min 240 mm when the masonry is confined.

To ensure stability of walls, the ratio of the effective wall height to

wall thickness should be max 15. To ensure load-bearing capacity

of masonry walls with openings the length of a structural wall

should be at least 1/3 of the greater clear height of the openings

adjacent to the wall in the case of confined masonry.


77

Figure 4: Flemish bond for one brick thick wall

Figure 4a: English bond for one brick thick wall

3. Confined brick masonry action (11)

3.1 Earthquake resistance of masonry walls

In the event of an earthquake, apart from the existing gravity loads,

horizontal racking loads are imposed on walls. However, the

unreinforced masonry behaves as a brittle material. Hence if the

stress state within the wall exceeds masonry strength, brittle failure

occurs, followed by possible collapse of the wall and the building.

Therefore unreinforced masonry walls are vulnerable to

earthquakes, and should be confined and/or reinforced whenever

possible. Masonry walls resisting in-plane loads usually exhibit the

following three modes of failure:

• Sliding shear- a wall with poor shear strength, loaded

predominantly with horizontal forces can exhibit this failure

mechanism. Aspect ratio for such walls is usually 1:1 or less

(1:1.5).

• Shear- a wall loaded with significant vertical load as well as

horizontal forces can fail in shear. This is the most common

mode of failure. Aspect ratio for such walls is usually about

1:1. Shear failure can also occur for panels with bigger

aspect ratio ie. 2:1, in cases of big vertical load.


79

• Bending- this type of failure can occur if walls are with

improved shear resistance. For bigger aspect ratios ie. 2:1

bending failure can occur due to small vertical loads, rather

than high shear resistance. In this mode of failure the

masonry panel can rock like a rigid body (in cases of low

vertical loads).

Failure modes for masonry walls subject to in-plane loads are

shown on Fig. 5

Figure 5: Failure modes for masonry walls subject to in-plane loads

3.2 Mechanical properties

In order to estimate the resistance of masonry walls, the following

mechanical properties for the masonry needs to be determined:

• The compressive strength- f

• The shear strength- fv

• The bending strength- fx

• The stress-strain relationship, s-e

Other essential mechanical characteristics of masonry:

• The tensile strength- ft, as an equivalent to shear strength- fv

• The modulus of elasticity- E

• The shear modulus- G

• The ductility factor-m

The ductility factor is determined only for a specific structural

element (specific proportions, boundary conditions etc). It cannot be

determined for the masonry itself. Mechanical characteristics of

masonry are determined by testing standard specimens of masonry

wallets and walls according to code EC.

3.2.1 Compressive strength

Compressive strength is determined by testing masonry specimens

of at least 1.5 units length and 3 units height or by testing walls of

1.0-1.8 m length and 2.4-2.7 m height.

In cases where the masonry specimen is slender

(height/thickness>20), lateral displacements at the mid height of

the wall are measured. The slenderness can be taken into account

using the measured value for this displacement d and the thickness

of the wall t. Thus the measured compressive strength can be

increased by the following factors:

t/(t-d), provided the increase is not more than 15%. According to

EC three identical specimens are tested and the results evaluated.

In cases where the measured mean compressive strength f of

masonry is different from the one of its constituents ( masonry

units and mortar) by 25% the value of f is modified. The

characteristic compressive strength of masonry fk is determined as

the smaller value of either fk=f/1.2 or fk=fmin. When verifying load

bearing masonry and test data is not available, the characteristic

compressive strength of plain masonry made with general purpose

mortar may be calculated on the basis of normalised compressive

strength of masonry units fb and compressive strength of mortar fm

as follows:

fk = K*(fb0.65)*(fm0.25) [MPa],


81

and fm is less than 20 MPa or 2fb, whichever is the smaller. The

value of constant K depends on the classification of masonry .

Below are shown recommended values for K:

• 0.60 for group 1 masonry units in a wall without longitudinal

mortar joint,

• 0.55 for group 2a masonry units in a wall without

longitudinal mortar joint,

• 0.50 for group 2b masonry units in a wall without

longitudinal mortar joint, and for group 1 masonry units in a

wall with longitudinal mortar joint,

• 0.45 for group 2a masonry units in a wall with longitudinal

mortar joint,

0.40 for group 2b masonry units in a wall with longitudinal mortar

joint, and for group 3 masonry units.

3.2.2 Shear strength

Shear strength of masonry is defined as a combination of initial

shear strength under zero compressive load and increase in

strength due to compressive stresses perpendicular to the shear

plane. Initial shear strength at zero compressive stress is denoted

with fvko. This property is determined according to EN 1052-3 by

testing a triplet specimen such that only shear stresses develop in

the mortar to masonry unit contact planes. A minimum of five

triplets are tested. The minimum acceptable value of fvko is 0.03

MPa. The characteristic shear strength of plain masonry is then

calculated as follows:

fvk = fvko+0.4*sd,

where sd is the design compressive stress perpendicular to the

shear plane. The value of sd should be greater than 0.065fb and a

limiting value specified in EC 6 depending on masonry unit's group

and mortar quality.

3.2.3 Bending strength

In cases where the masonry needs to be verified for out-of-plane

loads the bending strength is the governing factor. The bending

strength parallel to bed joints (see Fig. 6) is denoteed with fx1 and

the bending strength perpendicular to bed joints (see Fig. 7) is

denoted with fx2. According to EC 6 the value of fx1 should be

taken as zero when evaluating seismic resistance.

Figure 6: Vertical orientation of failure plane and corresponding bending

strength normal to bed joints


83

Figure 7: Horizontal orientation of failure plane and corresponding

bending strength parallel to bed joints

3.2.4 Elastic properties

The modulus of elasticity E of masonry can be determined after

compression tests. The elastic modulus is defined as a secant

modulus at service load condition. This load level corresponds to

1/3 of the maximum vertical load. When determined by testing E

modulus value is not available the following equation may be used:

E=1000fk

However in the calculated value of E modulus may not be correct.

Reliable E values are the one in the margin:

200fk <= E <= 2000fk

Theoretically and as specified in EC 6 the G modulus is evaluated

as being 40% of the E modulus. In reality the values of shear

modulus G are much lower. Reliable G values are the one in the

margin:

1000ftk <= G <= 2700ftk

The discrepancy between experimental and predicted values for the

mechanical properties of masonry can be explained with the

composite nature of the material. There are wide variety of not only

masonry units but also mortars and various composition of the

masonry wall itself. Therefore the testing of masonry is essential

step in seismic resistance verification of masonry houses.

4. Planning and layout

Surveys of earthquake damaged residential masonry wall houses

and analysis of the causes of damage indicate that well tied

buildings with well defined, continuous load path to the

foundations perform much better in earthquakes than building

lacking such features. Well defined, continuous load path can be

achieved with regular structural layout and uniformity both in plan

and elevation. The degree of symmetry is also found to have a

significant influence on earthquake resistance. Damage can be five

to ten time worse in irregular buildings compared to regular ones.

Thus satisfactory seismic behavior can be guaranteed by following

the requirements for regular and uniform layout both in plan and

elevation, interconnectivity between structural members and

strength of materials summaries an earthquake resistant structural

form for masonry wall structure is the one which is:

• Regular both in plan and elevation i.e. uniform and

symmetrical

• Redundant - capable of providing adequate resistance even

after a failure of a structural member

• With rigid floors interconnected with walls that ensure

diaphragm action

Stable foundation should be provided able to transmit the

maximum seismic loads from the superstructure to the foundation

soil. Masonry buildings with horizontal irregularities and lack of

symmetry may have considerable eccentricity between the mass

centre and stiffness centre giving rise to damaging coupled

lateral/torsional response. Horizontal irregularities in the form of

extensions, projections etc. may cause stress concentration and

local failures since these extensions are prone to vibrate separately

from the rest of the structure. On the other hand vertical

irregularity in masonry building may cause stress concentration at

a horizontal plane that can lead to total collapse. In order to achieve

satisfactory redundancy at least to lines of load bearing walls are

required in each principal direction of the building. Lack of rigid

floors will prevent proportionate load transfer onto walls at each

floor level as well as will not provide out of plane restraint. Not

supported masonry walls at floor level tends to separate at corners

and/or fail out of their plane, causing collapse of floor or roof.

According to EC 8 the following general criteria for structural

regularity in plan and elevation should be considered:

The building structure is approximately symmetrical along each

principal axis in plan, for both stiffness and mass distribution. A

sufficient number of load bearing walls with approximately the

same stiffness, should be provided in both principal direction of the

building -see Fig. 8


85

Figure8: Structural wall distribution in plan

The plan shape should be simple. Total dimension of projections,

reentrant corners or recesses in one direction is limited to 25% of

the overall dimension of the building in the corresponding direction

-see Fig. 9

Figure 9: Examples of regular configuration of masonry houses in plan

The length of a single portion of the building is limited to four times

its width. In cases where longer building is required, a separation

joint is necessary. The separation should be min 50 mm – Fig. 10

Figure 10: Irregular configurations in plan should be separated in regular

portions

Vertical regularity is achieved by uniform distribution along the

height of the building of stiffness and masses. Lack of vertical

regularity may lead to horizontal plane of weakness/stress

concentration and collapse. Mixed structural systems, such as a

combination of masonry structural walls in one level and RC frame

in the next are not allowed. For planning flexibility is possible

combined system consisting of RC columns and masonry shear

walls. For such configurations the masonry bearing walls should be

reinforced and the RC members should be connected into RC floors

forming frames. The vertical reinforcement of the masonry shear

wall should be anchored into the floor to ensure loads transfer,

The floors are rigid in their plane providing diaphragm action and

interconnected with masonry walls. To this end the floors should be

constructed in a single plane. In cases where large openings are

present in the floor, such as for stairways the contour of the

opening should be strengthened with a bond beam. Also two-way

slabs are preferred to one-way slabs, as they distribute the vertical

gravity loads more uniformly onto the masonry walls.

5. Plan dimensions and height or number of storeys

Limitations concerning the height of masonry wall houses have

been set in most existing seismic codes. Currently EC 8 limits the

construction of confined brick masonry houses located in seismic

zones with high seismic risk ie. ag => 0.3g to four storey houses.

However confined brick masonry wall buildings which conform with

the specifications for structural configuration and quality of

materials, the dimensions of the building are not limited by the


87

code. In this case the dimensions of the house are determined by

design calculations based on the load bearing capacity of the

masonry. The building should be verified according to ultimate limit

states. On the other hand based on the experience from past

earthquake as well as the existing technologies for masonry

housing construction it is recommended that the height and

number of storeys.

6. Distance between masonry bearing walls and wall openings

In EC 8 there is no requirement for maximum distance between

walls. However based on experience for different type of masonry

houses it is recommended that the distance between walls conform

to Table 5.

Table 5- Recommended maximum distance between structural walls

Another essential factor is the structural wall continuity. This

means that the size and configuration of openings in walls should

be carefully planned. The following recommendations regarding the

configuration and size of openings should be observed:

• Openings should be vertically aligned from storey to strorey

• The top ends of openings in the storey should be horizontally

aligned

• Openings should not stop continuous RC bond beams (at

lintel and/or roof level)

• Openings should be located symmetrically in the plan of the

building so that not to get in the way of the uniform

Design ground

acceleration ag

< 0.2

[g]

0.2 - 0.3

[g] >= 0.3 [g]

Unreinforced masonry [m] 10 8 6

Confined Masonry [m] 15 12 8

Reinforced masonry [m] 15 12 8

distribution of strength and stiffness in two orthogonal

directions.

7. Simple houses

According to EC 8 certain class of masonry housing can be exempt

from seismic resistance verification provided that the quality of

materials and construction rules specified in the code are met.

Such houses are named "simple buildings". According to EC 8

simple buildings are regular buildings with an approximately

rectangular plan. The ratio between the long to shorter side of the

house is no more to four and the projections or recesses from the

rectangular shape are not greater than 15% of the length of the side

parallel to the direction of projection. Such houses have the

following limitations regarding number of storeys above ground

(Table 6)

Table 6: Number of storeys above ground, allowed for simple buildings

For masonry house to comply with a simple building a number of

specifications are given for the masonry walls. The structural walls

should be symetrically located in plan in two orthogonal directions.

A minimum of two structural walls per orthogonal direction. The

length of each wall should be greater than 30% of the length of the

building in the wall plane and the distance between these walls

should be maximum 75% of the size of the building in the other

direction. The minimum cross sectional area of the structural walls

is also specified in EC 8. At every floor, the area of the structural

walls in two orthogonal directions is provided as a percentage of the

total floor area above the level considered. Table 7 below gives the

minimum horizontal structural wall cross-section.

Table 7: Minimum horizontal structural wall cross-section, given as 96 of

the total floor area above the level considered (6)

Design ground

acceleration ag < 0.2 [g] 0.2 - 0.3 [g] >= 0.3 [g]

Unreinforced masonry 3 2 1

Confined Masonry 4 3 2

Reinforced masonry 5 4 3

Design ground < 0.2 [g] 0.2 - 0.3 [g] >= 0.3 [g]


89

To enforce reguliarity, the difference in structural walls cross-

sectional area in two orthogonal directions from storey to storey

should be maximum 20%. The difference in the mass of structural

walls in two orthogonal directions from storey to storey should be

as well maximum 20%. For such buildings it is also required that

75% of the vertical load is carried from the structural walls.

8. Details for seismic resistance

8.1 Concept

The performance of building subject to an earthquake motions is

governed by the inter-connectivity of structural components as well

as the individual component's strength, stiffness and ductility.

Thus the details to provide seismic resistance can be classified in

two categories:

Details for complete load path

• Provide wall-to-wall connection ie. tying of walls

• Provide means for walls to foundations connection

• Provide connection of bond beams to roof

• Provide connection of walls to bond beams

• Provide stiff in their plane floors/roofs

Details to improve structural components strength and ductility

• Improve the compressive strength of structural components

acceleration ag

Unreinforced masonry 3 5 6

Confined Masonry 2 4 5

Reinforced masonry 2 4 5

• Improve the bending strength of structural components

• Improve the shear strength of structural components

• Improve the ductility, m of the structural components

9. Bond beams

In the case of confined masonry construction bond beams are

constructed as part of the vertical and horizontal masonry confining

elements. Bond-beams should be constructed in-situ from

reinforced concrete and cast simultaneously with the floor slab.

Bond-beams should be cast on top of all structural walls at every

floor level. The minimum bond beam's cross section is

recommended to be 150x250. The bigger dimension being the

thickness of the wall. Typical examples of monolithic cast in-situ

RC bond beams with RC slabs are shown below on Fig. 11.

Figure 11: Details of cast in-situ RC slabs with bond beams

Maximum vertical distance between bond-beams is 4 m. Bond-

beams are constructed because:


91

• Forms confined masonry shear walls in combination with tie-

columns

• Improves the in-plane stiffness of floors to provide diaphragm

action

• Transfers the horizontal load from the diaphragm to the

structural walls

• Connects the structural walls together and provides out-of-

plane support

• Connects the RC tie-columns

EC8 specifies the following minimum requirements:

Concrete of class 15 should be used

• Cross section size should be not less than 150x150 mm

• Four mild steel rebars with total area 240 mm2

• To ensure integrity of the bond beam the longitudinal rebars

at corners and wall intersections should be spliced a length

of 60f

• Transverse reinforcement-stirrups rebars f6 @ 200 mm

intervals (Fig. 12)

Figure 12- Detail of RC bond showing splicing of rebars at wall corners

According to EC 8 the resistance of the RC bond-beam should not

be taken into consideration in the design calculations.

Consequently there is no mandatory design through calculation for

the bond-beams. As was discussed in the confined masonry section

the design parameters are determined on empirical basis. In Table 7

the members reinforcement can be determined based on the

seismicity of the location the number of stories and position.

Table 8:Recommended reinforcement of horizontal

RC bond-beams

Number

of

storeys

Position

(storey)

Low:

< 0.2 [g]

Moderate:

0.2 - 0.3 [g]

High:

>= 0.3 [g]

2 1-2 4 bars,

#8 mm

4 bars, #10

mm

4 bars, #12

mm


93

10. Tie-columns

Although the tie-columns and bond beams do not provide frame

system adequate splicing and anchoring of rebars is required at all

joints. Sixty rebar diameters splices are required according to EC8.

The cross-sectional area of rebars for tie-columns can be selected in

dependence of seismicity of the location and number of storeys in

the house. On Fig. 13 below is illustrated the splicing of rebars

between bond beam and tie-column.

4 1-2 4 bars,

#10 mm

4 bars, #12

mm

4 bars, #14

mm

4 2-4 4 bars,

#8 mm

4 bars, #10

mm

4 bars, #12

mm

6 1-2 4 bars,

#12 mm

4 bars, #14

mm

4 bars, #16

mm

6 3-4 4 bars,

#10 mm

4 bars, #12

mm

4 bars, #14

mm

6 5-6 4 bars,

#8 mm

4 bars, #10

mm

4 bars, #12

mm

Figure 13: Construction of tie-column for confined brick masonry house

11. Floors and roofs

In EC 8 it is specified that the floor and roof structure can be

constructed in timber or reinforced concrete, provided a diaphragm

action can be achieved. When building confined masonry houses,

RC floor slabs cast in-situ are preferred.

Apart from developing diaphragm action and transfer of the seismic

forces onto the walls the floors and roof should support the walls

out of their plane, ie. all structural walls should be restrained at

floor/roof level. In the case of RC slab the connection is provided

naturally by constructing RC bond beam onto the structural walls.

12. Lintels and cantilever elements


95

Lintels are load-bearing elements which support the weight of the

wall and floor above opening. Lintels can be made from in-situ

reinforced concrete, timber and reinforced masonry. In seismic

zones cast in-situ RC lintels are recommended. If the distance

between the top of the opening to the top of the floor above is less

than 600 mm the lintel can be cast simultaneously with the bond

beam and floor slab as shown on Fig. 14. In cases where the

distance is bigger the lintels can be cast separately (Fig. 14) and

care should be taken to bond the RC lintels to the masonry of the

adjoining wall through horizontal rebars.

Figure 14: Requirement for lintels in seismic zones

13. CONCLUSIONS

Confined brick masonry is also quite ductile, and it can absorb

significantly high energy and undergo large deformation during

earthquake. The confined brick masonry technology also ties up the

entire building together for better seismic performance. Confined

brick masonry construction makes a building very safe against

differential settlement and wind load. Also, the confined brick

masonry construction results in better aesthetics and is convenient

to construct using economically available local material and labour.

In the IS 4326-1993 there exists provision for tying up the building

members together, but the concept of confined brick masonry is not

utilized except in clause 8.5. However, there is clear provision for

confined brick masonry in the Euro Code 8, 1998. The

recommended technology is fully supported by Euro Code 8 and IS:

4326-1993 (Clause 8.5), and therefore, there should be no

hesitation in application of the technology in the Gangetic plain, as

detailed in the report.

REFERENCES

[1] Brzev, S. Sinha, R.., ‘Unreinforced brick masonry building

with RC roof slab’, World Housing Encyclopedia,

Report/India, EERI and IAEE.

[2] IS: 1905-1980, ‘Indian Standard Code of Practice for

Structural Safety of Buildings-Masonry Walls’, Second

Revision-1981, Bureau of Indian Standards, New Delhi.


Earthquake Resistant Design and Construction of Buildings’,

Bureau of Indian Standards, New Delhi.


97


Improving Earthquake Resistance of Low Strength Masonry

Buildings’, Bureau of Indian Standards, New Delhi.


Ductile Detailing of Reinforced Concrete Structures

Subjected to Seismic Forces’, Bureau of Indian Standards,

New Delhi.


Repairs and Seismic Strengthening of Buildings-Guidelines’,

Bureau of Indian Standards, New Delhi.

[7] National Building Code of India 2005’, Bureau of Indian

Standards, New Delhi.

[8] Eurocode 8: ‘Design provisions for earthquake resistance of

structures. Part 1-2: General rules- General rules for

buildings’. ENV 1998-1-2: 1995 (CEN, Brussels, 1995).

[9] Eurocode 8: ‘Design provisions for earthquake resistance of

structures. Part 1-3: General rules- Specific rules for various

materials and elements’. ENV 1998-1-3: 1995 (CEN, Brussels,

1995).

[10] Eurocode 6: ‘Design of masonry structures. Part 1-1:

General rules for buildings. Rules for reinforced and un-

reinforced masonry’. ENV 1996-1-1: 1995 (CEN, Brussels,

1995).

[11] Singh, k. Pramod, (2006), ‘A Report on Composite

Confined Brick Masonry Construction for Four Storey

Apartment Buildings in The Gangetic Plain,

Report/Department of Civil Engineering, IT, BHU, Varanasi.


101

Earthquake Resistant Confined Brick Masonry

Buildings

P. K. Singh

Professor & Head, Department of Civil Engineering, Institute of


1. INTRODUCTION

As per Euro Code 81, a construction system where plain masonry

walls are confined on all four sides by reinforced concrete members

or reinforced masonry is called confined brick masonry (CBM). In

case of CBM buildings the design philosophy adopted is that

neither the brick masonry nor reinforced concrete gets damaged

during earthquake condition.

Euro Code 8 limits the construction of CBM houses located in

seismic zones having design ground acceleration (ag) greater than

or equal to 0.3g to four storeys (Table 1).

Table 1: Recommended maximum height of building (H) and number of

storeys (n).

Design ground

acceleration ag

< 0.2

[g]

0.2 - 0.3

[g]

≥ 0.3

[g]

Unreinforced

masonry

H [m] 12 9 6

n 4 3 2

Confined

Masonry

H [m] 18 15 12

n 6 5 4

Reinforced

masonry

H [m] 24 21 18

n 8 7 6

1.1 Significance of Brick Masonry Buildings in India

As being followed in China and Chile brick masonry apartment

buildings can be the future of the apartment buildings in India also.

Since 1990, base isolated brick masonry buildings with reinforced

concrete floors/roof have been used more widely in China.

Figure 1: Brick Masonry buildings in China

Figure 2: Brick Masonry buildings in Chile

Buildings of confined brick masonry type (Fig 2) are found in all

regions of Chile.

1.2 Socio-Economic Impact


103

Figure 3: Seismic zones II, III & IV of the Gangetic plain having more than

75m soil cover

Seismic zones II, III & IV of the Gangetic plain are shown in Fig 3,

where alluvial soil deposit is having a depth of more than 75m and

goes up to few kilometers in some areas. Total population residing

in the area is approximately 32.91 crores. Therefore, earthquake

resistant confined brick masonry building for the area will have

very high socio-economic impact.

1.3 Technical Details

According to Euro Code 8 the cross-sectional area of rebars for tie-

columns can be selected in dependence of seismicity of the location

and number of storeys in the house. In composite confined brick

masonry buildings the column shall be of 230 mm x 230 mm

having 4 bars of 12 mm diameter as longitudinal reinforcement and

6 mm diameter stirrups at the spacing of 150 mm centre to centre.

The details of column are shown in the Fig. 4.

230 mm

4 x1 2 dia

ba rs

M20

C oncrete

23

0 m

m

6 dia stirru ps @ 150 c/c

6 d ia stir rups @

85 c/c

150

mm

230 mm

4x12 dia ba rs

a) C olum n de tailsb ) Band details

Figure 4: Details of composite column and Lintel level band

The foundation details corresponding to allowable bearing capacity

of 100 kN/m2 is given in Fig.5. The width of strip footing for brick

masonry shall be 1200 mm and the dimensions for column footing

shall be 1200mm x 1200 mm. The column footing shall be

reinforced with 6 bars of 10 mm diameter in both the directions.

4x12 dia

bars

230 mm

250

mm

Lean Concrete

6 dia stirrups @

150 c/c

M20

Concrete

0

6x10 dia bars either way

1200 mm

10

0 m

m

Figure5: Details of foundation for a four storey CBM building

1.4 Site Effect

Seismic effect of local soil conditions on peak ground acceleration

are shown in figure 6.


105

Figure 6: Effect of local soil conditions on peak ground acceleration.

From the figure, it is seen that maximum seismic acceleration is

considerably lower in the alluvium deposit in comparison to the

rock mass.

Fig.7 gives a relationship between the natural period of soil and

alluvium depth.

Figure 7: Relationship between the natural period of soil and

alluvium depth

As the depth of soil deposit increases, fundamental period of the

deposit also increases. Due to plastic deformation and cracking of

the soil, high frequency content of the earthquake waves can not be

supported by the soil, and it quickly dies out in the soil. Therefore,

in deep alluvial soil deposit area only low frequency and high

amplitude earthquake waves are experienced at the ground level.

Fig. 8 shows relationship between damage and the fundamental

period of the soil in the 1967 Caracus earthquake.


107

Figure 8: Damage and the natural period of the soil in the 1967 Caracus

earthquake

As reported for the 1967 Caracus earthquake , buildings up to 3 to

5 storeys constructed at places having soil cover more than75

meters suffered minimal damage, and suffered considerable damage

at places having soil depth less than 75 meters. Similarly, buildings

up to 10 to 14 storeys suffered considerable damage at places

where soil cover was more than 75 meters and suffered minimal

damage at places where soil cover was less than 75 meters.

1.5 Structural Action of CBM

Some structural actions of CBM are presented here for its clear

structural understanding.

1.5.1 Load Sharing

In the CBM building, flexible nature of the slab and the lintel level

band, helps the brick masonry wall and the reinforced concrete

column to act together to support all the vertical loads in direct

compression. Load redistribution between reinforced concrete

column and brick masonry wall mainly at the offsets, ensures equal

strain in the reinforced concrete and the brick masonry at their

interface (Fig. 9).

Brick Masonry

RC Column

a) From column to wall b) From wall to column

Approximately 1 mm

Figure 9: Load redistribution between reinforced column and brick

masonry wall

1.5.2 CBM Action under In-Plane Static Loading

Singh et al.2 conducted experiment on three models, namely; (i)

Reinforced concrete frame without infill (ii) Brick masonry infilled

reinforced concrete frame having no shear connection, and (iii)

Brick masonry infilled reinforced concrete frame with shear

connection. The Load Deflection Curves for the tested models are

given in Figure 10.


109

Loa

d (

kN

)

0

0

2

50

100

150

200

250

300

10

Deflection (mm)

4 6 8 12 13

Infilled frame withoutshear connection

Infilled frame with shear connection

Frame without infill

Figure 10: Load deflection curves for static loading

According to them, in-plane strength of CBM wall may go up

approximately 10 times in comparison to unconfined brick

masonry.

1.5.3 Continuous Lintel Band Action

Effect of continuous lintel band on out of plain vibration of the wall,

and in plane strength of the wall are discussed below.

a) Out of plane effect

In case of CBM building continuous lintel band is provided all

around the building. This lintel band breaks the wall height and

thereby increases stiffness of the wall and results in its reduced

deflection to about one fifth (Fig. 11).

ww

wslab

slab

slab

slab

lintel band

Figure 11: Deflection of BM wall with and without lintel band

Deflections of the wall for the two cases are compared below.

Deflection without lintel band action, ∆ = EI

wh

384

4

Deflection with lintel band action = EI

wh

384

44

3

2

=

38481

16 4 wh ≈

∆/5

Thus, out of plane deflection of the wall reduces to about 1/5th due

to the continuous lintel band where the wall is assumed to be

supported.

b) In-plane effect

Singh, P.K. et al.3 have reported experimental results of in-plane

effect of continuous lintel band. They have tested models of infilled

frame without opening, infilled frame with opening having

continuous lintel band, and infilled frame with opening having

isolated lintel band.

The ultimate load carrying capacity of infilled frame with opening

having continuous lintel band was reported to be 1.7 times that of

the infilled frame with opening having isolated lintel.

1.5.4 Separation of Orthogonal Walls at the Corner

In the CBM buildings, the corner column, which is tied at the lintel

and floor level, provides flexural support to the two orthogonal walls


111

at the corner. This action prevents wall separation at the corner

during earthquake.

2. MODEL TESTS

2.1 Scale effect

The models have been prepared and tested on 1/5th scale. In the

direct model study response of the prototype is directly determined

from measurement of response of the model(5).

2.2 UBM Building model Test

An UBM 2-storey model was prepared on geometrically reduced

scale of 1/5th which is seen as mounted on the shake table in Fig.

13.

Materials used for the brick masonry and RC works in the

experiment are; (i) Portland pozzolana cement, (ii) 1st class country

bricks of size 46×23×14mm, having average water absorption of

10.7% and compressive strength of 35MPa, (iii) Coarse sand of size

4.0mm downgraded to 1mm having FM of 6.29 used as coarse

aggregate, and (iv) washed locally available Ganga river sand used

as fine aggregate having FM of 2.81.

Concrete mix of 1:1.5:3 by weight with water cement ratio of 0.5

was adopted for all RC works, and cement mortar of ratio 1:3 by

weight was used for the brick masonry.

Figure 13: UBM model Mounted on shake table

2.3 Test Results

The building model after failure is seen in Fig. 14.

2.3.1 Amplitude Measurement

Detailed measurements are taken using Laser sensors and

CATMAN Easy software. The amplitude at the roof slab level was

also measured by using a scale mounted on the stand and a pointer

fixed to the model, with the help of video recording.


113

Fig. 14 Test results viewed from East Face

2.3.2 Frequency and amplitude

Amplitude of the shake table was fixed at ± 10mm. In this model

rate of change of frequency was 0.033Hz /s. Total number of cycles

subjected to the model was 122 cycles in 87sec. The time interval

was kept as 3sec for each frequency step.

The plots between time and amplitude at top of the model are

shown in figures 15 and 16, which represent plot for the first 27sec,

and last 54 to 84sec, respectively. From figure 16, it is clear that

model vibrated with maximum amplitude of ± 12.5mm at the top,

with a storey drift of 2.5mm.

Figure 15: Time Vs amplitude plot of UBM model (0-27sec)

Figure 16: Time Vs amplitude plot of UBM model (54-84sec)

g- level at failure of the UBM model

If the displacement / amplitude is given by;

y = a sin ωt

Then, ý = a ω cos ωt

ÿ = - a ω2 sin ωt

And, ÿ max = - a ω2

Maximum acceleration at the base level= -a ω2

= 0.010 * 17.582

= 3.09 m/ sec2

= 0.32g

Maximum acceleration at the top slab level= -a ω2

= 0.0125 * 17.582

= 3.86 m/ sec2

= 0.39g

-15-13-11

-9-7-5-3-113579

111315

0123456789101112131415161718192021222324252627282930

am

pli

tud

e in

mm

Time in sec

Series1

-15

-10

-5

0

5

10

15

5455565758596061626364656667686970717273747576777879808182838485

am

pli

tud

e in

mm

Time in sec

Series1


115

The main conclusion drawn from the above experimentation is that

the UBM building model failed at 0.32g level in bending. Therefore,

reinforcement at the corners was found to be necessary to enhance

the g-level of the building before failure.

3. Building Model as per IS 4326-1993.

A building model geometrically similar to UBM model was prepared

as per IS 4326-19934 provisions, except confinement of openings (

Fig. 17 and Fig. 18).

Fig.17 shows the model where masonry up to window sill level with

corner reinforcement welded to the base plate is completed.

Figure17: Masonry up to window sill level with corner reinforcement

The complete building model mounted on the shake table is seen in

Fig.18.

Figure18: Building model as per IS4326 with Laser sensors

3.1 Test Results

The building model after failure is seen in Fig. 19. The failure took

place at 0.65g level at base level and 1.04g at top. The mode of

failure was failure of the corner steel in tension.

Figure 19: Test results viewed after failure


117

3.3 Frequency and amplitude

Amplitude of the shake table was fixed at ± 5mm. In this model rate

of change of frequency was 0.037Hz /s. The amplitude with respect

to time at roof slab level, middle floor slab level and base level are

measured with laser sensors.

The plots between time and amplitude at top, middle and bottom of

the model are shown in figures 20, 21 and 22, which represent plot

for 14.9-16.08sec, 75.82-76.8sec and 98.8 to 100.0sec,

respectively. Observed maximum amplitude at the top is ± 8.0mm.

Figure 20: Time Vs amplitude plot of UBM model (14.9-16.08sec)


-10

-5

0

5

10

14.5 15 15.5 16 16.5

Series1

Series2

Series3Am

pli

tud

e in

mm

-10

-5

0

5

10

75.5 76 76.5 77

At Top

At Middle

At Bottom


The main conclusion drawn from the above experimentation is that

the building model failed at 5.7Hz frequency i.e. at 0.65g level by

the way of rupture of vertical reinforcement at base level. No other

failure mode was noticed.

4. CBM Building Model

A building model on 1/5th scale and geometrically similar to UBM

model was prepared (Fig.23). Reinforcement details adopted in the

CBM model (at 1/5th scale) are given in the table 6.

Table 6: Reinforcement details of the model

Sl.

No

Particulars Dia of rebars

As per Euro

code 8

Nos./spacin

g

Dia of

wire

used in

the

model

1. Column

reinforcement

12 mm 4 2.4 mm

2. Slab reinforcement 8 mm 22mm c/c 1.6 mm

3. Beam

reinforcement

12mm 2.4 mm

4. Stirrups 6mm 2- legged @

17 mm c/c

1.2 mm

5. Lateral ties 6mm 2- legged @

30 mm c/c

1.2mm

6. Binding wire 22 gauge 26

gauge

-10

-5

0

5

10

98.5 99 99.5 100 100.5

Top

Middle

BottomTime in sec


119

Figure 23: Complete CBM model after mounting on shake table

The model after test is shown in figure 24. As seen from here, there

is no damage to the model at all after the test. The test had to be

stopped due to limitation of the shake table which became unstable

at 7.2Hz frequency.

Figure 24: CBM model after test

4.1 Frequency and amplitude

Amplitude of the shake table was fixed at ± 5mm. Time vs.

amplitude plot for the CBM model are shown in figures 25 and 26.

The CBM model was subjected to amplitude at the top level of

model of ±9mm and base amplitude of ±5mm.

Figure25: Time vs amplitude plot for CBM model ( 0-27sec)

Figure26: Time vs amplitude plot for CBM model (111- 120sec)

The CBM model did not fail even at 7.2 Hz frequency. The

maximum g-level of CBM at the base level was 1.04g and at the top

slab level it was 1.88g.

-6-5-4-3-2-10123456

0 1 2 3 4 5 6 7 8 91011121314151617181920212223242526272829

am

plitu

de

in

mm

Time in sec

-10-9-8-7-6-5-4-3-2-10123456789

10

111 112 113 114 115 116 117 118 119 120 121

am

plitu

de in

mm

Time in sec


121

5. CONCLUSIONS

On the basis of experimental study, the following salient

conclusions are drawn.

1. In the UBM model initial cracks due to flexure appeared at

0.32g – 0.39g at the base in the horizontal direction at a

frequency of 2.82Hz, which lead to final failure. Therefore, there

is need for sufficient vertical reinforcement at the corners to

prevent this type of failure.

2. Building model as per IS4326-1993 failed at the maximum

shake table frequency of 5.7Hz. Fixed amplitude at the base

level was +5mm, and the observed maximum amplitude at the

roof slab level was +8.0mm. The corresponding g-level at the

base was 0.65g and at the roof slab level was 1.04g. Model failed

by the way of corner reinforcement rupture at the base level.

Hence, it is concluded that single bar as the vertical corner

reinforcement is insufficient.

3. The CBM model was subjected to maximum practically possible

frequency of the shake table of 7.2Hz ( ± 5mm base amplitude)

in 486 cycles. No damage to the model was observed, and the

model remained intact after the test.

4. In case of CBM, the shake table amplitude was fixed at ±5mm,

and maximum roof slab amplitude of the model was observed to

be ±9mm. The corresponding g-level at maximum possible

frequency and amplitude was 1.04g at the base, and it was

1.88g at the roof slab level.

5. In the CBM model no separation of the brick masonry and RC at

the interface was observed even at 1.88g level. Therefore, it is

concluded that for the CBM buildings, there is no necessity to

provide offsets, as given in Euro code 8.

6. In the CBM Building model there is no failure or crack observed

at openings and at junction of concrete and masonry even at

1.88 g-level. Hence, it is concluded that there is no need to

confine the openings as given in the IS 4326-1993.

7. The CBM building model, tested without bond beams, exhibited

no deficiency during the test even at 1.88g level. Therefore, it is

concluded that in CBM buildings provision of bond beam below

the slab level, as given in Euro code 8, is not necessary.

ACKNOWLEDGEMENT

The research work was carried out under Special Assistance

Program of the University Grants Commission, New Delhi in the

Department of Civil Engineering, Institute of Technology, Banaras

Hindu University.

REFERENCES

[1] Euro Code 8: ‘Design provisions for earthquake resistance of

structures.’ Part 1-2: General rules for buildings’. ENV 1998-1-

2:1995 (CEN, Brussels, 1995).

[2] Singh, P.K. Saxena S. and Roy. B N (2001) 'Behavior Of Brick

Masonry Infilled Reinforced Concrete Frames Subjected to Static

Loading’ Journal of the Institutions of Engineers (India), vol 82,

no 01, pp 23-29.

[3] Singh, P.K., Singh ,V. and Yadav, S. (2006) ‘Effect of Opening on

Behavior of the Infilled Frame with and without Continuous

Lintel Band’ Journal of the Institutions of Engineers (India), vol

87, pp 33-37.


123

[4] IS 4326-1993, ‘Earthquake resistant design and construction of

buildings – Code of Practice’, Bureau of Indian Standards, New

Delhi.

[5] Ganeshan, T.P. ‘Model Analysis of Structures’ , University Press

(India) Limited.


125

Analysis of Confined Brick Masonry Buildings

P. R. Maiti

Assistant Professor, Department of Civil Engineering, Institute of

Technology, Banaras Hindu University, Varanasi-221005, India

1. INTRODUCTION

Confined brick masonry construction is an alternative to

unreinforced masonry (load bearing) structures and RC frame

construction. It has some features of both the technologies. It

consists of masonry walls with horizontal and vertical RC confining

members built on all four sides of the wall. Vertical members

resemble the columns in RC frame construction but they are of

smaller cross section, they are called tie columns. Similarly

horizontal members are called tie beams. Generally they are

termed as horizontal ties and vertical ties. These members are

effective in i) enhancing the strength of masonry walls under lateral

loads; ii) reducing the brittleness of masonry walls hence improving

their earth quake performance; and iii) confining the members to

restrict damage to masonry wall. The different components of a

typical confined brick masonry building are shown in Figure 1.

Figure 1: Components of confined brick masonry building

Masonry walls: Transmit the gravity load from the slab above

down to the foundation. The walls act as bracing panels, which

resist horizontal earthquake forces. The walls must be confined by

concrete tie- beams and tie-columns to ensure satisfactory

earthquake performance.

Confining elements (tie-columns and tie-beams): Provide

restraint to masonry walls and protect them from complete

disintegration even in major earthquakes. These elements resist

gravity loads and have important role in ensuring vertical stability

of a building in an earthquake.

Floor and roof slabs: Transmit both gravity and lateral loads to the

walls. In an earthquake, slabs behave like horizontal beams and are

called diaphragms.

Plinth band: Transmits the load from the walls down to the

foundation. It also protects the ground floor walls from excessive

settlement in soft soil conditions.

Foundation: Transmits the loads from the structure to the ground.

It must be noted that horizontal and vertical ties may be of various

kind of materials apart from reinforced concrete like steel, timber

etc.

1.2 History and Extent of Application of Confined Brick Masonry Construction

• 1908 - First known use of confined masonry construction

was in reconstruction of buildings destroyed by the Messina,

Italy earthquake of magnitude 7.0.

• 1930 – Confined masonry construction started in Chile (after

1928 Talca earthquake of magnitude 7.8) and Colombia

• 1939 – Another earthquake of 7.8 magnitude hit Mid-

Southern Chile which established the confined masonry

construction as a better earthquake resistant construction.

• 1940 – Confined masonry construction introduced in Mexico

City to control wall cracking caused by large differential

settlement under soft soil condition.

• Confined masonry construction is in practice over last 33

years in Mediterranean Europe, Latin America, Middle East,

Indonesia etc.


127

1.3 Confined Brick Masonry and Reinforced Brick Masonry

In confined brick masonry, reinforcement is restricted in the

confining members and no reinforcement is provided in masonry

itself whereas in reinforced brick masonry construction, vertical

reinforcement is provided in the hollow space in bricks which is

later grouted with a cement grout to avoid corrosion of

reinforcement. Horizontal reinforcement is provided in ladder form

i.e. in horizontal joints. Additional vertical reinforcement is provided

at corners, joints, openings and wherever necessary depending on

the expected severity of seismic load.

1.4 Confined Brick Masonry and RC Frame Construction

In general observation both confined brick masonry construction

and RC frame construction look alike but in reality they are vastly

different in manner to resist gravity and lateral load as well as in

the sequence of construction. Major differences are as following:-

Figure 2: Confined Brick Masonry Construction

In confined brick masonry construction walls are constructed first,

then vertical ties are constructed and at last horizontal ties are

constructed with floor/roof slab whereas in RC frame construction

frame is constructed first and then masonry walls are constructed

since walls are non structural members in this case.

Masonry walls are the main load bearing structures in confined

masonry construction, expected to bear both gravity and lateral

loads while in RC frame construction, all loads are resisted by RC

frame and walls are non load bearing part.

While strip footing is used in confined brick masonry structures,

RC frame structures require isolated footing.

Confining elements are not built to resist moment hence they have

relatively simple reinforcement detailing which simplifies design

and facilitates construction.

Smaller cross section area makes confined masonry structures

cheaper than their RC frame counterparts.

It is to be noted that even with smaller beam/column size and

inadequate detailing too RC frame structures would not perform as

good as confined masonry construction under seismic load due to

inadequate design and construction.

1.5 Failure of Confined Masonry Structure

Failure mechanisms of confined masonry wall panels depend on the

direction of earthquake loading. There are two possible scenarios:

a) Earthquake ground shaking in the direction parallel with the

longitudinal wall axis, also known as in-plane seismic loading, or

b) Earthquake ground shaking perpendicular to the longitudinal

wall axis, or out-of-plane seismic loading.

Mechanisms of seismic response due to in-plane and out-of-plane

seismic loading are discussed in the following sections.


129

1.5.1 In-plane failure mechanisms

There are two major failure mechanism observed due to in-plane

seismic loading in confined masonry buildings

• Shear failure mechanism

• Flexural failure mechanism

a) Shear failure mechanism

It is due to in-plane seismic loads. It is characterized by distributed

diagonal cracking in the wall. These cracks propagate into the tie-

columns at higher load levels, as shown in Figure 3.

Initially, a masonry wall panel resists the effects of lateral

earthquake loads by itself while the confining elements do not play

any significant role. However, once the cracking takes place, the

wall pushes the tie-columns sideways. At that stage, vertical

reinforcement in tie-columns becomes engaged in resisting tension

and compression stresses. Damage in the tie-columns at the

ultimate load level is concentrated at the top and the bottom of the

panel. These locations, characterized by extensive crushing of

concrete and yielding of steel reinforcement, are called plastic

hinges (Figure 4). It is to be noted that the term plastic hinge has a

different meaning in the context of confined masonry components

Figure 3: Shear Failure of the Wall

than that referred to in relation to RC beams and columns, where

these hinges form due to flexure and axial loads.

In confined masonry construction, tie-beams and tie-columns resist

axial loads. Shear failure can lead to severe damage in the masonry

wall and the top and bottom of the tie-columns.

b) Flexural failure mechanism

It is caused by in-plane lateral loads and is characterized by

horizontal cracking in the mortar bed joints on the tension side of

the wall, as shown in Figure 5. Separation of tie-columns from the

wall was observed in some cases (when toothed wall-to-column

connection was absent). Extensive horizontal cracking, which

usually takes place in tie columns, as well as shear cracking can be

observed on Figure 5.

Figure 4: Plastic Hinge in a Confined Brick Masonry

Figure 5: Flexural Failure Mechanism


131

1.5.2 Out-of-plane seismic effect on wall

Seismic shaking in the direction perpendicular to a masonry wall

(also known as out-of-plane seismic loading) causes bending and

shear stresses in the wall. This may result in cracking and possible

wall collapse by overturning. Due to an increase in spectral

accelerations up the building height, the out-of-plane seismic

effects are more pronounced at higher floor levels, as shown in

Figure 6a. In the area affected by the 2010 Maule, Chile

earthquake, wall cracking due to out-of-plane seismic effects was

observed at the top floor level, as shown in Figure 6b (no damage

was observed at lower floors in the same direction). The building

had RC floors and timber truss roof.

The extent of damage and a likelihood of wall collapse depends on

the type of roof and floor diaphragm (rigid or flexible), and how well

the wall is attached to its confining elements (if any). The out-of-

plane bending mechanism is critical mainly for buildings with

flexible diaphragms, which are not capable of transmitting the

lateral forces to the stiffer walls oriented in the direction of the

seismic action. In some cases, this mechanism can also be critical

in buildings with rigid diaphragms due to inertia forces generated

by transverse wall vibrations, as shown in Figure6a. To prevent the

occurrence of this failure mechanism, it is important to restrict the

maximum spacing of tie-beams and tie-columns and ensure tooting

and the interaction between the walls and the confining elements.

Figure 6: Out-of-plane seismic response of confined masonry walls: a) mechanism of seismic response (Tomazevic, 1999), and b) observed damage at the top floor level of a building after the 2010

Maule, Chile earthquake (M. Astroza)

A possible out-of-plane failure mechanism for walls in buildings

with rigid diaphragms is similar to that characteristic of a two-way

slab supported on all ends and subjected to uniformly distributed

loading, as shown in Figure 7a. This damage pattern was observed

at the second floor level of a three-storey building damaged in the

2010 Maule, Chile earthquake, as shown in Figure 7b.

1.5.3 Seismic Response of Multi-Storey Confined Masonry

Building

In multi-story confined masonry buildings, earthquake-induced

lateral forces peak at the ground floor level and cause significant

shear cracking. Under severe earthquake ground shaking, the

collapse of confined masonry buildings may take place due to a soft

story effect (similar to that found in RC frames with masonry

infills), as shown in Figure 8. This behavior was confirmed by

experimental studies (Ruiz and Alcocer, 1998; Alcocer et al., 2004).

Figure 7: Out-of-plane seismic effects in confined masonry walls: a) two-way slab mechanism, and b) evidence from the 2010 Maule, Chile earthquake (S. Brzev)

Figure 8: Soft-story collapse mechanism for multi-storey confined

masonry buildings (Alcocer et al., 2004)


133

1.6 Key Factors Influencing the Seismic Resistance of Confined

Masonry Structures

a) Wall density

Wall density is the key parameter influencing the seismic

performance of confined masonry buildings. Evidence from past

earthquakes show that confined masonry buildings with adequate

wall density were able to resist the effects of major earthquakes

without collapse. The wall density is quantified through the wall

density index d, which is equal to d = AW/AP .

Where, AP is area of the building floor plan, as shown in Figure 9,

and AW is equal to the cross-sectional area of all walls in one

direction, that is, a product of the wall length and thickness. When

performing the AW calculations it is not necessary to deduct the

area of tie-columns and area of voids in hollow masonry units. It is

very important to note that wall cross-sectional area should not be

included in the Aw calculation in the following cases:

a) Walls with openings, in which the unconfined opening area is

greater than 10% of the wall surface area, and

b) Walls characterized by the height-to-length ratio greater than

1.5.

The d value should be determined for both directions of the building

plan (longitudinal and transverse).

Figure 9: Wall Density Index Parameter

b) Masonry Unit

The following types of masonry units are acceptable for confined

masonry construction:

1) Solid concrete blocks

2) Hollow concrete blocks

3) Solid clay bricks

4) Hollow clay tiles (blocks).

The hollow units are those having, in their most unfavorable cross

section, a net area at least 50% of the gross area, and exterior face

shell thickness of not less than 15 mm (Figure 10a). For hollow

units with two to four cells, the minimum thickness of the interior

webs is 13 mm. Multi-perforated units are those with more than

seven perforations or cells (Figure 10 b). For multi-perforated units

having perforations of the same dimensions and distribution, the

minimum thickness of the interior webs is 7 mm.

Hollow masonry units should be used with caution in non-

engineered buildings. To ensure satisfactory seismic performance of

masonry walls built using concrete blocks, it is critical that the

minimum material strength and construction quality

recommendations outlined in this document have been met. Note

that wall density index are by 33% higher for walls built using

hollow concrete blocks compared to those built using solid units.

Figure 10: Different types of hollow bricks


135

Perforations in solid masonry units are permitted. However, the

ratio of net to gross area should be greater than 75%.

The following types of units are not recommended for confined

masonry construction:

1) Masonry units with horizontal perforations, and

2) Natural stone masonry and adobe (sun-dried earthen units).

c) Mortar

It should be noted that hydraulic cement is commonly used for

masonry wall construction. Masonry cement is pre-mixed in a plant

and it consists of a mixture of Portland cement and plasticizing

materials (such as limestone or hydrated or hydraulic lime), and

other materials introduced to enhance one or more properties such

as setting time, workability, water retention and durability.

Masonry cement is not commonly used for load bearing wall

construction, except for rendering wall surfaces to avoid the mortar

shrinkage cracking

d) Masonry

Masonry strength has a significant influence upon the seismic

resistance of a confined masonry buildings and life safety of its

inhabitants. It is therefore extremely important to perform basic

tests using local masonry materials; this is particularly important

for projects involving several buildings.

Compressive strength is a very important property of masonry, and

it may be highly variable depending on local materials and

construction practices. The design compressive strength (fm) for the

combinations of typical masonry units and mortars used in local

housing construction practice should preferably be determined by

testing prism specimens made of the masonry units and mortar

used at construction sites. The prisms should be tested using same

procedures as other masonry wall applications (NTC-M, 2004).

e) Tie – Columns

Tie-columns significantly influence the ductility and stability of

cracked confined masonry walls. The provision of closely spaced

transverse reinforcement (ties) at the top and bottom ends of tie-

columns results in improved wall stability and ductility in the post-

cracking stage (Alcocer and Klingner, 1994).

f) Horizontal wall reinforcement

In many countries where confined masonry construction is

practiced, reinforcement is usually not provided in masonry walls.

However, in four-to-five storey construction in Peru there is a

tendency to provide horizontal joint reinforcement in the form of

one or two wires laid in the mortar bed joints. The Mexican Code

NTC-M 2004 prescribes that the horizontal reinforcement, when

provided, be placed continuously along the wall length. Horizontal

rebars should be anchored into the tie - columns; the anchorage

should be provided with 90o hooks at the far end of the tie-column.

The hooks should be embedded in the concrete within the tie-

column. The bar diameter should be larger than 3.5 mm and less

than ¾ the joint thickness. Research studies have shown that

horizontal reinforcement has a beneficial effect on wall ductility.

Specimens with horizontal reinforcement showed a more uniform

distribution of inclined shear cracks than the unreinforced

specimens. Cold-drawn steel wires are used as horizontal

reinforcement in Mexico; these wires are made of steel without a

defined yield plateau, where strain hardening develops at very small

strains (0.002 to 0.0025). The type of steel used for horizontal

reinforcement influences its effectiveness in enhancing masonry

shear resistance. Early experimental studies used horizontal

reinforcement made of high carbon steel that exhibited elasto-

plastic behavior.

g) Openings

An experimental research study showed that, when the opening

area is less than approximately 10% of the total wall area, the wall

lateral load resistance is not significantly reduced as compared to a

wall without opening (Yanez et al. 2004). The walls with larger

openings develop diagonal cracks (same as solid walls), except that

the cracks are formed in the piers between the openings; thus,

diagonal struts form in the piers, as shown in Figure 11. Most


137

building codes prescribe the maximum permitted opening size

beyond which the tie-columns need to be provided.

1.7 Types of damages

Brick masonry is a common construction material in India because of its abundance, low cost, good sound and heat insulation properties and availability of skilled labor. Masonry is extensively used in India as in infill walls in reinforced concrete buildings. When buildings are subjected to earthquakes, various states of damages occur.

a) Nonstructural damage b) Slight structural damage c) Moderate structural damage d) Severe structural damage e) Collapse

It is obvious that unreinforced masonry buildings are among the most vulnerable structures in earthquakes. Fragility curves are used by different researcher to asses the probabilistic damages of a building in earthquake. Fragility curves provide a powerful tool for anticipating the damage to structures in future probable

Figure 11: Failure mode in confined masonry with wall openings

earthquakes. The effects of different parameters on the seismic behavior of these structures can be investigated through using fragility curve. Fragility curves for a specific type of buildings is a probabilistic method to estimate the probability that the building will exceed a specific state of damage for a definite value of seismic intensity parameter. In the present study, different types of failure of brick masonry building during past earthquake are critically pointed out from existing literature in introduction portion. The stress analysis of brick with mortar joints are analyzed numerically using ANSYS software. A prototype is presented here for experimental analysis in shaking table. 2. Stress analysis of brick masonry using numerical modeling

A best numerical model is the one that represents the maximum

characteristics of the actual model. The process of representation of

an actual object into a numerical model in particular software

needs continuous refinement. The existing numerical models for

masonry have been divided into two groups, the heterogeneous and

homogeneous models. The heterogeneous models analyze the

masonry walls discretizing bricks and mortar separately through

finite element and or interface elements. A suitable constituents

relationship is then assumed for each component. In this way it is

possible to take account with particular accuracy, the characteristic

of mortar joints, which play very important role in the global

behavior for masonry.

Numerous finite element programs are available now a day for

numerical modeling of structures including SAP, ETABS, ADINA,

ABAQUAS, ANSYS. In case of heterogeneous models elastic

properties of brick unit and mortar joints are assigned separately to

numerical models. A certain value of modulus of elasticity E has

assigned to the solid elements representing mortar and different

values of E is assigned for brick.

Finite element is a mathematical method which makes calculations

by dividing complex structures into very little elements. ANSYS

program is a program which puts forth the performance and

possible fracture loads of constructions into consideration in virtual

medium. The program puts forward how a whole construction

collecting the behavior and effect of every little piece in the system

will display behavior. The results can be obtained as tables or

graphics. The solution of very complex systems as geometrical scale

or an equation can be made with ANSYS program. Therefore, it can


139

be used in the modeling of confined brick masonry constructions

effectively.

SOLID65 element may be used in modeling of RC confining

elements as well as the masonry prism. SOLID65 (shown in Figure

12) may be used for the 3-D modeling of solids with or without

reinforcing bars (rebar). The solid is capable of cracking in tension

and crushing in compression. The element is defined by eight nodes

having three degrees of freedom at each node: translations in the

nodal x, y, and z directions. Up to three different rebar

specifications may be defined. The most important aspect of this

element is the treatment of nonlinear material properties. The

concrete is capable of cracking (in three orthogonal directions),

crushing, plastic deformation, and creep. The rebar are capable of

tension and compression, but not shear. They are also capable of

plastic deformation and creep.

LINK8 element has been used for modeling of main bars as well as

stirrups in reinforcement. LINK8 (Fig. 13) is a spar which may be

used in a variety of engineering applications. This element can be

used to model trusses, sagging cables, links, springs, etc. The 3-D

spar element is a uni-axial tension-compression element with three

degrees of freedom at each node: translations in the nodal x, y, and

z directions. Plasticity, creep, swelling, stress stiffening, and large

deflection capabilities are included. The element is defined by two

nodes, the cross-sectional area, an initial strain, and the material

properties.

Figure 12: SOLID 65 element

COMBIN39 (Fig. 14) is used for modeling the mortar-masonry unit

interface. It is a unidirectional element with nonlinear generalized

force-deflection capability that can be used in any analysis. The

element has longitudinal or torsional capability in 1-D, 2-D, or 3-D

applications. The longitudinal option is a uniaxial tension-

compression element with up to three degrees of freedom at each

node: translations in the nodal x, y, and z directions. The element

has large displacement capability for which there can be two or

three degrees of freedom at each node.

2.1 Example problems

Marinilli, Angelo and Castilla, Enrique presented a paper in 13th

World Conference on Earthquake Engineering, 2004 held at

Vancouver Canada, titled as Experimental Evaluation of Confined

Masonry Walls with Several Confining Columns. In present study

their experimental models are used to make a model in ANSYS and

results would be compared with the experimental findings. The

same model as used in literature is used here to model in ANSYS.

The specification and dimensions are kept same in the ANSYS

model.

Figure 13: LINK 8 geometry

Figure 14: COMBIN39 GEOMETRY


141

Model Specifications

As shown in Figure 15, the first specimen “M1” consisted of one

panel and two confining-columns. The second specimen “M2”

consisted of two panels and three equally spaced confining-

columns. The third specimen “M3” also consisted of two panels, but

the central confining-column was located at ⅓ of the specimen

length. Finally, the fourth specimen “M4” contained three panels

and four equally spaced confining-columns.

H = 2.3 m and L = 3 m

Bottom beam cross section = 0.3 x 0.5 m

Top beam cross section = 0.2 x 0.15 m

Confining column cross section = 0.15 x 0.15

Concrete blocks were used as masonry units with

dimensions of 0.4 x 0.15 x 0.2 m

4: 1:1Sand, lime cement mortar was used

Masonry units of compressive strength = 8.5 N/mm2

Mortar of compressive strength = 7.0 N/mm2

Concrete of grade M25 was used to make RC confining

elements

Reinforcement detailing of RC confining elements

Steel of nominal yield strength = 420 N/mm2

The confining- columns and the top beam were reinforced

lengthwise with four 12Φ bars. The confining-columns were

reinforced transversally with 10Φ stirrups at 60 mm intervals at the

400 mm ends of the elements and at 120 mm intervals in the

remaining portions. The top beams were reinforced transversally

with 10Φ stirrups at 100 mm intervals.

Figure 15: Model used by Marinilli and Castilla (2004)

ANSYS Model of specimen M1, M2, M3 and M4

Figure 16a: Volume plot of model M1

Figure 17a: Meshing of model

M1

Figure 17d: Reinforcement detailing of model M2

Figure 16b: Volume plot of model M2

Figure 17c: Meshing of model M2

Figure 16c: Volume plot of model M3

Figure 16d: Volume plot of M4

Figure 17b: Reinforcement detailing of model M1


143

Physical Testing of any model is a cumbersome task. It not only

involves manpower but also ample time and resources. Hence

physical test may not be done frequently and for every model,

especially for buildings of lesser importance. The maximum

displacement response of the top corner of the right side of each

model has been calculated for different frequencies and presented

in the Figure 18. In the Figure 18 the Y-axis VALU represents the

maximum displacement and X-axis applied frequency.

Figure 17e: Meshing of model M3 with boundary condition

Figure 17f: Reinforcement detailing of model M3

Figure 17g: Meshing of model M4

Figure 17h: Reinforcement detailing of model M4

It is observed from Figure 18 that maximum deflection in x-

direction of the top right corner shows that at lower frequencies

deflection value is quite less and it increases almost linearly but

after a certain limit, which obviously is different for different

specimen, the deflection value shoots to very high value. This

abrupt change in deflection value indicates the brittle behavior of

confined masonry. And the point at which this happens may be the

point when cracking starts. As the frequency is increased;

deflection at the top corner increases slowly but after a certain

frequency deflection response changes abruptly. This abrupt

change in displacement may be due to brittle failure of the confined

masonry.

Figure 18a: Frequency Vs Max deflection curve for the top right corner of specimen M1

Figure 18b: Frequency Vs Max deflection curve for the top right corner of specimen M2

Figure 18c: Frequency Vs Max deflection curve for the top right corner of specimen M3

Figure 18d: Frequency Vs Max deflection curve for the top right corner of specimen M4


145

3. Modeling of a masonry building

It is obvious that masonry buildings are among the most vulnerable

structures in earthquake. Therefore, evaluating the seismic

performance of these buildings is essential against earthquakes

toward the hazard mitigation and risk assessment. It is very

difficult to perform test of the building in laboratory. One may

construct the model of the building by simulation. In this section

one model is presented and described how one can model the

prototype and use for experimental purpose. Here one modeling for

laboratory experiments is illustrated for analysis.

Prototype

The masonry building dimension is approximately 4.0 m x 4.0 m

and the overall height of the building is 3.0 m. The building has one

door and two windows opening. The dimension of the door and

windows are taken as D= 1.2 m x 2.1 m and W=1.2m x 1.0 m

The building is 1:10 prototype.

Similitude between model and prototype

1. Geometric similarity

2. Kinematic similarity

Geometric similarity: Length scale ratio mr

p

LL

L=

Here 1

10rL =

Kinematic similarity:

Time scale ratio mr

p

TT

T=

Velocity scale ratio

m

m m rr

pp r

p

LV T L

VLV TT

= = =

Acceleration scale ratio 2

rr

r

La

T=

Both model and prototype subjected to gravity loads, therefore

acceleration scale ratio 1ra =

From which 1

10rT = and

1

10rV = and Frequency scale ratio

10rF =

Figure 19: Plan and Elevation of the model

Figure 19: Elevation of the model


147

Weight calculation of the model

Brick size: 30mm x 15 mm x 10mm

Weight of one brick = 9.56 gm (Reduced model of brick made by

clay in laboratory)

Weight of single storey

Weight of Masonry

Thickness of walls = 25 mm

Height of walls = 300 mm

Length of walls = 400 mm

Volume of Building = 4 x (25 x 300 x 400 – 25 x 25 x 300)

= 1.1 x 107 mm3

Deductions

Door opening = 120 x 210 x 25 = 6.3 x 105 mm3

Window opening = 2 x (120 x 100 x 25) = 6.0 x 105 mm3

Volume of columns = 4 x 25 x 25 300 = 7.5 x 105 mm3

Total volume of brickwork = 1.1 x 107 – (6.3 + 6.0 + 7.5) x 105

= 9.02 x 106 mm3

Volume of one brick (with mortar) = 31 x 16 x 11 = 5.4 x 103 mm3

No. of Bricks = 9.02 x 106 / 5.4 x 103 = 1670.37 = 1675 approx.

Weight of one brick = 9.56 gm.

Weight of brickwork = 1675 x 9.56 = 16013 gm = 16.013 kg.

Weight of mortar = .02 x 9.02 x 10-3 x 1540 = 0.271 kg.

Weight of concrete (column and slab) = (25 x 25 x 300 + 10 x 400 x

400) x 2.4 x 10-6 = 4.29 kg.

Total weight = 16.013 + 0.271 + 4.29 = 20.57 kg.

2. Total weight of building Model

Total weight = 20.57 x 4 = 82.3 kg.

4. Brick and mortar strength test in laboratory

The failure of brick masonry depends on the strength of mortar and

bricks used. The test of bricks may carry out in the laboratory. Few

schematic view of the failure of brick masonry is presented in this

section.

Figure 20: Failure of brick masonry under compression testing Machine

4. Concluding remarks

Present study covers all the aspects of Confined Brick Masonry

structures. Starting from its various parts to its seismic behavior

and after that construction guidelines were discussed in detail. A

thorough survey of literature gave us insight in the research area of

Confined Brick Masonry clearly. This clearly shows that though

much has been done in this field when it comes to actual testing of

prototypes but the area is still underdeveloped in terms of

analytical modeling.


149

References

[1] Alcocer, S. M., Arais, J. G. and Vazquez, A. (2004). Response

Assesment Of Mexican Confined Masonry Structures Through Shaking Table Tests. Proceedings of 13th World Conference on Earthquake Engineering, Vancouver, B.C., Canada, Paper No. 2130.

[2] Alcocer, S.M. and Klingner, R. (1994). Masonry Research in the Americas. Masonry in the Americas, ACI Publication SP- 147, American Concrete Institute, Detroit, pp.127-169

[3] Astroza, M., Cabezas, F., Moroni, M., Massone, L., Ruiz, S., Parra,E., Cordero,F., and Mottadelli, A., 2010. Intensidades Sismicas en el Area de Daños del Terremoto del 27 de Febrero de 2010, Universidad de Chile, Santiago (in Spanish).

[4] Ali, S. and Page, A., W.,” Finite Element Model for Masonry subjected to Concentrated Loads”, Journal Structure Division ASCE, Vol.114, No.8, 1988, pp.1761-1784.

[5] Asteris, P.G., Syrmakezis, C.A. (2005) Strength of Unreinforced

MasonryWalls Under Concentrated Compression Loads , Practice Periodical on Structural Design and Construction, ASCE, Vol. 10, No. 2, pp. 133-140.

[6] Kanit, R. and Donduren, S. (2010). Investigation of Using ANSYS Software in the Determination of Stress Behaviors of Masonry Walls Under Out of Plane Cycling Load. International Journal of the Physical Sciences, Vol 5(2), pp 097-108.

[7] Kazemi, T. M., Asl, M. H., Bakshi, A. and Rofooei, R. (2010). Shaking Table Study Of A Full- Scale Single Storey Confined Brick Masonry Building. Transaction A: Civil Engineering, Vol. 17, No. 3, pp 184-193.

[8] Marinilli, A and Castilla, E (2004) Experimental evaluation of confined masonry walls with several confining columns, Proceedings of 13th World on earthquake engineering, Vancouver, B. C, Canada Paper No-2129.

[9] Milani, G., Lourenço, P.B., Tralli, A., “ Homogenised Limit Analysis of

Masonry Wall, Part I: failure Surface” ,Computers & Structures, 84(3-

4),2006, pp. 166-180

[10] Parikshit Verma (2011) Analysis of confined brick masonry, M. Tech Dissertation, Department of Civil Engineering, IT BHU Varanasi.

[11] Tomazevic, M. (1999). Earthquake-Resistant Design of Masonry Buildings. Imperial College Press, London, U.K.

[12] Moroni, M., Astroza, M., and Mesias, P. (1996). Displacement Capacity And Required Storey Drift In Confined

Masonry Buildings. Proceedings of 11th World Conference on Earthquake Engineering, Acapulco, Mexico, Paper No. 1059.

[13] Yoshimura, K., Kikuchi, K., Kuroki, M., Nonaka, H., Kim, K. T., Wangdi and Oshlkata, A. (2004). Experimental Study for Developing Higher Seismic Performance of Brick Masonry Walls. Proceedings of 13th World Conference on Earthquake Engineering, Vancouver, B.C., Canada, Paper No. 1597.

[14] Yanez F, Asrroza M, Holmberg A and Ogaz O. (2004) Behaviour of Confined Masonry Shear walls with large openings, Proceedings of 13th World conference on earthquake Engineering, Vancouver, B. C Canada paper No-3438.


153

A Study on Indian Codes and Performance Based

Design

Dipendu Bhunia

Assistant Professor, Department of Civil Engineering, Birla Institute

of Technology & Science, Pilani, Rajasthan, India

1. INTRODUCTION

The calamities due to earthquake prove to be much more disastrous

in India in comparison with the similar incidences in developed

countries. Authors believe that this is due to lower, nonrealistic

design standards and differences in the quality of constructions

and construction practices. The process of Earthquake-disaster

mitigation starts with acquiring the state-of-the-knowledge. Equally

important and rather difficult is to translate and communicate that

knowledge so as to put into state-of-the-practice. Seismic design

codes are the tools by which knowledge in Earthquake Engineering

is conveyed to the field. Often, the aseismic design and construction

is considered as merely a dynamic analysis of structure. Even

today, Indian seismic codes suffer serious shortcomings, including

conceptual errors. This paper is an attempt to address problems

with some of the provisions of IS 1893 (Part 1): 2002 through the

concept of performance based design.

1.1 Seismic Design

Current approach of codified seismic design in most of the

countries is about 70 years old and is based on satisfying force

demands. One major drawback of the approach is that, it does not

directly address the inelastic response in terms of either the forces

or deformations. It has been recognized that losses due to

nonstructural damage and loss of utility services of modern society

could be much greater than structural damage (ATC40, 1996,

FEMA273, 1997). The Loma Prieta Earthquake of 1989 and the

Northridge Earthquake of 1994 resulted in large-scale unacceptable

damage to modern structures, which complied with the prevailing

building codes. The earthquakes of magnitude even greater than

these were anticipated in the US codes. Owing to this, Structural

Engineers Association of California (SEAOC) felt the need for

development of a new, performance based design philosophy. The

post-earthquake studies showed the flaws of using force-based

designs. The inconsistency between prescribed linear analyses

techniques and specifications of reduced seismic loads for design

based on ductile nonlinear behavior led to unacceptable seismic

performance in large number of structures. (Vipul Prakash, 2004)

1.2 Performance Based Design

In 1992, Federal Emergency Management Agency (FEMA)

sponsored the development of national consensus guidelines for

seismic retrofit of buildings, the ATC-33 project. This was the first

effort to standardize the performance based design (PBD) approach.

This project documented the qualitative descriptions of performance

levels. The approach used, was quickly adopted by SEAOC’s Vision

2000 committee and extended to include the design of new

buildings. Together, the FEMA-273 NEHRP (National Earthquake

hazard Reduction Program) Guidelines and Vision 2000 Report

have defined the current state of practice in performance based

design and created awareness among engineering fraternity. The

intent was to establish a design framework that leads to structures

of predictable performance during different levels of seismic

shaking. The structure needs to satisfy certain performance criteria

in order to achieve specified performance objectives for different

levels or damage states. A design performance objective is an

expression of desired performance level for the building for each of

the considered earthquake level. In short performance level is

indicative of anticipated and acceptable damage state (ATC40,

1996; Bertero)

2. Design Criteria

Once the performance objectives are selected, the associated

limiting values become the acceptability criteria to be checked at


155

concerned stages of the design. For a given performance objective,

calculated response quantities must not exceed appropriate

performance limits. The limiting values of the responses are

associated with the damage levels for a specified earthquake ground

motion. The responses include system level building response (e.g.

lateral drift) and element responses. Usually three different

performance levels or limit states are specified (Bertero; Dowrick,

2003):

Serviceability: corresponds to limit state of serviceability for a minor

earthquake. It is desired that structure deform in elastic range and

do not suffer any damage.

Repairability: corresponds to damage control limit state for a

moderate earthquake, structures may have entered in an inelastic

range but damage is repairable.

Safety: corresponds to limit state collapse for a severe earthquake,

structures enter well in to the inelastic range so that damage need

not be repairable but total collapse is not allowed.

ATC-40, FEMA-273 documents prescribe more elaborate

performance levels, which are based on specific requirements of

owners.

3. Capacity Design Basis

Capacity design is a seismic design approach in which distinct

structural components, such as plastic hinges in members, are

chosen and detailed for energy dissipation according to a desired

mechanism of nonlinear lateral deformation. All other structural

components and actions are provided with sufficient strength to

prevent failure under the chosen mechanism. FEMA 273 and ATC

40, establish in part, this approach. This approach is in fact a

prerequisite of using a nonlinear static procedure, however, the

FEMA 273 document does not explicitly explain that a capacity-

design approach must be followed.

Failure Mode Control: In general, good designs not only seeks to

keep the overall probability of failure below a given level but it

arranges the system such that less desirable modes are less likely

to happen than other modes of failure. This increases the reliability

of the design by decreasing the potential for damage and increasing

the overall safety. The less desirable modes of failure for a structure

are (1) those resulting in total collapse of the structure, e.g. failure

of vertical load carrying system, (2) those involving sudden failure of

a member/structure e.g. shear and/or torsion modes. The number

of possible failure modes are substantially reduced by suppressing,

the chances of occurrence of undesirable failure

mechanism/modes. The capacity design principles aim at achieving

this (Dowrick, 1994)

4. Trend of the Seismic Design Codes

ATC 34 and Vision 2000 Report presents the goal of the future

seismic design code/s. They include the short-term goal, mid term

goal and long term goal. The long term goal is to draft the guidelines

of the performance based design in to code form and to complete

the overhaul of seismic design practice during the period 2000 to

2005. Hamburger R., Whittaker A. et al have presented the

summary of ATC 58 project related to development of codes of next

generation.

A brief report on The ATC 58 Project mentions about the Two Stage

Implementation Plan for the project. The first phase will comprise

development of performance verification procedures that will permit

an engineer to evaluate the performance of an existing building or

of a proposed design for a new building using the same terms

defined by decision makers. Verification procedures will include

rules to model buildings and simulate their response to a range of

earthquake events, each having different intensity. The goal of the

second phase will be to develop design and stakeholder guidance to

use performance-based design. The acceptability of current code

performance will be evaluated and appropriate minimum

performance levels for structures of differing occupancies will be

recommended. The final documents would be in the form of design

guidelines and resource documents for use in developing future

building codes.

4.1 Scenario in India


157

Jain S. K. (2003) has presented a review of IS 1893-2002

(Part 1), suggesting major improvements in the future edition. The

gap of 18 years in revising the penultimate version of the code is

referred to as a barrier in inclusion of advances that have occurred

in the knowledge related to earthquake resistant design of

structures during those 18 years. Some of these new developments

have been incorporated in the 2002 version of the code, while many

others have been left out so that the implementation of the code

does not become too tedious for Indian engineers. Vipul Prakash

(2004) has critically reviewed the development of the Indian seismic

codes starting with IS 1893-1962 to 2002 (Part I) and IS 13920-

1993. The paper addresses an important issue of exploring

potential for implementation of PBE in India. The author expresses

an urgent need to incorporate the elements of earthquake

engineering at the basic level of the undergraduate program of

engineering education, as a prerequisite. Two-level performance

criterion and the way of achieving it are suggested.

5. Provisions of IS 1893 (Part I): 2002

Design Criteria: (IS 1893 (Part I): 2002 is referred to as IS 1893

henceforth in this paper.)

This is stated as: to ensure that the structures possess at least a

minimum strength to withstand minor earthquakes (<DBE), which

occur frequently, without damage, resist moderate earthquakes (DBE)

without significant structural damage though some non-structural

damage may occur and aims that structures withstand a major

earthquake (MCE) without collapse.

Thus a three level performance is desired. IS 1893:1962, 1966,

1975 and 1984 specified earthquake loading corresponding to

single seismic event for use in force based analysis and design, even

though the stated performance objective specified two levels of

earthquake ground shaking intensities: moderate and heavy. IS

1893 (Part I): 2002 specify two levels of earthquakes: maximum

considered earthquake (MCE) and design basis earthquake (DBE),

but desires three level performance criteria (Vipul Prakash, 2004).

While defining the design basis earthquake IS1893 states that it is

the earthquake, which can reasonably be expected to occur at least

once during the design life of the structure. The definition becomes

vague in light of a loosely used term ‘design life’. The very first issue

that bothers the structural engineers around the world is the life of

the structure. Neither any of the IS codes define this nor this is

covered in academic curricula in India. A foreword of IS 1893

mentions about the seismic hazard level with respect to ZPA at 50%

risk level and 100 years service life. If the service life is considered

as design life, dose it mean that all structures should have the

same life span irrespective of the structure type and the

maintenance levels of the structure? Maximum considered

earthquake is simply defined as the most severe earthquake effects

considered by the code. There is no mention of the probability of

exceedence of the level considered.

6. Concept of Response Reduction Factor ‘R’

This was introduced to permit elastic force based design for a

system that is expected to respond inelastically in the design

earthquake/s. This was a necessity in context of the knowledge

then available in 1970s. But this resulted in problems, which exist

even today. The factor was assumed to be period independent and

the values assigned to the structural systems were purely empirical

and judgment based. Over the years the values might have been

refined but the basis has remained the same. Second important

consideration is that the R value is system based but is used to

derive the seismic design forces at component level without paying

attention to the redistribution of the forces due the presence of

inelasticity. There is no explicit relationship of this factor with the

fundamentally depended attributes of the structural systems:

ductility, overstrength and the redundancy. If the reduction from

MCE level to DBE level (dividing the zone factor ‘Z’ by 2) is

considered to depend on possible overstrength in material and

members then ‘R’ is left with redundancy and ductility. Among

these, redundancy is a term, not yet adequately defined. If it refers

to statical indeterminancy, it gets exploited while designing:

member sizes for indeterminate systems are substantially lower

than corresponding determinate system. If availability of alternate


159

paths for load transfer are considered to be synonymous for

redundancy, it conflicts with consideration of inelastic effects which

are distributed throughout the system possibly not leaving behind

any UNUSED path for load transfer. Finally it may be only the

ductility, which is relied upon for the use of ‘R’. Further this refers

to system/global ductility demand (lateral drift at roof) and member

or section ductility is few (3 to 6) times of that. Designers and

executors must be made aware rather cautioned about the

manifestation of this. IS1893 specifies these values and IS 13920

states the rules for ductile design but there is no procedure to

ensure whether this ductility has really manifested in the design.

The need to rectify the inconsistencies in R-value is one of the main

arguments for implementation of the performance based design.

6.1 Computation of Fundamental Time Period

The objective of clause 7.8.2 is to check the overestimation of the

time period via through dynamic analysis methods in order to

reduce the design base shear. The lower limit on time period is

imposed through empirical formulae based on experimental and

field investigations. These formulae are independent of stiffness

and/or mass distribution in the system. IS 1893 prescribe formulae

in this regard for (1) moment resisting frame (steel/concrete)

without brick infill panels; (2) for all other buildings and for the

case of moment resisting frame with brick infills.

ATC 3-06 prescribes almost the same formulae with an exception

for RC frames, which is:

0.75T= 0.061 h

{0.75

aT = 0.075 h for moment resisting RC frame building as per IS

1893}

As mentioned earlier, these formulae are empirical and in case of

ATC specifications, based on data from buildings in USA. If ATC

formulae are adopted in IS 1893, is it based on assumption of

similar conditions of buildings in India and USA? If the minor

change in formula for RC frames is considered as indication of

thought given to the prevailing differences in the two countries, it

leads to confusion with formula for estimation of modulus of

elasticity of concrete. IS 456-2000 defines this as:

c ckE 5000 f=

ACI 318M-99 defined this as:

c ckE 4700 f=

It can be seen that the IS 1893 period formula overestimates the

period in comparison with ATC formula where as it uses higher

value of modulus of elasticity of concrete than that given by ACI.

This clearly shows the inconsistency in the IS 1893 specifications.

Majority of structural engineers in India use approximate

fundamental time period method to estimate the design lateral

forces. IS 1893’s inappropriate basis in this context has not affected

their designs because most of our buildings are of medium rise type

having time period (actual) in the range of 0.25 to 1.0 second.

Fortunately the short period range of IS 1893 response spectra

covers time periods from 0.1 to 0.67 seconds giving a constant

spectral acceleration value, making the process of time period

estimation insensitive to design lateral forces in this range.

The objective of the fundamental time period method is to ensure

provision of certain minimum strength. In this context provision of

UBC 97 appears to be more rational. Along with response spectrum

approach, UBC 97 specify a formula for base shear, which is

independent of time period as well as response reduction factor. It

depends on seismic weight of the structure and seismic zone

characteristics, as was there in Seismic Coefficient Method of IS

1893-1984. This could have been modified and retained to provide

the lower bound on design base shear.

7. Regularity of Building Configuration

Clause 7.1 defines the regular and irregular configurations through

the aspects of geometrical details and distribution of stiffness and

mass. Almost five-page information through 30 diagrams is

presented. The source of the information contained under this

clause can be found in any standard text on the subject:

Earthquake Resistant Design of Buildings/Structures. The


161

requirements of regularity in various forms are basically a useful

guideline to planning authority and of little use to design authority.

An attempt to codify/document such requirements will rarely be

fruitful because there can be infinite number of possibilities in this

aspect.

e.g. Table 5, Sl.No.(1), (iii) define vertical geometric irregularity. If

the clause is interpreted in its stricter sense, pyramid type geometry

of particular dimensions can be shown as irregular structure.

A better way to address this problem is to incorporate the

guidelines on understanding and interpreting the mode shapes of

the structure. This will necessitate the 3-D model of the structure

and free vibration studies compulsory for all the cases. Probably to

avoid this, with a fear that it may be beyond the scope of average

professional in India or an attempt is made to oversimplify the

situation, in either case the purpose is not served.

8. Load Combinations

Clause 6.3.1.1.2 and 6.3.1.2.3 states the load combination of

1.7(DL ± EL) for Plastic design of steel structures and 1.5(DL ± EL)

for limit state design of RC structures respectively, along with other

combinations with imposed load. The combinations referred here

are the critical one in most of the cases as experienced by the

authors in routine design work. In case of the lateral load

estimation (i.e. earthquake load, EL) the seismic weight of the

building is estimated. This includes full dead load and partial

live/imposed load (25% or 50% as the case may be). The issue is

100

180

•Floor heights are not equal. •Important point is that the dimensions of adjacent storey can be in excess of 150% of storey under consideration.

why this partial live load is not reflecting in load combinations

those involve the earthquake loading? If this part of gravity load is

duly acknowledged in load combinations, the effect of lateral load

gets reduced resulting in lower design moments for the members.

The clause specifies 1.7(DL ± IL ± EL) or 1.2(DL ± IL ± EL) but this is

with full-imposed load, this makes some sense but the earlier

referred cases certainly need modifications.

9. Analysis Procedures

IS 1893 allows response spectrum method or time history analysis

for estimation of lateral forces for structures satisfying certain

limitations based on regularity of structure type, height and

location as per the seismic zone. A few certain types of irregularities

are allowed if within specified limits to get combined in with above

stated analysis methods. For remaining cases there is no guidance

so also is the case of deciding the necessity of nonlinear analysis. At

this point it should be noted that the response spectrum

recommendations are essentially forced based elastic procedures

and the time history procedure, which can be an alternate to it is

the linear time history analysis. This is contradictory with the aim

of IS 1893 which is to safeguard against the collapse through

inelastic deformations during a major earthquake. If life safety is

the target performance of the structure and damage (no matter how

much) is accepted in the design, the linear procedures become the

improper tools because linear behavior implies designing for no

damage condition/s.

Maffie (2000) and Vipul Prakash (2004) have discussed the

appropriateness of analysis methods in conceptual sense. It is

common to assume that, compared to the conventional design,

performance based design always requires a more complicated

structural analysis, such as a nonlinear static procedure, which

need not be true. In many instances, performance-based design is

used when better than life safety performance: either immediate

occupancy or some measure of damage control is desired. Nonlinear

procedures are actually less important to use for immediate

occupancy or damage control performance levels than they are for

life safety or collapse prevention performance levels. The reason is

that for the stricter performance levels such as immediate


163

occupancy, nonlinear response is limited, and thus an elastic model

can acceptably capture the behavior. Currently, in US, nonlinear

procedures are used more frequently for existing buildings than for

new buildings. This is somewhat illogical because many of the

existing buildings being evaluated have little ability to achieve

nonlinear deformations, whereas new buildings are intended to

have high ductility capacities.

It is necessary that IS 1893 incorporates the damage estimation

procedures, which are sensitive to displacement/strain type

quantities rather than forces or stresses. It can be concluded at this

stage that the desired performance objective would dictate the

analysis procedure for evaluation of the performance.

10. Basis of Design

IS 1893 allows Limit State Method for RC structures and Plastic

Method for steel structures. In both these methods, the calculated

loading is multiplied by certain safety factors to arrive at the final

design forces. The aim of these methods is to ensure that the

structure will not attain the limit state under consideration or will

not become unfit for use. However with earthquake loading it

becomes illogical to reduce the loading by a large margin e.g.

minimum of 6 times, reduction in case of RC structure, R=3 for

OMRF, during estimation of loading and then again increasing it by

[1.5 (partial safety factors for loading) X 1.17 (partial safety factor

for material: reinforcing steel) =] 1.75 during member design and

claim that it is designed for not attaining the limit state. This is with

the limit state of collapse, how one should check the limit state of

serviceability or the other limit states? No factors are specified for

these cases. This is obvious because in the first paragraph of this

paper the design criteria of IS 1893 is reproduced and it states that

minimum strength for DBE behavior is ensured but behavior under

MCE is only aimed at.

11. Need for Performance Based Engineering (PBE) Approach In

India

It has been realized that strength (concept of design base shear) is

not the only parameter that decides the extent of safety and the

extent of damage. Increase in strength need not lead to enhanced

safety and reduced damage. It is recognized that structure would

perform better when the distribution of strength is paid due

attention (Bertero). The reasoning of what led to development of

PBE in western countries is an eye-opening lesson for developing

countries like ours. It is necessary for the architects and structural

engineers in India to improve the gloomy picture in the field of

consultative professional practices. PBE can be considered to offer a

promising solution in this regard because of its inherent elements:

• PBE requires elaborate analysis and design procedures ensuring

multilevel performance

• Behavior of the building is defined in terms of measurable

performance characteristics, which are easy to understand, by the

building owners/users.

• Role of the client/statuary body, the architect, the structural

engineer and other concerned agencies can be clearly defined with

respect to the responsibility and the liability.

• Quality control procedures can be made more stringent and

realistic.

• Possibility of improving the professional relationship of the key

elements such as the owner/user→ architect/structural

engineer→ structural engineer/architect→ contractor. This will be

helpful in monitoring the post-built behavior of the building,

especially in the event of failure. The investigations would focus

on finding of what went wrong and learning from the failure

rather than finding a scapegoat.

Apart from this and advantages of this approach, there are other

aspects such as globalization. According to “The Agreement on

Technical Barriers to Trade (TBT): 1995” of the World Trade

Organization: When appropriate, technical regulations should

specify products in terms of performance rather than design or

descriptive characteristics. India is a member country of WTO since


165

01/01/1995. It is not far away that Indian Codes will have to

completely switch over to Performance Based Design Philosophy.

12. Historical Evidence for PBE in India

Normally it considered that PBE came from automobile

industry (Naiem F. 2001). As regards the civil engineering, the first

documents based on PBE were produced under FEMA sponsored

ATC 33 project in 1992. ATC, FEMA, SEAOC, NISEE Berkeley, and

many others have published a series of documents then after.

However Hamburger and Mohle claim that PBE in US dates back to

1927 and all the US codes developed so far are PBE based. Akiro

Inokuma (2002) mentions that PBE approach was first adopted in

building construction industry of Japan in 1963 itself. This may be

true. But then the basis on which Hamburger or Inokuma try to

justify their claim, can be extended to today’s Indian Codes. Not

only this, it can be traced much back (in B.C.) in Indian History. A

list of classics on building construction science

(STHAPATYASHAASTRA)* from Indian Literature is presented

below. The information contained in these granthas* was based on

the data collected over a period of time not less than few hundred

years or number of generations.

Table1: Famous Architects (STHAPATIS)* of Ancient India and Their

Published Literature

No. Name of the Architect Title of the Publication

1 Vishwakarma (Sthapati of Gods

or universe)

1 VishwaKarma

Vastushastra

2 Kshirarnava

3 Dynanprakash Deeparnava

2 Sage Mansara Mansara Shilpshastra

3 Maya ( Sthapati of Asuras or

Demons)

Mayamtam

4 Maharaja Bhojdeo Samarangan Sutradhara

5 Mandan 1 Rajvallabha Mundan

2 RupMandan

3 Devata Moorti Prakaran

4 Prasad Mandan

6 Kumarmuni Shilparatnam Vol. 1 and

2

7 Varaha Mihir Bruhatsanhita

8 Thakkar Pheru Vastusara

9 Jin Datta Suri Vivek Vilasa

10 Vidyadhara, MahaGovind Literature not available but the

structures built by them exist

still today.

* These are the devnagari words (in languages prevailing at that time e.g.

Sanskrit, Prakrit, etc.). The meaning appears prior to the brackets.

These classics cover very finer aspects and details of building

construction process, which was practiced as a science. Guidelines

and rules are listed with reference to:

• Duties of every individual, directly/indirectly associated with the

construction project.

• Selection of proper sites for construction and thereby

establishment of cities (town planning)

• Requirements of plot area and building area for individuals on the

basis of their social status.

• Construction planning on the basis of seasonal changes or

weather conditions

• Details of construction procedures for different forms of

structures and quality control.

• Requirements of structural details of members as well as for the

system as a whole.

• Penalty and bonus/reward clauses for contractors.

It is beyond the scope of this paper to discuss all these aspects in

detail. But it is worth mentioning about the last point. The penalty

clause was simple: tit for tat or blood for blood, if structure fails to

serve its intended purpose and causes loss to the owner, the

contractor (if found guilty) was liable for paying the compensation:

if owner/his family/users suffer physical disability the contractor

would face the same disability as a punishment this included even

the death sentence to contractor or his family. It may seem cruel

today, but it was the period of royal monarchy and the crime was

looked upon as a sin. The strict rules need to be considered as

indicative of the importance given to these activities. The aspect of

legal enforcement of the guidelines or rules (what we call it as a

code today) was properly addressed, the front on which we lack

even today. Further a careful look at the information presented here


167

would reveal that it contains all the elements needed for

performance based engineering and its implementation. Thus it can

be said that India is the originator of concept of PBE and it was in a

well-developed form in ancient India. All this was gradually lost

during the invasions by foreign rulers. India may be the only

country in the world to face the invaders from many other countries

and religions. Rather, this proves that there existed a rich,

advanced and established civilization.

13. CONCLUSIONS

The basic philosophy of drafting IS 1893 code has remained to be a

force based procedure and that the drawbacks mentioned, are

basically by-products of attempts to incorporate the features of

otherwise three dimensional and/or inelastic behavior of structures

in to two dimensional elastic procedures. In principle code has not

retained any method suitable for hand computations hence it would

be appropriate to insist the 3-D modeling and include the modeling

guidelines for analysis. It is necessary to provide the guidelines to

evaluate the nonlinear behavior. Evaluation procedures provide the

key to ensure the behavior: which is the core issue in PBE. IS1893

needs to be reformed in its future edition. Time tasted performance

of ancient Indian structures is a sufficient proof that designs based

on performance concept are not new for India.

References

[1]Applied Technology Council, “Tentative Provisions for the Development of Seismic Regulations for Buildings”, ATC 3-06, NBS Sp-510, NSF 78-8, N.B.S., USA (1982)

[2] Akiro Inokuma, “Basic Study of Performance Based Design in Civil Engineering”, Journal of Professional Issues in Engineering, Education and Practice, January 2002.

[3] Applied Technology Council, Redwood City, California: ATC-40 Report (1996): Seismic Evaluation and Retrofit of Concrete Buildings, Volume I

[4] Bertero V. V. “Performance Based Seismic Engineering; a Critical Review of Proposed Guidelines”, Seismic Design Methodologies for the Next Generation of Codes, Proceedings of the International Workshop, Slovenia.

[5] Dowrick D. J.,“Earthquake Resistant Design For Engineers and Architects”, 2nd

edition-1994, John Wiley and Sons Ltd. ISBN: 0 471 91503 3

[6] Farzad Naiem, “The Seismic Design Handbook”, Kluwer Academic Publishers, Massachusetts ISBN 0 7923 7301 4, 2001

[7] Federal Emergency Management Agency (FEMA): FEMA-273: 1997 [8] Hamburger R. O. and Mohle J.P., “State Of Performance Based-

Engineering In United States”, web: peer.berkeley.edu/moehle/papers/State_of_PBEE_in_US.pdf

[9] Hamburger R., Whittaker A. and et al, “The ATC-58 Project: Development Of Next-Generation Performance-Based Earthquake Engineering Design Criteria For Buildings”, Paper No. 1819, 13Th World Conference On Earthquake Engineering, Vancouver, Canada

[10] Jain S.K., “Review of Indian Seismic Code, IS 1893 (Part 1): 2002”, The Indian Concrete Journal, November 2003.

M.1) Maffei Joe (2000), “Suggested Improvements To Performance Based Seismic Guidelines”, 12th WCEE, Auckland, New Zealand, February 2000.

[11] Vipul Prakash (2004), “Whither Performance Based Engineering in India?” Journal of Indian Society of Earthquake Technology, Vol.41, No.1, March 2004.

[12] Vipul Prakash, Prajapati G. I. “Lecture-notes: EQ 512: Earthquake Resistant Design of Structures”, PG Course in Earthquake Engineering, IIT Roorkee (India) (Unpublished) (2004)

[13] The World Trade Organization (WTO) Agreement on Technical Barriers to Trade (TBT): Implications For Developing Countries, March 1995 No. 44


171

Earthquake Scenario of India and Its Relation to

Various Rock Types

Medha Jha

Assistant Professor, Department of Civil Engineering, Institute of


1. INTRODUCTION

The increasing number of natural disasters, together with the

increase in number of victims and significant impacts on socio -

economic infrastructure requires active development of prevention

and mitigation measures, to reduce the number of disasters

and / or reduce the damage to infrastructure, economics and social

life; better understanding of the mechanism behind natural hazards

is required. Geology has a central role in identifying areas of natural

hazard risk and recommending the appropriate mitigation

measures. Good understanding of all geological processes will lead

to better insight in how and what type of prevention or mitigation

measures should be taken.

Earthquakes are also notorious natural hazards and have

enormous impact on life and infrastructure. Understanding the

subsurface with respect to active and passive faults, stress and

strain are required for earthquake risk assessment.

1.1 Causes of Earthquake

Earthquakes are mostly associated with the Plate Boundaries.

There are three types of plate boundaries: Divergent, Convergent

and Transform. Movement and slipping along each of these types of

boundaries can form an earthquake Figure 1. Depending on the

type of movement, the earthquakes occur in either a shallow or

deep level in the crust. The majority of tectonic earthquakes

originate at depths not exceeding tens of kilometers.

In subduction zones, where old and cold oceanic crust descends

beneath another tectonic plate, “Deep Focus Earthquakes” may

occur at much greater depths (up to seven hundred kilometers!).

These earthquakes occur at a depth at which the subducted crust

should no longer be brittle, due to the high temperature and

pressure. A possible mechanism for the generation of deep focus

earthquakes is faulting. Earthquakes may also occur in volcanic

regions and are caused there both by tectonic faults and by the

movement of magma (hot molten rock) within the volcano. Such

earthquakes can be an early warning of volcanic eruptions.

Figure 1: Tectonic Setting of Earthquakes

2. Tectonic set up of India


173

Peninsular India constitutes one of the largest Precambrian shield

areas of the world (Figure 2). The Indo-Gangetic Alluvium Plain

(IGAP) separates the Himalaya to the north and the Peninsular

Shield to the south. The Shillong Plateau in northeast India

constitutes an outpost separated from the main shield by the

Bengal Basin and from the Himalaya by the Brahmputra River.

The Peninsular Shield of India is made up of three main cratonic

regions (Figure 2); the Aravalli, the Dharwar and the Singhbhum

which are separated by Proterozoic rifts and mobile belts. The major

prominent rifts that separate the southern and northern blocks of

the shield are the Narmada Son Lineament (NSL) and the Tapti

Lineament (TL), together called the Son-Narmada Tapti lineament

(SONATA). The other rift basins are the Kutch, Cambay, Godavari,

Cuddapah etc.

Figure 2: Seismo-tectonic map of India

The Himalayan region is very much associated with a high degree of

seismicity in comparison to that of Peninsular India, and making

the Himalayan region seismically more vulnerable to earthquake

damage (Zone V) than that of Peninsular region(Figure 3).

Figure 3: Seismic zonation within India.

3. Tectonic setting of Central India


175

Tectonically the earthquake-affected area of Central India

encompasses two crustal provinces of central Indian shield, namely,

the Northern Crustal Province (NCP) and the Southern Crustal

Province (SCP) (Acharrya and Roy, 1998). The two provinces are

separated by a Central Indian Suture (CIS, Jain et.al., 1995).The

southern part of NCP, containing the Satpura and Son Narmada

(SONA) valley geographic domains is known as Central Indian

Tectonic Zone, (CITZ; Radhakrishna and Ramakrishna, 1988). The

boundaries of CITZ are marked by Narmada North Fault (NNF) in

the north and CIS in the south (Acharrya, 1997). The main

earthquake affected area lies in SONA lineament zone, which forms

the northern unit of CITZ. The SONA zone is about 1600 km long

and 150 km – 200 km wide extending from the southern margin of

Kathiawar Peninsula in the west to the margin of Vindhyan basin in

the east (Crawford, 1978; Ahmed, 1964). The zone has been a major

centre of tectonism with evidences of reactivation. The E-W to ENE-

WSW trending Narmada and Tapti lineaments form a prominent

tectonic belt Son Narmada Tapti lineament (SONATA) in midplate

continental India Narmada tectonic line and its presumed eastward

extension, Son have been considered as a major Precambrian deep

crustal features (Auden, 1949; West, 1962) and possibly a paleorift

extending hundreds of kilometers in E-W direction (Mishra, 1999).

Correlation of structural and geophysical data shows that the Son

Narmada Tapti lineaments together represent an intraplate rift with

a central (Satpura block) horst bounded on either side by grabens;

the Narmada graben on the north and the Tapti graben to the south

(Mishra, 1999). The trace of the Narmada South Fault (NSF) was

noted in the Jabalpur area. The seismicity pattern of the

earthquake has a correlation with the ENE- WSW structural feature

of the terrain. The main shock of Jabalpur earthquake of May 22

1997 and its after shocks are interpreted to have generated as a

result of reactivation of the NSF at the crust mantle boundary

(Gupta et.al, 1997;Acharryya 1997; Acharrya et.al, 1998, Devarajan

et.al, 1998). Intraplate seismicity may be due to reactivation of

preexisting faults and stress concentration which may be caused by

lateral variation in crustal structure, density, lithologic boundaries

and stress concentrations along the edges of the structures; strain

on the other hand is concentrated along the faults and shear zones,

resulting in their reactivation. As regards seismicity in the Narmada

valley, the reactivation of faults or shear zones would be favoured

over new fault generation since the SONA fault is in a high shear

stress orientation.

4. Case Study of Jabalpur Earthquake

An earthquake of magnitude 6 rocked a large portion of the shield

area of Peninsular India in the early hours of May 22, 1997 around

Jabalpur, M.P. The epicenter of the shock was about 20 km E-SE of

Jabalpur at 23.08°N latitude and 80.06°E longitude and Focal

depth was 35 kms. As it was the summer time most people were

sleeping outside their houses, there were fewer fatalities. The

seismicity pattern of the earthquake has a correlation with the

ENE- WSW structural feature of the terrain. The Jabalpur

earthquake of 1997, fall under zone III of seismic zoning map of

India (IS: 1893 - 1984). The main shock of Jabalpur earthquake

and its after shocks are interpreted to have generated as a result of

reactivation of the Narmada South Fault (NSF) at the crust mantle

boundary. For classification of building types and vulnerability

classes and for establishing the damage grade guidelines provided

by Grunthal in 1993 was adopted.

Jabalpur area represents a complex of igneous, sedimentary and

metamorphic rocks. The analysis of the damage patterns was done

on the basis of seismic rigidity of these litho units and it shows that


177

the seismic rigidity characteristics of the litho-units have played a

main role in accentuating the damage. The Jabalpur earthquake is

an important event for India from the point of view of seismic

preparedness and expertise in repair of seismically damaged

structures.

4.1 Geology of the Area

Jabalpur is a historic town named after Rishi Jabali in the state of

Madhya Pradesh. Petrographically, this area comprises of all types

of rocks namely igneous eg granites and basalts, sedimentary eg.

sandstones, limestones, shales and clays and metamorphic rock eg.

marble, schist and gneiss. Stratigraphically it comprises of the

litho-units belonging to geological age from palaeoproterozoic to

tertiary lavas and recent compacted alluvium (Matley, 1921). Thus,

it allows a vision from recent to the remotest end of the beginning of

the geological time. (Table1). Geomorphologically, the area reveals

dominant imprints of structural control and lithological

differentiation and mixed topography, which is a combination of

plains, inselbergs, highlands, and trappean plateau. Narmada is the

main river that confines the area in the south. It occupies the

tectonically active linear valley on the face of the Peninsula (Project

CRUMANSONATA, 1995).

Table 1: Litho-stratigraphic Succession of the Area

Age Super group/

Group/ Formation

Lithology

4.2 Geotechnical assessment of the area

The foundation of construction in different areas of the town rests

over various lithounits, viz Madanmahal granite, Gondwana and

Lameta sediments which comprises of sandstones, clays and

limestones, Deccan trap basaltic flows and alluvium. Constructions

have also been made in reclaimed land fills. The analysis of the

damage pattern was based on European macroseismic scale- 1992

(updated MSK scale, Grunthal,1993), which uses the parameters of

degree of damage to man made structures of various vulnerability

Quaternary Recent Alluvium along river Narmada

and its tributaries

Lower Eocene to

Upper Cretaceous

Deccan trap Basalts

Upper Cretaceous Lameta group Sandstone, shale marl, Impure

cherty limestone

Lower

Cretaceous to Permo

-Carboniferous

Gondwana Super

group

Sandstone, clay, shale,

Conglomerate in basal part

Meso to

Neoproterozoic

Vindhyan Super

group

Essentially sandstone shale and

limestone

Palaeoproterozoic Madanmahal Pink porphyritic and Non-

porphyritic granite

Late Archaen to

Palaeo-proterozoic

Mahakaushal

Acid and Basic intrusive

Conglomerate, qua rtzite

quartz mica schist,

chert breccia, dolomitic

marble chert and quartzite

bands and amphibolites,

phyllite, metabasalt, quartz

schist, banded haematite

quartzite, quartzite and

amphibolites


179

classes seismogeological and hydrological changes. Jabalpur has

1.5 lakhs house of vulnerability classes A to D (Grunthal, 1993).

The ratio percent for class A: B: C is nearly 20:75:5. A few class C

structures are G+2 to G+4 storied with a maximum height of nearly

20 meters.

The analysis of the damage patterns shows that the seismic rigidity

characteristics of the litho-units have played a main role in

accentuating the damage. As originally defined by Reid (1908), the

seismic rigidity is the product of rate of propagation velocity of

longitudinal seismic Waves (Vc) and density. Seimogeological

changes are restricted to the units of low and moderate seismic

rigidities. Thus the structures over high seismic rigid terrain have

comparatively less damage as compared to those over low and

moderate seismic rigidity.

The areas, which are situated directly on the granite basement, are

Madan Mahal, Adhartal, Ghamapur, Shobhapur, Gokalpur. In

these areas, most of the houses were of class A and class B. Most of

the vulnerability class A houses and many class B types houses

with brick,mud mortar tiled roof have grade 1 damage. A few class

A houses have grade 3 damage. The less degree of damage in these

areas may be due to the high seismic rigidity of granite which

ranges between 13 -16.

The construction in the northwestern, south central and

southeastern parts of the Jabalpur are either over Gondwanas or

on the soil cover developed over Gondwanas with seismic rigidity of

9. Most of the class A and B and some class C structures have

grade 2-3 damage. Some class A and B buildings with grade 3-4

damage have also been reported. Many ground fissures were

noticed in these areas.

In the areas located on the Deccan traps with seismic rigidity (13

-16) with varying soil cover most of the class A houses with 2-4

damage; most of class B and a few class C houses with grade 1

damage and some class B houses with grade 2 damage have been

reported. The areas located on / near river banks and alluvium(of

varying thickness)includes Gwarighat, Jilherighat, Gauriyaghat

area . Maximum damage occurred in these areas due to low seismic

rigidity which ranges from 3-5.Most of the class A structures have

grade 2-5 damage, many class B type with grade 1-2 damage and a

few with damage upto grade 4 have been reported. In ancient times

Jabalpur was the land of tals and talaiyas . With the passage of

time these tals and tanks were filled and reclaimed. These areas

have vulnerably low seismic rigidity, which is less than1. Intensity

accentuation is also maximum in reclaimed zones with high

moisture content. The area mainly includes Marhotal, Ranital,

Gulowa chowk, Phoolnagar, Sharda Chowk, Gangasagar.Most of

the class A and B structures have grade 2-4 damage and some

class C have grade 5 damage and a few class A have grade 5

damage.

In addition to the damages to the houses gas emanations, landslips

and bank failure, ground fractures and changes in ground water

regime were also recorded. The gas escaping from the standing

water column on channel part of Narmada river was reported.


181

Emanations of gas charged with dust from ground fissures was

reported from many localities.

Extensive ground fissures were noticed in residual soils and

alluvium. Most of these were irregular in outline, varying in length

from 1m to as much as 50m. Alluvium and soil on reclaimed areas

as well as on Gondwana have poor cohesion. In Dhanwantri nagar,

enechelon ground fissures, predominantly trending ENE-WSW and

N-S were developed. In Supatal area, discontinuous, parallel

tensional fissures trending E-W and running for nearly 25m were

observed on the Chuikhadan hillock exposing Gondwana

sediments. A 3m long hairline crack trending WNW-ESE had

developed in the Sharda Chowk area on compacted earth.in Gulowa

Chowk area, E-W trending 3-5m long tensional fissures had

developed in soil. Ground cracks trending E-W and NW-SE

developed in the soil in the Nehrunagar (Medical college) area. In

Lalpur, on the banks of Narmada, conjugate hairline cracks

trending WNW-ESE and NE-SW were seen in semi-consolidated

alluvium. In Tilwaraghat, 2.5m long lunate ground cracks were

noted in compacted soil in the school playground.

Hydrological changes occurred both in surface and subsurface

water regimes. Strong agitation of standing water bodies

accompanied by mud churning emergence and disappearance of

natural springs occurred in surface water regimes. Changes in

colour, turbidity, taste and odour were observed in addition to

fluctuation in the water table in groundwater regime. The localities

over the Gondwanas sediments, alluvial tracts and the reclaimed

areas showed the maximum changes in the groundwater regime. In

the northeastern part of the city the people reported muddy water

with brownish black colour. In Supatal area, water became

brackish, laden with mud and was rendered viscous. In the

Shivnagar area groundwater became turbid and white accompanied

by rise in water table. In the area around Medical College appeared

in many dry bore wells. This water was reddish and moderately

turbid. In the southern part of Jabalpur town, turbidity remained

for 4 to five days after the earthquake. These changes may be due

to the less cohesive nature or loosening of the soils or formations

due to shaking.

5. CONCLUSIONS

Seimogeological changes are restricted to the units of low and

moderate seismic rigidities. Thus the structures over high seismic

rigid terrain suffered comparatively less damage as compared to

those over low and moderate seismic rigidity.The less degree of

damage in the granitic areas may be due to the high seismic rigidity

of granite which ranges between 13 -16.The construction in the

northwestern, south central and southeastern parts of the Jabalpur

is either over Gondwanas or on the soil cover developed over

Gondwanas with seismic rigidity of 9.Medium damage was reported

from these area.The areas located on / near river banks and

alluvium (of varying thickness)includes Gwarighat, Jilherighat,

Gauriyaghat area . Maximum damage occurred in these areas due

to low seismic rigidity, which ranges from 3-5.


183

REFERENCES

[1] Acharrya, S.K., Kayal, J. R., Abhinaba Roy and Chaturvedi, R. K.

(1998) Jabalpur earthquake of May 22, 1997, Constraints from

aftershock study, Jour. Geol. Soc. India.

[2] Acharyya, S. K. (1997) Jabalpur earthquake of May 22, 1997.

Journal Geol. Soc. India. V.50, No.3, pp. 375.

[3] Acharyya, S.K. and Abhinaba Roy (1998): Thermal-mechanical

history of Central Indian Tectonic Zone and Reactivation of major

faults – Chapman conference on SCR earthquake, NGRI,

Hyderabad.

[4] Ahmed, F. (1964) The line Narmada – Sona Valley. Curr. Sci. 33: pp

362 – 363.

[5] Auden, J.B., (1949) Dykes in western India – A discussion of their

relationship with Deccan traps , Trans. Nat. Inst. Sci., India, Vol 3,

pp 123 – 157.

[6] Crawford, A .R. (1978) Narmada – Son lineament of India traced into

Madagaskar. Jour. Geol. Soc. India, Vol. 19, No. 4, pp 144 – 153.

[7] Devarajan, M.K. et al (1998) Seismotectonic studies of Jabalpur

earthquake of 22 May, 1997. Indian Min. Vol. 50, No. 4, pp. 377-

396.

[8] Grunthal, G. (1993) European Macro seismic Scale, 1992. European

Seismological Commission, Luxembourg. pp. 1 – 79.

[9] Gupta, H. K., Chada, R. K., Rao, M. N., yana, D. L., Mandal, P.,

Ranikumar, M. and Kumar, N. (1997) Jabalpur earthquake of May

22, 1997; Jour. Geol. Soc. of India; 50, pp. 85 – 91.

[10] Jain, S.C., Nair, K. K. K., Yedekar, D. B., (1995) Geology of the Son

– Narmada – Tapti Lineament Zone in Central India. In: Project

CRUMANSONATA, Geol Surv. Ind. Spec. Pub.No. 10. pp. 1 – 154.

[11] Matley, C. A., (1921) The rocks near Lametaghat, Jabalpur district

Rec. Geol.Surv.India. Vol. 53 (2) pp.165 –169.

[12] Mishra (1999): Prediction, exemplified by Garm area of the

Tadzhik,SSR, Akad.Nauk. USSR Inst. Fiz. Semli, Moscow, pp 72 –

99.

[13] Radhakrishnan, B.P. and Ramakrishnan, M. (1988) Archaean –

Proterozoic boundary in India, Jour. Geol. Soc. India. V.32, pp. 263

– 278.

[14] Reid, H.F. (1998) The California earthquake of April, 1906, Report

of the state earthquake investigation commission, V 2. The

mechanics of earthquake. Carneige Institute. Washington.

[15] West, W. D. (1962) The line of Narmada – Son Valley, Curr. Sci.,

31. pp. 143 – 144.


185

The Effect of Dynamic Loading on Structural

Integrity Assessment

Debashis Khan

Assistant Professor, Department of Mechanical Engineering,

Institute of Technology, Banaras Hindu University, Varanasi, U. P.

1. INTRODUCTION

Now-a-days in the design of civil engineering structures, static

loads like dead weight, superimposed loads and environmental

loads (wind or waves) are considered together with the time varying

load or dynamic load. Previously, people used to evaluate the effects

of dynamic loading by use of an equivalent static load or by a

modification in the factor of safety value. It is very important to

consider dynamic loading effects in the construction of tall

buildings, long bridges under wind-loading conditions, buildings in

earthquake zones, any component subjected to vibrations due to

equipment or machinery, impulsive load produced by blasts etc. It

has been observed in literature that rapid loading of a structure can

come from a number of sources and it affects not only the

structural behavior but also may affect the material properties. In

reality, an inertia effect from dynamic load can source plastic

behavior. In many cases, dynamic loads give rise to high stress

levels near cracks and fracture takes place so rapidly that there is

insufficient time for large scale yielding to develop. With increase in

strain rate, it has been noticed that there is increase in yield stress

and ultimate tensile stress. Also under high loading rate, the

fracture toughness for cleavage fracture is reduced. Therefore, it is

very important to consider the effects of dynamic loading on

fracture in detail [1].

The structural engineering earthquake design community became

very upset after observing the effects of the earthquakes at

Northridge, California in 1994 and at Kobe, Japan in 1995. There

were widely spread fractures within welded steel moment resisting

frames. These frames have been originally designed to be strong

enough to resist the stresses and also ductile enough to

accommodate the distortions generated by a severe earthquake.

There are many such examples of brittle fractures which occurred

at the connection between the beams and columns at lower load

and deformation. These failures have encouraged the engineering

community to investigate the reason and as part of this to explore

alternative connection types. A great deal of research and

laboratory testing have been carried out in order to identify better

moment connections for new steel moment connection in buildings.

Failures in such kind of engineering structures made of materials

with high toughness and low strength may occur due to pre-

existing flaws/ defects or through nucleation of crack and its

subsequent growth into the defect free regions with disastrous

consequences to human lives, often involving large scale financial

loss. It is therefore essential to characterize quantitatively the

residual strength of material in the presence of cracks, as the

presence of a crack reduces the structural strength [2].

Currently, various numerical techniques like finite element,

boundary element method are used to study the local behavior of

connections with defects within a complete building frame under

the dynamic loading. The purpose of the numerical analyses is to

identify the effects of crack length, connection design, and material

properties on the local behavior of sub model connections located in

the full steel frame building under such loading. The stress

distribution in the region of the column and beam flange connection

is also considered. Over the years, the development of methodology

and criteria for accurate failure prediction has been the focus of

quantitative fracture mechanics, which is based on the energetic

concepts correlated to crack extension.


187

2. THE CONSEQUENCE OF DYNAMIC LOAD ON THE

PROPERTIES OF STRUCTURAL STEEL

In structural design the influence of dynamic load on material

properties is often ignored. However, it is a fact that an inertia effect

from dynamic load can cause plastic deformation. From the

research work in the last few decades, it is known that increase in

loading rate affects the material properties of steel. Normally, the

quasi-static tests of yield stress are conducted at low strain rates.

Under seismic loading conditions for short periods the local strain

rates in structures may be causing increase in yield stress of 30%

[1]. It was investigated by Manjoine that the lower yield stress and

ultimate tensile stress are increased with increase in strain rate [3].

In a separate study it was shown by Campbell and Cooper that the

fracture strain decreases with increasing strain rate [4]. This

implies that the material becomes more brittle when the strain rate

increases. Cowper and Symonds suggested the following important

relationship of strain rate, static flow stress and dynamic flow

stress [5].

�̇ = !"#$"# − 1&'̇, ()* ≥ ()

Where ()′ is the dynamic flow stress at a uni-axial plastic strain

rate �̇, () is the static flow stress. D and q are constant for a

particular material. Figure 1 represents dynamic uni-axial tensile

tests on mild steel at various mean plastic strain rates

Figure1. Dynamic uni-axial tensile tests on mild steel at various

mean plastic strain rates. A: �̇ = 106,-.; B: �̇ = 55,-.; C: �̇ = 2,-.; D:

�̇ = 0.22,-.; E: �̇ = 0.001,-.. 1 unit of ordinate is 6.895 MPa [1].

In a separate study Wakabayashi et. al. [6] found smaller dynamic

enhancements than those given by Symonds [7] from tests which

resulted in the following expression:

/01/0 = 1 + 0.0473234 5 �̇�̇)6

Where �)̇=50 x 10-6 s-1

In Figure 2, the curve corresponding to k = 1 represents a very

limited ductility and k = 10 represents fully ductile behavior.


189

Figure 2. Dynamic enhancement of yield stress of steel as a

function of period of vibration and maximum strain reached [1].

3. THE INFLUENCE OF LOADING RATE ON THE FRACTURE

PROPERTIES OF STEEL

The fracture toughness of structural steels under dynamic loading

without significant effects of stress wave normally increases with

decreasing loading rate and increasing temperature, as shown in

Figures 3 and 4.

Figure 3. The effect of temperature and loading rate on KIc [8].

Figure 4. The effect of loading rate on KIc [8].

Generally loading rate is proportional to strain rate which in turn

implies that the material cleavage fracture toughness decreases

with increasing strain rate. In 1987, Barsom and Rolfe classified

the loading rate for fracture analysis and testing of steel into three

categories, as shown in Table 1, [8].

Table 1 Type of load, strain rate and example


191

Type of Load Strain rate Example

Static and Quasi-static < 10-5 Steady-state, creep and

Relaxation

Dynamic 10-2

– 10-1

Traffic, Machinery

10-1

- 101

Earthquake, Crane

101 - 10

6 Explosion, Blast load

Barsom and Rolfe [8] also presented typical results of Charpy V-

notch impact tests which are shown in Figure 5. It is being

observed that brittle to ductile transition behavior occurs at lower

temperature for slow loading tests compared to dynamic loading

tests. The rate of change of energy absorbed in the dynamic loading

test is higher than that for the slow-bend test. The brittle-to-ductile

transition temperature is quantified in terms of a temperature shift.

In the region of temperature shift, loading rates reduce the fracture

toughness rapidly and increase the propensity for brittle fracture of

the steel. However, at temperatures lower than the transition

region, the loading rate does not have much effect. In the upper

region, dynamic loading tends to increase the toughness behavior of

steel. Also, at high loading rates, the local temperature at the tip of

a crack may increase due to absorption of energy from local plastic

work thereby affecting the fracture toughness but this effect is

automatically taken into account in dynamic fracture toughness

tests at the appropriate loading rate [1].

Figure 5. The temperature shift in CVN and upper-shelf level due to

strain rate [8].

4. FRACTURE PARAMETERS FOR EVALUATING THE CRACK

TIP SEVERITY

Contrary to the advances in static fracture mechanics, significantly

fewer reliable facts with established criteria and solved problems

(for both stress intensity factors and J integral) are found in

dynamic fracture mechanics. The two kinds of problems which are

generally dealt under the domain of dynamic fracture mechanics

are fast fracture mechanics and impact fracture mechanics. Fast

fracture mechanics treats various behaviors of fast propagating

crack tips such as growth initiation, propagation, arrest, kinking

and curving, branching etc and in such conditions the effects of

crack velocity play significant roles. On the other hand, impact

fracture mechanics deals with various fracture behaviors under

impact or dynamic loading, wherein the effects of material inertia

and stress wave interactions play significant roles, [9], [10] – [17]. In

the case of a sudden or impact loading, fracture can occur

unexpectedly, which may be the main concern for a failure analysis.


193

Since the work of Eshelby, Cherepanov and Rice [18] – [20], the

subject of the so-called path-independent integrals has received

much attention both in linear elastic and more complex nonlinear

elastic plastic fracture mechanics applications due to its many

advantages. In this case, the Eshelby-Cherepanov-Rice J integral

has played a very significant role in the advancement of static

fracture mechanics. From the theory and computational point of

view, the static J integral comprises the salient features like it has

the physical meaning of energy release rate; it has the property of

path independence; and it can be related to the stress intensity

factors by shrinking the integral path to the crack tip, [9]. A local

value of the strain energy release rate for non-linear elastic

material, denoted J, is given by:

Where W is the strain energy per unit volume, 7 = ∫ (9:;) <=9:, Ti are

components of tractions and ui are the components of

displacements. The crack extends along the x-axis, and s is the arc

length along an arbitrary contour traversed counter clockwise from

the lower face of the crack around the tip to the upper face.

The J-integral defined by above equation is valid for most types of

monotonic loading for elastic–plastic material. In the dynamic case,

the J-integral is not path independent due to the presence of

material inertia in the vicinity of the crack. Also if unloading occurs

for elastic–plastic material, the J-integral will no longer represent

strain energy release rate. Therefore the calculation of the J-integral

under dynamic loading for non-propagating cracks should be

developed by including the kinetic energy density of material at the

crack tip in the same manner as the strain energy density [10].

Thus

where W and T are the stress–work density and kinetic energy

density per unit volume at t = 0; Γ is a vanishingly small contour

which lies in the principal normal plane at s, and n is the unit

vector normal to Γ. Pij denotes the non-symmetric first Piola–

Kirchhoff stress tensor which is work conjugate to the displacement

gradient expressed on the t = 0 configuration >?@>AB, i.e., the stress–

work rate is simply Pij>?@>AB per unit volume at t = 0. All field quantities

are expressed in the local orthogonal coordinate system, X1–X2–X3,

at location s on the crack font.

In many monotonic loading conditions, the calculation of J-integral

can be carried out using general purpose finite element programs

such as ANSYS/ ABAQUS provided crack tip inertia effects are not

considered. The J-integral formulation is similar to the static case

as shown in Rice’s J-integral. Due to the omission of the inertia

effect term in the standard J-integral formulation, the J-integral

option in ANSYS/ ABAQUS cannot in general be used for dynamic

conditions. In order to evaluate the integral an alternative approach

is to develop a post processing code which will use the stress-strain

data once finite element stress analysis by the package is done.

Extensive research for developing the finite element software for

calculating dynamic non-linear fracture of solids has been carried

out by Dodds et al. [15] leading to the introduction of a program

called WARP3D. This WARP3D software is mainly used to analyse

3D solid models subjected to static and dynamic loads. A general J-

integral computation facility (with inertia, face loading, thermal

loading, anisotropic materials) is also included in WARP3D. If there

is no sign of inertia effects WARP3D will calculate J value by using


195

the conventional J-integral formulation. When inertia effects are

present, it will calculate separately the static and dynamic terms.

An alternative method of defining crack tip severity is to use crack

tip opening displacement (CTOD) values.

5. CONCLUSIONS

From the current review work it is clear that the dynamic load can

affect structural behavior, material properties and fracture

toughness of structural steel to a great extent. An overview of

various numerical techniques which can handle the influence of

dynamic loading in tall building, bridges, various equipments

subjected to impact load has also been presented. However, it is

being advised that results generated from the above mentioned

software are to be compared with the available analytical solutions

to validate the methodology and accuracy.

6. REFERENCES

[1] Kuntiyawichai, K. and Burdekin, F.M., Engineering assessment

of cracked structures subjected to dynamic loads using fracture

mechanics assessment, Engg. Fract. Mech., Vol. 70, pp. 1991 –

2014, 2003

[2] Khan Debashis and Biswas K. Circular arc crack under dynamic

load: a generalized approach for energy release rate, Int. J. Fract.,

Vol. 141, pp. 27-35, 2006

[3] Manjoine MJ. Influence of rate of strain and temperature on

yield stresses of mild steel. J Appl Mech 1944;11:211–8.

[4] Campbell JD, Cooper RH. Yield and flow of low-carbon steel at

medium strain rates. In: Proceedings of the conference on the

physical basis of yield and fracture. Institute of Physics and

Physical Society; 1966.

[5] Cowper GR, Symonds PS. Strain hardening and strain-rate

effects in the impact loading of cantilever beams. Report No. 28,

Department of Mathematics, Brown University; 1957.

[6] Wakabayashi M, Nakamura T, Iwai S, Hayashi Y. Effect of strain

rate on the behaviour of structural members subjected to

earthquake force. In: Proceedings of the eighth world conference on

earthquake engineering, San Francisco, vol. IV; 1984.

[7] Symonds PS. Viscoplastic behaviour in response of structures to

dynamic loading. In: Huffington, editor. Behaviour of materials

under dynamic loading. New York: ASME; 1965.

[8] Basom JM, Rolfe ST. Fracture and fatigue control in structures

applications of fracture mechanics. Englewood Cliffs, NJ: Prentice-

Hall; 1987.

[9] Nishioka T, On the dynamic J integral in dynamic fracture

mechanics. FRACTURE: A Tropical Encyclopedia of Current

Knowledge (Dedicated to A. A. Griffith), Edited by G. P. Cherepanov,

Krieger Publishing Company, Melbourne, USA, pp. 575–617, 1998

[10] Nakamura T, Shih CF, Fround LB. Analysis of a dynamically

loaded three-point-bend ductile fracture specimen. Engng Fract

Mech 1986;25:323–39.

[11] Kanninen MF, Popelar CH, Advanced fracture mechanics.

Oxford University Press, New York, 1985

[12] Nakamura, T., Shih, C.F., and Freund, L. B., Three-

dimensional transient analysis of a dynamically loaded three-point-

bend ductile fracture specimen, ASTM STP 995, Vol. I, American

Society for Testing and Materials, Philadelphia, pp. 217 – 241, 1989


197

[13] Freund LB, Dynamic fracture mechanics. Cambridge University

Press, Cambridge, 1990

[14] Guz, A. N. and Zozulya, V. V., Problems of dynamic fracture

mechanics without contact of the crack faces, Int. Appl. Mech., Vol.

30 (10), pp. 735 – 759, 1994

[15] Dodds RH, Gullerud A, Koppenhoefer K, Ruggieri, Warp3d-

release 13: 3-D dynamic non-linear fracture mechanic analysis of

solids using parallel computers and workstations. Structural

Research Series, 607, UILU-ENG-95-2012, University of Illinois at

Urbana-Champaign, 1999

[16] Zaho, W and Burdekin, F.M., Dynamic structural integrity

assessment for offshore structures, J. Offsh. Mech. Arct. Engg.,

ASME, Vol. 126, pp. 358 – 363, 2004

[17] Anderson, T. L, Fracture mechanics: fundamentals and

applications, CRC Press, Taylor and Francis Group, Boca Raton,

USA, 2005

[18] Eshelby, J New York.. D., The continuum theory of lattice

defects, Solid State Phy., 3, Academic Press, New York, 1956

[19] Cherepanov, G. P., Crack propagation in continuous media,

Appl. Math. Mech., 31, 3, pp. 467 – 488, 1967

[20] Rice, J. R., A Path independent integral and the approximate

analysis of strain concentration by notches and cracks, J. Appl.

Mech., vol. 35, pp. 379-386, 1968

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