lessons learned from the northridge earthquake

ELSEVIER PII: S0141--0296(97)00031-X

Engineering Structures, Vol. 20, Nos 4-6, pp. 249-260, 1998 © 1997 Elsevier Science Ltd

All rights reserved. Printed in Great Britain 0141~)296/98 $19.00 + 0.00

Lessons learned from the Northridge earthquake Duane K. Mil le r

The Lincoln Electric Company, 22801 St Clair Avenue, Cleveland, OH 44117-1199, USA

The performance of welded connections in special moment resisting frames (SMRF) during the Northridge, California earthquake in January 1994 was influenced by many variables. The actual ground motions experienced by the buildings, the design details employed in the connections, material properties, fabrication workmanship and inspection practices. The evidences of workmanship-related problems were primarily the result of lack of conformance to the requirements specified in the AWS Structural Welding Code D1.1. Lack of emphasis on in-process visual inspection may have permitted questionable fabrication practices to go undetected. A fundamental expectation regarding the performance of the SMRF is that of 'ductility'. In concept, the seismic energies are to be absorbed by the formation of plastic hinges within the structure. The pre-Northridge connection detail is examined in light of quality, fracture and multi-directional stresses, and its relative performance regarding expectations of ductile behavior. Post- Northridge research specimens are examined with respect to the same factors, along with their relative level of ductile behavior. Details that enhance ductile behavior and the reasons that this behavior can be expected are presented. © 1997 Elsevier Science Ltd.

Keywords: welded connections, seismic design, ductility

1. Introduction

The performance of welded connections in special moment resisting frames during the Northridge, California, earthquake of January 1994 may have been influenced by many variables, including:

• ground motions • design assumptions and details • materials properties • workmanship • inspection practices

All of these will be reviewed, but this paper will focus on the design expectations regarding ductility, why these expectations were not met and how ductility can be enhanced.

2. Background

On January 17, 1994, a fairly moderate earthquake struck a northwestern suburb of Los Angeles. Suddenly, the city of Northridge became well-known to structural engineers

around the country and, indeed, the world. The earthquake struck at 4:30 A.M. on a Monday morning. The early morning time frame helped to minimize loss of life. The damage was widespread, ranging from homes, apartments, high- ways and bridges, to utilities, gas lines and buildings. The three-story apartment buildings that sustained the most severe damage typically consisted of a ground level garage, second-story living quarters and third-story bedrooms. The lack of lateral bracing around garage door openings may prove to be the most significant structural deficiency. Steel bridges performed excellently and there have been no reports of significant damage to these structures. Pre- stressed concrete bridges performed largely as expected and those that had received retrofit improvements since previous earthquakes performed well. Those that had not been retrofitted collapsed, as would have been expected. Reinforced and pre-stressed concrete parking garages perhaps were the most problem-prone structures, resulting in total collapse in many situations. Reinforced concrete buildings were significantly damaged and some experienced collapse. Yet, amidst all the destruction surrounding the area, steel structures stood. There were no deaths associated with steel-framed buildings, no structural collapses

249

250 Lessons learned from Northridge earthquake: D. K. Miller

and, in general, only minor nonstructural damage to steel framed buildings. First reports were that steel structures were unscathed and that as in past earthquakes, steel structures had outperformed those constructed of alternate materials.

Three years after Northridge, the most important facts about steel-framed buildings remain unchanged: no deaths, no collapses. Even the initial reports of minimal nonstructural damage continued to hold true. Unfortunately, closer inspection after the earthquake has revealed damage to the beam-to-column connections in over 100 steel-framed buildings in the area. The apparent inconsistency between the structural damage and minimal nonstructural damage will, perhaps, supply some insight into why these structures sustained the amount of damage they did.

2.1. Special moment resisting frames The system employed in the majority of buildings in Northridge was the special moment resisting frame (SMRF). This approach uses no diagonals, but rather relies on the relative strength of the beam as compared to the column to develop a 'strong column, weak beam' relationship. As the structure is subjected to lateral forces, the rec- tangular panels in the structure tend to form parallelograms. However, the connection is assumed to be strong enough that the lateral forces will be absorbed in the floor beam, causing displacement to occur in this region. It is possible to design a building to utilize either a few very massive moment connections per floor, or to replicate a smaller detail throughout the structure. In either approach, the concept is the same, namely, the floor beam will be the member that experiences elastic and in-elastic (plastic) deformations should the seismic forces become great enough. The SMRFs, however, did not behave as expected during the Northridge earthquake.

2.2. Damage to the SMRFs A pattern quickly emerged among SMRF buildings that were damaged in the earthquake. The damage was typically confined to the lower flange-to-beam portion of the connection and the top beam flange-to-column flange remained intact (see Figure 1). In some cases, the bolted shear tab experienced sheared bolts, tears through the tab between the bolt holes or tears of the fillet weld from the column face. This type of damage to the shear tab, however, occurred only in the presence of damage to the bottom

Figure 1 Typical beam-to-column connection for SMRFs. The boxed area is where damage typically occurred.

flange. The top flanges generally remained intact, this was attributed to the influence of the slab which generated some composite strengthening to the connection.

The nature of the fracture to the bottom flange, however, varied from structure to structure, and even within a specific building. Figure 2 illustrates eight types of cracks that have been observed in buildings. Of the eight, seven have their point of initiation at the intersection of the bottom side of the bottom flange to the column flange. This is in the area where the fusible steel backing intersects the column and beam. The very specific delineation between fracture types was made in order to classify the damage prior to repair. Depending on the nature of the damage, the approach to repair varies.

Type A cracks initiate at the region of the backing bar where the beam flange and the column flange come together at a 90 ° angle. The cracks initiate at the root of the weld, following the zone immediately between the weld and the column flange material, or propagating into the weld metal itself. A Type A crack is defined as extending upwards less than one-half the beam flange thickness.

Type B cracks are similar to Type A, except by definition, they extend more than half way through the bottom flange. However, they do not exit to a surface. Again, the only reason for distinguishing between these two types of cracks is that the repair procedures employed are slightly different between the two details. From the underside of the beam, Type A and B cracks are evident in the separation of the backing from the column, as shown in Figure 3.

Type C cracks can be viewed as an extension of a Type B with a crack exiting at the weld toe. An example of Type C from an actual structure is shown in Figure 4. For pur- poses of definition, the exit point is at the weld face, weld toe, or within 1/4 in of the weld toe, or within 1/4 in of the weld toe in the column flange. These cracks typically follow immediately along the fusion zone. At first glance, this crack type has the general characteristics associated with either lamellar tearing or underbead cracking. A cross- section of an actual Type C is shown in Figure 5.

Type D cracks begin like type As, but the crack turns into the column flange material, and exits well above the toe of the groove weld. The fracture is clearly in the base metal. As viewed from the end of the floor beam, looking toward the column face, Type D cracks usually occur on either side of the web, creating two flattened arch-like exit points that can be observed, see Figures 6 and 7. In the center of the face of the column flange, the Type D may become a Type C.

Type E cracks are very similar to Type D, but they are buried within the column flange and do not exit to a free surface. Type E cracks can only be positively identified with ultrasonic (UT) inspection. During repair of damaged structures, it has been found that fractures expected to be Type A or B are frequently Type E (see Figure 8).

Type F cracks are similar to Type E, but the crack moves into the column flange, and may exit on the interior face of the column (see Figure 9). The exit point is typically at the toe of the weld that joins the continuity plates to the inside of the flange, there are some reports of the exit point being below the continuity plate, although this is the excep- tion.

Type G cracks are akin to Type F, but the crack con- tinues into the column web, as shown in Figure 10. Perhaps the most disturbing of all crack types in some structures, there are Type G cracks that initiate at either side of a col-

Lessons/earned from Northridge earthquake: D. K. Miller 251

~ T ~ p e A Type B ~ Type C

~ Type F

i~: ~i~m~ii~

~pe D

I Type H

Figure 2 Crack Types

Figure 3 Backing bar has separated from the column face, indi- cating Type A or B

Figure 5 A Type C crack from an actual structure. Notice the fracture does not intersect the weld metal, but follows the fusion zone

Figure 6 Type D crack. On the opposite side of the web, the arch-like contour is mirrored

Figure 4 Type C crack from earthquake damaged building

umn, resulting in complete horizontal fracturing of the column. Of course, the columns are in compression and some lateral stability is offered by the presence of the deck. The final type of crack is identified as Type H. It initiates at the tow of the weld, or at the intersection point of the weld access hole and the bottom flange of the floor beam. The crack may go through the weld metal, the base metal, or a combination of the two.

3. P o t e n t i a l contributing variables

3.1. Ground motions

Northridge was not the 'big one' - - actually, it was not even close. It measured only 6.8 on the Richter scale, mak- ing it moderate by those standards. However, the Richter scale measures total energy released during the earthquake. Northridge was a very short event, with the greatest per- centage of the activity taking place in a 6 s period. Rather than a gradual shaking, the community of Northridge experienced a severe shock or impact. Very high accelerations were reported. In one case, 1.8 g in the horizontal

2 5 2 Lessons learned from Northridge earthquake: D. K. Mil ler

Figure 7 Type D cracks pull a piece of the column material from the flange

Figure 10 Type G cracks extend into the web region. The f loor beam is on the left

Figure 8 Type E cracks do not extend out of the column

Figure9 Type E crack from an actual building. During repair, the backing was removed, the crack was gouged away, and a Type E was discovered where initially, it had been thought that a Type A or B was present

direction and 1.2 g in the vertical direction. These structures experienced 'uplift' and, unlike other seismic events, the vertical and horizontal accelerations were in phase one with another. The characteristics of the ground motions experienced at Northridge are beyond the scope of this paper. However, the (apparent) success of the connection detail pre-Northridge suggests that the ground motions experienced during the Northridge earthquake may have had unique aspects that are not yet fully understood. Possible ground motion similarities between Northridge and the Kobe earthquake (which occurred exactly one year later) support the need for more extensive evaluation of this variable.

4. Design

4.1. Design expectations

The basic design expectation regarding the performance of the SMRF is that the material property of ductility will enable the frame to deform inelastically, absorbing imposed seismic energy in the process. This deformation results in plastic hinge formation, the product of concentrated forces acting on ductile material. Most designs expect these hinges to form in the beams (as opposed to the column panel zones) and all designs expect the connection to remain undamaged. In Northridge, however, the hinges did not form and the connections were damaged.

Even though the SMRF systems have been used exten- sively, their variability in performance was well documented in previous research. Because the fractures associated with the testing of research specimens were in the vicinity of the weld, welding practices had been frequently questioned by the researchers as well. Perhaps a more rigor- ous examination of failed laboratory test specimens could have circumvented the need for the reexamination of the connection that is taking place at this time. The issues were well-documented in the open literature, however, which served as an eerie foreshadowing of some things to come. Northridge provided the real life laboratory that duplicated some of the laboratory experiences. The larger sized members that had become a commonly used configuration in Northridge had never been subject to laboratory testing. Size effects, particularly as they relate to triaxial stresses, resulted in differences in behavior between laboratory results and real-world behavior.

4.2. Connection details The moment connection of interest utilizes a shear tab that is shop welded to the column. In the field, the beam web is bolted to this shear tab, facilitating alignment and erec- tion of the member. The top and bottom flanges are field welded. The joint detail typically consists of an AWS D1. I prequalified complete joint penetration (CJP) groove weld detail, namely a TC-U4a. Most field contractors utilize a root opening of 9 mm and an included angle of 30 °. The D1 Code requires the application of 'weld tabs' to either end of the joint to facilitate quality deposition of weld metal across the entire joint, A weld backing is placed under the weld joint to support the molten weld metal. The D1 Code

Lessons learned from Northridge earthquake: D. K. Miller 253

requires this weld backing to be thoroughly fused by the weld metal.

For 'static' structures, the D1.1 Code permits weld backing to be left in place, while it is required to be removed from 'dynamic' structures when the backing is perpendicular to the applied load. Most buildings, even those in seismic zones, were considered 'static', primarily because the dynamic criteria in the Code was directed toward fatigue- prone structures (e.g. bridges). In T-joints subject to bend- ing moments, the naturally occurring unfused surface of the weld backing that lies parallel to the column face acts as a stress raiser, particularly when the weld root is placed in tension. This condition occurs in SMRFs due to the lateral seismic forces.

Depending on the ratio of Zf/Z (Z is the beam's plastic section modulus and Zf is the plastic section modulus of the beam flanges only), supplemental welds of the tab to the beam web may be required. The moment connection, therefore, consists of welded flanges, bolted web and, in some cases, supplemental fillet welds of shear tab to web. This connection, as indicated before, has shown widely varying performance characteristics under laboratory conditions. Significantly better performance has been seen when the web is directly welded to the column. The improved performance has been attributed to the better transfer of the moment capacity of the web through this welded connection. Most designs have assumed the moment capacity of the beam to be transferred through the flanges, although there may be substantial web moment capacity depending on the Zf/Z ratio. If this force is not transferred through the web connection, additional forces are passed through the fl~tnges.

To facilitate welding, weld access holes are provided in the beam web. These holes are further discussed below. They affect the design because the beam section properties are reduced by the elimination of material in that area. The greatest applied moment (due to lateral forces) will occur at the column face. This moment must be resisted by the beam's section properties, which are lowest at this location. The hinge should therefore occur in this region - - at the location of the weld. Rather than ductile stretching, however, brittle fracture occurred.

4.3. Materials propertie~ Base metal properties of interest include the yield strength, tensile strength, relative strength of beam steel to column steel, the through-thickness strength properties of the column steel, and base metal toughness. Also, the heat-affected zone (HAZ) properties are of interest (by definition, base metal). The design assumptions regarding SMRF's assume inelastic strain will be concentrated in the beam, ultimately resulting in the formation of plastic hinges. This necessitates a 'weak bean~, strong column' relationship. In part, this is accomplished by specifying ASTM A36 steel for the beams and A572 Gr50 for the column. ASTM specifications are not tight enough, however, to ensure that the material in the beam will be lower in strength than that in the column. Since the weld metal is generally higher in strength than either the beam or the column, the 'weak link' in the system can be the fusion zone in the column.

ASTM specifications for A36 control miinimum yield strength (but not maximum) and the tensile strength range. ASTM A572 Gr50 also controls minimum yield strength, but only minimum tensile strength. Therefore, as permitted by ASTM specifications, A36 beams could, and often did,

have yield and tensile strengths that exceeded the A572 Gr50 columns.

Some examples of low toughness base metals have been reported, but these appear to be exceptions. However, there is increasing evidence of very high heat input welding procedures being used, exceeding 4 KH/mm in some cases. This, coupled with high interpass temperatures (over 300 ° C), may have significantly degraded HAZ toughnesses. The use of high interpass temperatures and heat inputs probably reduced the yield and tensile strength of some welds as well.

Weld metal notch toughness is another area of concern. Applicable pre-Northridge codes did not specify minimum toughness levels, permitting the use of a commonly selected self shielded flux cored electrode, E70T-4, that does not have minimum toughness requirements. Actual welds commonly had toughnesses less than 15 J at 20°C and some less than 7 J. This low toughness, combined with details such as left-in-place weld backing and variable quality of workmanship, offered little resistance to brittle fracture, when subjected to seismic loads.

4.4. Workmanship It is possible to weld across the full width of the top flange without interruption. The weld that attaches the bottom flange of the beam to the column is more difficult to make, because the beam web prohibits the deposition of a continu- ous weld along the flange width. The welder is required to extend the electrode through the weld access hole and travel to the edges of the flange, terminating the weld on the weld tab. Properly sized weld access holes (frequently called 'rat holes') facilitate adequate access and visibility for quality welding. The D 1.1 Code prescribes certain minimum weld access hole dimensions. However, these minimum dimensions may be too small for some applications. It is imperative that the welder be given ample space to facilitate the deposition of quality weld metal. Detailers must specify the dimensions of weld access holes to facilitate quality workmanship. In at least one building, generously sized weld access holes were utilize, but the shear tab was so long that it interfered with the access hole.

Post-earthquake evaluation of the damaged connections revealed a host of workmanship related problems, including slag influsions, lack of fusion, inadequate penetration (to the weld backing), inadequate 'tie-in' between weld beads, poorly prepared weld access holes and the use of unauthor- ized 'weld dams' instead of the code required weld tabs. These problems were generally concentrated in the portion of the weld directly under the beam web. This is the area where one would anticipate the greatest number of diffi- culties. It is imperative that welders be specifically trained to deposit sound weld metal along the entire length of the joint.

A common source of many weld quality problems was the deposition of large, thick, wide weld passes. This is the direct result of high deposition weld procedures that utilize slow travel speeds, resulting in high heat input levels. The proper method to control the practice is either heat input control or (as is done in the DI.1 Code for prequalified procedures) control of weld bead sizes. Examples exist where welds are more than 12 mm thick, twice the 6 mm limit imposed by the Code.

The resultant weld quality was affected in three ways: (a) the area of soundly fused metal was diminished, increasing local stress levels; (b) harmful notches were


introduced, reducing the fracture resistance of the connection; and (c) the properties (toughness) of the weld and HAZ were reduced, further decreasing the fracture resistance of the connection.

4.5. Inspection practices The primary method used by the D 1.1 Code to ensure quality is through in-process, visual inspection. While ultrasonic inspection (UT) is a valuable tool to verify weld soundness, it cannot determine if critical welding variables have been properly controlled. These variables include preheat and interpass temperatures, welding amperage, voltage, travel speed, electrode extension, polarity, and bead sequence. UT cannot verify that the proper electrode was employed. In- process inspection is essential to ensure weld quality.

Unfortunately, current inspection practices dispro- portionately emphasize NDT and often overlook the importance of visual inspection. Specifications other than D I.1 appear to further this trend by providing specific UT requirements, but failing to require visual inspection requirements. The situation is further complicated by the specific geometry associated with the SMRF that makes it impossible to fully inspect the bottom beam flange weld because of the presence of the beam web.

When these joints are subject to UT inspection, the center of the length of the weld is difficult to inspect because of the interference of the beam web, and the radius between the web and flange. Unfortunately, this region is also the most difficult to weld area, leading to a most unde- sirable situation: the most problem-prone region is the least inspectable. In addition, interpretation of UT results of groove welds in T-joints with weld backing left in place is very complicated. Root defect signals are difficult to separ- ate from naturally occurring reflections from the weld backing.

5. Ductility

In the damaged structures of Northridge, there was only rare evidence that plastic zones actually were formed. Rather, the seismic energy was passed directly to the connection, overloading it and causing it to fracture. The area under the curve of a stress-strain diagram represents the total energy absorbed. It is essential that yielding take place to have significant plastic energy absorption. When the yield point is higher than expected, yielding will not occur, very little energy is absorbed in these members and greater energies are transferred through to the connection.

5.1. Ductility and materials

ASTM A36 steel has a minimum specified yield strength of 36 Ksi (250 MPa). The delivered steel will obviously have a higher value if it is within specification. Twenty years ago, the average yield strength for A36 was approximately 42 Ksi (290 MPa). In 1994, the average yield strength has increased to approximately 48 Ksi (330 MPa). This average value is 1/3 higher than the value assumed by many designs. Of course, while 48 Ksi (330 MPa) is the average value, some steel will have even higher values, with 55 Ksi (380 MPa) being fairly routine. For columns, the flange properties will typically be closer to minimum specified values, in part due to the thickness of the flanges. Furthermore, the tensile coupons for the mill test reports on rolled shapes are extracted from the web of the section.

The thinner webs routinely exhibit higher values than heav- ier flanges. According to mill test reports from steel in actual buildings, it is statistically possible to have floor beam material with higher actual yield strength properties than those of the column, even when materials are properly specified.

5.2. Ductility and notch effects Ductility of a material can only be exhibited in relatively smooth, notch-free configurations. In the presence of a notch, even a uniaxial tensile specimen will exhibit an increase in the apparent yield strength, and a greatly reduced elongation. As configured in the moment connection, geometric notches naturally occur as the horizontal beam flange meets the vertical column flange. The problem is exacerbated by the presence of a properly fused weld backing. It is further compounded by any regions of lack of fusion or slag inclusions in the weld and by poorly cut weld access holes. Under these conditions, ductile welds and ductile steel will not be able to exhibit ductility.

5.3. Ductility and triaxial stresses Most tensile and elongation data are obtained from uniaxial slowly loaded tensile specimens. When stretched, the elon- gated sample becomes thinner and more narrow. This reduction in the cross-section of the specimen can be measured after fracture and is generally expressed as reduction in area. It is not unusual for steel to exhibit 20-30% elongation under these test conditions.

When steel is simultaneously loaded in two or even three directions, however, it will not be able to exhibit its inherent ductility. Rather than behaving in a ductile manner, the steel will fracture without exhibiting any elongation and the fracture is typically termed 'brittle'. Yet, a uniaxial test specimen of the same material could exhibit tremen- dous elongation and ductility. Therefore, the designer must consider both the ductility of the material, and ductility of the configuration of the material. Ductile behavior is possible when:

(1) There is a shear stress (r) component resulting from the applied load.

(2) The shear stress component exceeds the critical shear stress of the material.

(3) The inelastic shear strain resulting from the shear stress is in a direction that will relieve the particular stress that is applied.

(4) There is a sufficient length of unrestrained material to permit a reduction in cross-sectional area (e.g. necking).

If these conditions are not met, ductile elongation cannot occur.

Restrained conditions reduce the shear stress component. Under ideally balanced triaxial stress conditions, there is no shear stress--hence, no ductile movement. With higher yield strengths (that is, higher shear strength), the movement will be reduced for a given force. A tensile test specimen with a short gage length will not deform as much as a specimen with a longer unrestrained length. For the SMRF, the connection is highly restrained, particularly when very large column sections are used. The behavior under these conditions can be examined mathematically. The strain (e) that results from applied stress (tr) can be found as follows:

Lessons learned f rom Nor thr idge earthquake: D. Ko Mi l le r 255

Figure 11

(0" 3 -- ~..~O" 2 -- j[..LO-I ) e3-- E

where /x = Poisson's raft() or 0.3 for steel. When written for all these principal axes and simultaneously solved for the principal stress, the following results:

= E[(1 - / x ) E3 + IJ,e2 +/zE,)]

(1 + /z ) (1 - 2/x)

This can be found for each principal axis. Figure 11 shows two regions in question. Point A is at

the weld joining the beam flange to the face of the column flange. Here there is restraint against strain (movement) across the width of the beam flange (E2). Point B is along the length of the beam flange away from the connecting weld. There is no restraint across the width of the flange or through its thickness.

Under highly restrained conditions (such as point A in Figure 11), el and IE 2 (strains in axes that are perpendicular to the direction of desired elongation) can be fixed at zero. The unit cube in Figure 12 represents point A. For a unit strain of 0.001 in the longitudinal direction, the resultant stresses are found to be

~ = 43.85 Ksi (302 MPa)

(r2 = 23.08 Ksi ( 159 MPa)

°c" I ............. ...... 2.7

Figure 13

o5 = 23.08 Ksi (159 MPa)

The ratio of these stresses will remain the same, so assuming a 70 Ksi (483 MPa) tensile strength steel, the stresses in the other two principal directions would be 36.84 Ksi (255 MPa). If the yield strength of the steel is 55 Ksi (379 MPa), the critical shear strength is one-half, or 27.5 Ksi (190 MPa). However, the greatest shear stress is as follows:

o-3 - o-2 T =

2

70 - 36.84

2

= 21.58 Ksi ( 149 MPa)

Since this is less than the critical shear stress (e.g. 27.5 Ksi), the critical tensile strength (70 Ksi) will be exceeded before elongation can occur. Brittle fracture will inevi- tably occur.

This can be easily visualized by considering Mohr 's circle of stress, illustrated in Figure 13. Stresses o-t and o'2 are plotted at 26.84 Ksi and (r3 at 70 Ksi. Note that the maximum shear stress is 20 Ksi, 7.5 Ksi less than the critical shear stress where yielding will occur.

In contrast, at a distance away from the column (see point B in Figure 11), the beam can be expected to behave differently. At a distance of 0.5 m from the column, for example, elongation in one direction can be facilitated by strains acting in the other two principal axes. Figure 14 represents a cube of material taken from point B. mathematically, the following may occur:

IJ 2

~a=+.O01

01

(32=0

i "''''''''''/-~:,.,'~II''" "i 03=30ksi

O~ =0

Figure 12 Figure 14

256 Lessons/earned from Northridge earthquake: D. K. Mil ler

• = 0.001in/in (mm/mm)

• = -0.0003

• = -0.0003

With these strains, at the ultimate tensile strength of the steel, the stresses are:

o-2 = 70 Ksi (483 MPa)

o-2 = o-J = 0 Ksi (OMPa)

The applied shear stress is 35 Ksi (241 MPa), exceeding the critical shear strength of 27.5 Ksi ( 190 MPa), resulting in ductility before fracture.

Mohr's circle of stress again helps to illustrate this state of stress. In Figure 15, o'1 and -o-2 are plotted as points set at zero. The third principle stress, o-3 is plotted at 70 Ksi. Notice that the maximum shear stress is 35 Ksi, 7.5 Ksi greater than the critical shear stress. Yielding, or ductile behavior, would be expected. In order to obtain ductility in a SMRF, it is critical that the location of the maximum stresses induced by the applied load occur in a region where the ductility of the steel can be facilitated; that is, in a region of uniaxial stress.

Two approaches have been successfully tested that illustrate the point. 'Stiffened' connections that utilize short coverplates, vertical ribs, or 'haunches' accomplish the desired behavior by (a) reducing the stress level in the connection and (b) pushing the expected hinge region into an area of uniaxial stress. The second approach deliberately reduces the beam capacity by eliminating some of the flange width for a small section. This reduced flange width concept (the so-called 'dog bone') clearly places the hinge formation region in an area where ductile behavior is possible. Test results to date (discussed below) corroborate this analysis.

6. Research activity

As the one-year anniversary of the Northridge earthquake approached, many who had been involved with the analysis of fractured connections, as well as the repair of damaged buildings, were acutely aware of the elapsed time. Some felt the degree of progress had been unfortunately slow and fewer answers were availaible than many had hoped, but as the anniversary approached, no one predicted that Janu- ary 17, 1995 would mark another tragic seismic event, this time on the opposite side of the ring-of-fire, in Kobe, Japan. That earthquake would prove to be far more damaging in

1:1.3=35

Oc,- 27:5 0 , ~ O~ 70

Figure 15

loss of life and cost of destruction. Despite the severe losses, Kobe provided yet another opportunity to better understand the behavior of structures. Analysis of the damage would reveal some similarities and some differences from the scene in Northridge. The differences in design, materials, and welding practices account for some of the disparities, but they also increased our understanding when similar behaviors obtained in spite of different practices. Perhaps one of the most beneficial outcomes of the Kobe earthquake, particularly with its coincidental date of occur- rence, is the degree of openness and cooperation that has been exhibited in the international engineering community since the event.

6.1. AISC research

One of the first organizations to rise to the challenge of Northridge was the American Institute for Steel Construc- tion. Under the leadership of the late Dr Jerry Haijer, a series of tests was quickly designed in order to increase the level of understanding regarding the connection and its performance. The primary investigator was Dr Michael Engelhardt, of the University of Texas, Austin. The results of these tests have been widely disseminated and much discussed in the open literature.

Some aspects of these tests have been criticized unjust- ifiably by other researchers and engineers. It must be understood that this was a quickly established test program, car- ried out under emergency-like conditions. Using the best available information, a series of test specimens was designed with the goal of obtaining a successful connection detail. That goal was achieved.

The test program employed relatively heavy columns (W14 x 455) and deep beams (W36 x 150). These member sizes were chosen since they duplicated a significant building under construction, and because these large member sizes had never been chosen before. The 'standard' detailing was never tested, as this was (at least at that time) assumed to offer little hope of success. From the program, valuable insight into several areas was made, including the following:

(1) Connection 'enhancements' (careful attention to weld quality, careful inspection of the completed joints, removal of weld backing from the bottom beam flange connection and removal of weld tabs) were insufficient to generate desired performance.

(2) The all welded connection (that is, welded beam web- to-column flange) performed slightly better than the connection with a bolted shear tab, but still fell well- short of the desired performance criteria.

(3) 'Reinforced' connection details that employed cover plates and vertical ribs performed significantly better, in general, than did the standard detail. These details move the region where the plastic hinge is expected to form away from the column face, into a region where the stress state is less complicated (e.g. uniaxial) and the steel is free to exhibit ductility.

(4) Even with the reinforced detail, adherence to written welding procedure specifications and code requirements is essential. When a variety of welding practices was used to fabricate the test specimens, analysis of weld metal extracted from the test specimens revealed significant differences in the notch toughness of the deposited weld metal, probably due to differences in heat input and interpass temperatures.

Lessons learned from Northridge earthquake: D. K. Mifler 257

The University of Texas tests showed conclusively that the problems were not limited to poor workmanship. In the summer of 1995, AISC sponsored a test of a concept commonly known as 'dog bones,' where the beam flanges have been shaved to provide for a reduced cross section a short distance away from the conaection. This approach attempts to create a 'weak link' away from the column connection in a region where the applied loads result in only uniaxial stresses. Limited testing has generated good results. The concept has the advantage of reducing the importance of maintaining tight control c,n base metal properties (yield strength), and reduces demand on the CJP groove welds. Most importantly, it places the greatest demand for ductility into a region where this behavior is capable of being delivered by the steel.

6.2. SAC activities Under funding by the Federal Emergency Management Agency (FEMA), a group known as SAC was formed to provide short-term underslumding of connection behavior. The SAC group was composed of three different groups, the Structural Engineers Association of California (SEAOC), Applied Technology Council (ATC) and Cali- fornia Universities For Research in Earthquake Engineering (CUREE). The SAC activities had two focuses: limited, preliminary research and technical advisories.

6.3. SAC research The primary thrust of the research activities was to test twelve assemblages that represented pre-Northridge practice. The test assemblages were sent to four universities for testing. All were tested following standard ATC-24 proto- col, and all failed prematurely, unable to deliver the degree of plastic rotation desired, and fracturing in a brittle manner rather than developing the desired plastic hinges. The poor performance of the connection surprised no one except for those who continued to focus on weld quality as the primary variable that would explain the Northridge behavior. It was a bit surprising to some, however, that not one specimen behaved as expected.

The type of fractures experienced duplicated those seen in the Northridge earthquake. This was extremely beneficial inasmuch as the predominant fractures that had been experienced in previous testing been predominantly near the weld joint. The column fractures had never been experienced in the laboratory.

A fortuitous seres of erogrs appears to have contributed to the column fractures. Due to a fabrication mixup, a test specimen was configured with beam material specifically produced to ASTM A572 grade 50 specificationns, this was welded to a column of the same grade of material. The actual values for the steel, however, were such that the beam material exceeded lx~th the yield and tensile of the column [beam: Fy = 62.6 Ksi (432 MPa), F, = 74.7 Ksi (5.5 MPa); column: Fy = 53.5 Ksi (369 MPa), F u = 72.5 Msi (500 MPa)]. This provided extremely valuable insight into the understanding of the unique column fractures.

After the specimens were tested and broke, each was repaired for further testing;. Two approaches were taken toward the repairs. The first simply attempted to restore the original intent of the connection, either replacing fractured metal with weld metal, or n~placing fractured materials and welding in new pieces of steel. The second approach involved deliberate modification of the connections, adding additional material to one or both sides of the connection.

The connections that received only weld repair were nonetheless different than their original configuration. Obviously, the materials had been strained and there was justifiable concern whether these members would be capable of delivering their expected performance. Secondly, the repaired assemblies would be geometrically different than their predecessors. The repair methodologies would rarely leave weld backing in place. Repair welding was performed using filler metals that were capable of delivering notch-tough weld metals. When steel was removed and replaced with other materials, the replacement piece was frequently thicker than the material removed.

The performance of the unmodified connections was similar to the original behavior. In some cases, the repair scheme had modified the structure and the performance was slightly enhanced over the original detail. In no case, however, was the degree of plastic rotation achieved acceptable. Still, these results are promising. It affirms the ability of repaired buildings to perform in their pre-earthquake mode. This is not to suggest that the repaired connections will never be damaged in subsequent earthquakes. Rather, since there was no collapses of steel-framed buildings associated with the magnitude 6.8 earthquake, the repaired building should be able to resist a similar earthquake in a similar manner. The natural enhancements that occur during repair welding may indeed generate a greater degree of confidence in their ability.

It has been suggested that the earthquake uniquely identified the weakest connections in buildings, perhaps those connections that had welds that were of lower quality than other connections. The demand on the remaining connections, according to this theory, was sufficiently low that the standard detail was adequate. If this is true, and if the repaired connections are equal to or superior to the original connections with high quality welds, these test results would indicate the building should perform better in subsequent earthquakes.

The modified connections, however, performed outstand- ingly well. Significantly increased levels of plastic rotation could be achieved, in several cases exceeding 5% radian rotation. These results imply that existing steels and their corresponding steel specifications, may be adequate providing the geometry of the connection is changed.

Several SAC and FEMA tests have been conducted at the University of California, San Diego, under Dr Chia- Ming Uang. The history of three specimens is illustrated in Figures 16-18. All three were originally constructed using pre-Northridge details and fabrication procedures. Good workmanship was achieved. All three were tested, and failed to achieve adequate ductility before failure (see Fig- ures 16(a), 17(a), 18(a).

Next, all three damaged specimens were repaired. During the original testing of specimen 1, the bottom beam flange connection broke. This particular specimen was going to be enhanced with the addition of a bottom side haunch. It was decided to test the repaired and modified detail without restoring the bottom flange connection. The weld was removed, the column face inspected for incipient cracks, and the haunch added (see Figure 16(b). Figure 16(c) demonstrates the improved behavior and the deformation of the beam indicative of hinge formation in Figure 16(d).

Specimen 2 was a 'repair-only' test. The attempt was to restore the connection to the previous condition without substantial changes. During the original test, some buckling of the top flange occurred, but Figure 17(a) demonstrates


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that the degree of rotation was not acceptable before fracture. Note, however, the better performance shown in Fig- ure 17(a) compared to Figure 16(a). To repair this specimen, a replacement flange piece was inserted (see Figure 17(b)). The performance of the previously tested specimen

was similar in the repaired-only condition (see Figure 17(c)). This was encouraging since it suggests the steel is not permanently damaged due to earthquake loading. The localized kinking resulted in tearing along the beam web- to-flange intersection (see Figure 17(d)).

Lessons/earned from Northridge earthquake: D. K. Miller 2 5 9

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Specimen 3 was repaired utilizing a combination of the previous two repair/modification approaches. The top flange buckled before the connection broke, so a flange splice was used. The bottom connection did not fracture, so unlike Specimen 1, this weld was still present. A haunch was applied, as in Specin~cn 1. Figure 18(b) shows the repair/modification configuration and Figure 18(c) shows the improved performance that was achieved. Hinge formation is apparent in Figure 18(d).

The SAC testing perfor~aed to date supports these conclusions:

(1) Good workmanship alone is not sufficient to produce acceptable results.

(2) Repaired (only) connections will perform like the original connections--no better, but also no worse.

(3) Modification of the joint can significantly improve connection performance, even on previously damaged connections.

These tests demonstrate that the pre-Northridge detail has some significant deficiencies. In addition, concepts that force plastic hinge formation into the unrestrained region of the beam enable inelastic deformations to occur, just as the analysis presented in section of this paper covering 'ductility' would predict.

6.4. SAC technical advisories The second major activity of the SAC organization was the preparation of Interim Guidelines. These will be made available through the sponsoring organization FEMA. This interim document provides the single best consolidation of technical information to date regarding repair procedures for existing buildings, as well as guidelines for new construction. Major topics include: damage classifications,

post-earthquake evaluation, post-earthquake inspection, post-earthquake repair and modification, new construction, metallurgy and welding, quality assurance, visual inspec- tions and non-destructive testing.

The Guidelines have affirmed the need for several connection detail changes, including the following:

(1) Weld backing should be removed from the bottom flange of beam-to-column connections. The resultant stress riser that is created at this 90 ° intersection, coupled with the uncertainty of inspecting the root of these welds and the high degree of difficulty associated with deposition of quality root passes in this region justify this requirement. In addition, it is recommended that a reinforcing fillet weld be applied to soften the transition at this 90 ° intersection. Removal of backing from other joints is not mandated as justification for these requirements is somewhat limited, and the activities associated with backing removal in other configurations may cause more harm than good.

(2) Weld tabs are required to be removed. In retrospect, an inappropriate amount of attention has probably been directed toward weld tabs and the use of the so-called end dams. The center of the beak-to-column flange joint, when the column is made of a I-shaped member, will be the most highly restrained portion of the connection. It is in this region that the most, if not all, of the fractures initiated. More will be said on this later. Nevertheless, the influence of weld tabs left in place, or the use of end dams, probably had little if any influence on the actual behavior of the connections in Northridge. Still, for highly strained members subject to seismic loading, this is a good recommendation that should be followed, particularly on an interim basis.

(3) For the beam-to-column flange welds, filler metal


should have specified minimum levels of notch toughness for deposit of weld metal. Current DI.1 code requirements do not mandate this requirement and existing AWS A5 Filler Metal Specifications do not require minimum specified levels of notch toughness for the commonly used E70T-4 electrode. To ensure modest levels of notch toughness, the Guidelines rec- ommend that filler metals be capable of delivering 20 ft pounds at 70F (27 J @ 20°C).

(4) Maximum welding interpass temperatures should be limited to 550F (287°C) when notch toughness is specified for the weld deposit. At the present time, the DI.1 code does not mandate an upper limit on the interpass temperature. When interpass temperatures are excessively high and specifically exceeding 600F (315°C), most filler metals will experience a deterio- ration in notch toughness, as well as a reduction in the yield strength. Properties of welds extracted from Northridge-damaged buildings have revealed evidence that interpass temperatures may have been exceedingly high. This, coupled with the high heat input levels associated with the sizes of beads observed in some fracture connections has prompted the guidelines to include a maximum interpass temperature of 550F (287°C) for prequalified WPSs.

Of equal importance to the additional provisions that have been listed in the Guidelines, it is worthwhile to exam- ine those issues which have been determined to be non- contributors to the Northridge damage. This is particularly important in light of several misconceptions which quickly surfaced as suspect contributors immediately after the earthquake. These issues include the following:

(1) The adequacy of the DI.1 code. While a few items have been identified as additional controls that need to be applied in the interim, the general overall suitability of the D 1.1 code has been reaffirmed.

(2) Welding processes. The Guidelines state that there is no evidence to suggest that one welding process is con- sistently superior to another, although the Guidelines also suggest that any process can be misused or misap- plied.

(3) Preheat. The Guidelines reaffirm the suitability of the AWS D1.1 Preheat Tables and suggests that additional preheat is not routinely required for new construction. The Guidelines go on to indicate that, under conditions of high restraint, additional preheat may be necessary.

(4) Peening. After the earthquake, a variety of specifications were written that required peening to be applied for routine fabrication, both new and repair. While the power of peening to reduce residual stresses is not dis- puted, neither should it be mandated for new construction or all repair. The Guidelines affirm this position.

(5) Post heat. Another post-Northridge reaction was to mandate post welds stress relief, or post heat operations. These should not be routinely required for most fabrication, either new or repair. Moreover, particularly when applied to repair, the application of unnecessary thermal activities may complicate the fabrication. The Guidelines prioritize the importance of control of hydrogen in the filler metals and proper application of preheat as more important than post heat, or, stress relief. While these operations may be useful, they

(6)

should be applied under the supervision of a knowl- edgeable welding engineer. Electrode diameter. The restriction on electrode diameter that was prevalent in many post-Northridge specifications has been found to be unwarranted, at least as a single, individual variable. While larger electrode diameters permit the use of higher deposition rates, a single variable of the electrode diameter cannot be iso- lated from the multitude of variables that influence weld quality. Many Northridge weld cross-sections revealed very large bead sizes, indicative of inappropriate welding procedures. High amperages, high wire feed speeds, high deposition rates, long electrical stick- outs, large weave passes and slow travel speeds all result in high heat input welding and large bead sizes. A restriction on electrode diameter alone would not directly address the problem of large bead sizes. Rather, the existing code restrictions on bead sizes have been reaffirmed as an adequate control when properly applied and enforced.

7. Conclusions

Many factors may have contributed to the damage seen in the Northridge earthquake. All issues must be addressed and controlled. Improved materials exist--we only need to see to it that they are specified. Better workmanship can be achieved--in general, better enforcement of existing specifications will be adequate to ensure improved quality. Qual- ity will also be enhanced by increased emphasis on in-process visual inspection. These issues can be addressed fairly easily. These measures alone, however, will not generate a connection that is able to behave as expected (that is, to exhibit ductility) until the basic details employed are changed. This will be achieved when the demand for ductility is concentrated in a region where the steel is capable of delivering ductile behavior.

References

1 American Welding Society, 'Structural Welding Code: ANSI/ANS D1.1-96,' Miami, Florida, 1996

2 American Welding Society, 'Structural welding committee position statement on Northridge earthquake welding issues,' Miami, Florida, November 10, 1995

3 Blodgen, Omer W. 'Design of Welded Structures', Cleveland, Ohio, The James F. Lincoln Arc Welding Foundation, 1996

4 Blodgett, Omer W. 'Details to Increase Ductility in SMRF Connec- tions," The Welding Innovation Quarterly, 1993, X(1 )

5 Blodgett, Omer W. 'The challenge of welding jumbo shapes, Part If: increasing ductility of connections,' The Welding Innovation Quar- terly, 1993, X(I)

6 Englebardt, M. D. and Sabol, T. A. 'Testing of welded steel moment connections in response to the Northridge earthquake,' Progr. Report to the AISC Advisory Subcommittee on Special Moment Resisting Frame Research, October, 1994, (2), 1996

7 Miller, Duane K. 'Ensuring weld quality in structural applications, part I: the roles of engineers, fabricators & inspectors,' The Welding Innovation Quarterly, 1996, XIII, No. 2

8 Miller, Duane K. 'Ensuring weld quality in structural applications, part II: effective visual inspection,' The Welding Innovation Quar- terly, 1996, XII |(3)

9 Miller, Duane K. 'Northridge: the role of welding clarified,' The Welding Innovation Quarterly, 1994, XI(2)

10 Miller, Duane K. 'Northridge: an update,' The Welding lnnowUion Quarterly, 1996, XIII(I)

11 Uang, C. M. and Latham, C. T. 'Cyclic testing of full-scale MNH- SMRF moment connections,' Structural Systems Research, Univer- sity of California, San Diego, March 1995

lessons learned from the northridge earthquake

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