recommended lrfd guidelines for the seismic design of highway bridges

1325 NCHRP 20-7(193) Task 6 Report.doc

NCHRP 20-07/Task 193

Task 6 Report for

Updating

“Recommended LRFD Guidelines for the Seismic Design of Highway Bridges”

Imbsen & Associates, Inc. - A TRC Company

1325 NCHRP 20-7(193) Task 6 Report.doc i

TABLE OF CONTENTS

Section No. Page No.

1 Review Reference Documents ............................................................ 1-1

2 Finalize Seismic Hazard Level............................................................. 2-1

2.1 Recommended Approach to Addressing Seismic Hazard ...................... 2-1

2.1.1 Background on Seismic Hazard................................................... 2-2

2.2 Proposed Seismic Hazard for Design of Normal Bridges ....................... 2-2

3 Expand the Extent of the “No Analysis” Zone.................................. 3-1

3.1 Introduction............................................................................................. 3-1

3.2 Proposed Range of Applicability of Analysis .......................................... 3-3

3.3.1 Column Shear Requirement for SPC B ..................................... 3-12

3.3.2 Column Shear Requirement for SPC C ..................................... 3-14

3.4 Drift Capacity for SPC B and SPC C .................................................... 3-15

3.5 Hinge Seat Requirement ...................................................................... 3-18

3.5.1 Minimum Edge Distance............................................................ 3-18

3.5.2 Other Movement ........................................................................ 3-19

3.5.3 Skew Effect................................................................................ 3-20

3.5.4 Relative Hinge Displacement..................................................... 3-21

1325 NCHRP 20-7(193) Task 6 Report.doc ii

4 Select the Most Appropriate Design Procedure for Steel ................. 4-1

4.1 General................................................................................................... 4-1

4.2 Design Examples.................................................................................... 4-2

4.3 Load Path and Performance Criteria ...................................................... 4-4

4.4 Summary ................................................................................................ 4-8

5 Recommend Liquefaction Design Procedure .................................... 5-1

5.1 Objective ................................................................................................ 5-1

5.2 NCHRP 12-49 Liquefaction Design Requirements................................. 5-1

5.3 Damage Severity in Past Earthquakes ................................................... 5-3

5.4 Proposed Liquefaction Design Requirements ........................................ 5-4

5.5 Summary ................................................................................................ 5-6

1325 NCHRP 20-7(193) Task 6 Report.doc iii

LIST OF FIGURES

Figure No. Page No.

Figure 2-1: Idealized Load – Deflection Curve of a Bridge................................... 2-6

Figure 2-2: Probabilistic to Deterministic Ratio at Selected Sites..................... 2-12

Figure 3-1: Elastic Response Spectra Curves (5% Damping) for Soil Profile Type D (M = 6.5 ± 0.25) (Caltrans SDC)................................... 3-6

Figure 3-2: Elastic Response Spectra Curve (5% Damping) for Soil Profile Type D (M = 8.0 ± 0.25) (Caltrans SDC)................................... 3-7

Figure 3-3: Core Flowchart ................................................................................... 3-11

Figure 3-4: Proposed Drift Capacity for SPC B and C ........................................ 3-18

Figure 3-5: Skew Effect Seat Width Amplification Factor for Various Skew Angles ................................................................................................. 3-20

Figure 3-6: Relative Seismic Displacement vs. Period Ratio ............................. 3-23

Figure 3-7: Proposed Seat Width Compared to NCHRP 12-49 and DIV 1A (H=20ft) ................................................................................................ 3-25

Figure 3-8: Proposed Seat Width Compared to NCHRP 12-49 and DIV 1A (H=30ft) ................................................................................................ 3-25

Figure 4-1: Seismic Load Path and Affected Components .................................. 4-6

1325 NCHRP 20-7(193) Task 6 Report.doc iv

LIST OF TABLES

Table No. Page No.

Table 2-1: Identified Sources of Conservatism ...................................................... 2-4

Table 2-2: Selected Sites for PSHA/DSHA Comparison ........................................ 2-8

Table 2-3: Design Spectral Acceleration based on NCHRP 1997 ......................... 2-9

Table 2-4: Probabilistic Spectral Acceleration for 10% and 5% in 50 Years..... 2-10

Table 2-5: Spectral Acceleration (Type B & D Soil) for 5% in 50 Years.............. 2-10

Table 2-6: One-Second Spectral Acceleration Comparison to USGS 1996 ...... 2-11

Table 2-7: Probabilistic to Deterministic Comparison of One-Second Acceleration........................................................................................ 2-12

Table 3-1: Proposed Partitions for Seismic Performance Categories A, B, C, and D............................................................................................................. 3-4

Table 3-2: 1.0 sec. Spectral Acceleration for Magnitude 6 .................................... 3-8



Table 3-5: 1.0 sec. Spectral Acceleration (Division 1A)......................................... 3-8

Table 3-6: Seismic Performance Category for Selected Sites .............................. 3-9

Table 3-7: Column Parameters .............................................................................. 3-17

1325 NCHRP 20-7(193) Task 6 Report.doc v

Table 4-1: Reduction Factors for Steel Superstructure Bracings......................... 4-8

Table 5-1: Damage Severity Description................................................................. 5-3

Table 5-2: Damage Severity Rating vs. Earthquake Magnitude............................ 5-4

1325 NCHRP 20-7(193) Task 6 Report.doc 1-1

TASK 1

1 REVIEW REFERENCE DOCUMENTS

A review of the pertinent documents and information that were available was

conducted and has been included in Tasks 2 thru 5 as needed. The reference

material that was selected for inclusion is attached as appendices for each of

the individual tasks. Their inclusion as appendices makes this Letter Report

somewhat self-contained and additionally, makes it more convenient for our

reviewers.

A separate section is included in this Letter Report for each of the tasks as

described below:

Section 2 presents the justification for the 1000-year return period (i.e., 5%

probability of exceedance in 50 years) as recommended for the seismic design of

highway bridges.

Section 3 includes a description of how the “no analysis” zone is expanded and

how this expansion is incorporated into the displacement based approach.

Section 4 describes the two alternative approaches available for the design of

highway bridges with steel superstructures and concludes with a

recommendation to use a force base approach for the proposed specification.

Section 5 describes the recommended procedure for liquefaction design to be

used for highway bridges. This aspect of the design is influenced by the

recommended hazard level and the no analysis zone covered in Tasks 2 and 3


respectively. The recommendations proposed are made taking into account the

outcome of these two tasks for Seismic Performance Category D.


TASK 2

2 FINALIZE SEISMIC HAZARD LEVEL

2.1 Recommended Approach to Addressing Seismic Hazard

The recommended approach to addressing the seismic hazard is based on the

following positions:

• Recommendations would be Primarily for Design against the Effects

Ground Shaking Hazard

• Selection of a Return Period for Design less than 2500 Years

• Inclusion of the USGS 2002 Update of the National Seismic Hazard

Maps

• Effects of Near Field and Fault Rupture to be addressed in a separate

later Task.

• Displacement Based Approach with both Design Spectral Acceleration

and corresponding Displacement Spectra provided

• Hazard Map under the control of AASHTO with each State having the

option to Modify or Update their own State Hazard using the most

recent Seismological Studies consistent with the Established Risk


2.1.1 Background on Seismic Hazard The current State of the Practice in addressing the seismic hazard for the

design of bridges in the U.S. has evolved from just conforming to AASHTO

Division 1-A requirements to adopting higher standards that take into account

the possible effects of larger earthquakes in the Eastern United States and the

impacts of major earthquakes that occurred recently in the Western United

States, Japan, Taiwan and Turkey. This change in the Seismic Hazard

Practice can be best illustrated in looking at the following sources:

• NEHRP 1997 Seismic Hazard Practice

• Caltrans Seismic Hazard Practice

• NYCDOT and NYSDOT Seismic Hazard Practice

• NCHRP 12-49 Seismic Hazard Practice

• SCDOT Seismic Hazard Practice

• Site Specific Hazard Analyses Conducted for Critical Bridges

Appendix 2A contains background on seismic hazard drawn from the above

mentioned sources.

2.2 Proposed Seismic Hazard for Design of Normal Bridges

In reviewing the seismic hazard practice in different regions as described

previously, it is apparent that some important aspects of this Practice need to

be taken into consideration when developing new Guidelines. These aspects

are pivotal in reaching the objective of producing Guidelines that are adoptable

by AASHTO.


These aspects include:

1) Consideration for lower return period for Design based on the Maximum

Considered Earthquake (MCE) maps published in 1996 with USGS 2002

Update shall be considered a minimum standard. Modification or

increase in the hazard intensity based on Seismological Studies needs to

be included as an option for states and agencies seeking a higher degree

of hazard identification to a specific region or bridge.

2) The reduction in the design intensity can be implicitly achieved by

considering applying a reduction/modifier factor for design spectrum

derived from USGS MCE maps. An alternative to this approach would

be embarking on developing new maps based on a modified new

definition of the MCE for Bridge Design.

3) Consideration of applying a reduction factor on the hazard intensity for

existing bridges or bridges located in rural areas.

4) Selection of a lower return period for Design is made such that Collapse

Prevention is not compromised when considering historical large

earthquakes. This reduction can be achieved by taking advantage of

sources of conservatism not explicitly taken into account in current

design procedures. These sources of conservatism are becoming obvious

based on recent findings from both observations of earthquake damage

and experimental data. Some of these sources are shown in Table 2-1.


Table 2-1: Identified Sources of Conservatism

Source of Conservatism Safety Factor

Computational vs. Experimental Displacement Capacity of Components

1.3

Effective Damping 1.2 to 1.5 Dynamic Effect (i.e., strain rate effect) 1.2 Pushover Techniques Governed by First Plastic Hinge to Reach Ultimate Capacity

1.2 to 1.5

Out of Phase Displacement at Hinge Seat Addressed in Task 3

1 The conservatism is directly coupled to the seismic reliability of the

structural system under consideration. The current state of the practice

favors continuous superstructures for the majority of bridges with an

objective of minimizing expansion joints to gain functionality, reduce

maintenance, and increase life cycle of the bridge. This selection has a

favorable impact on the earthquake redundancy of the bridge system.

Considering a single performance level of “No Collapse”, the seismic

redundancy of the bridge system is enhanced with the increase of the

number of plastic hinges that must yield and then fail in order to produce

the impending collapse of the structure. This enhanced redundancy

translates into a delayed failure (i.e. collapse) provided sufficient seat

width exists in the bridge system. Therefore two distinctly different

aspects of the design process need to be provided:

a) An appropriate method to design adequate seat width(s) considering out

of phase motions.

b) An appropriate method to design the ductile substructure components

without undue conservatism.


These two aspects are embedded with different levels of conservatism that

need to be calibrated against the single level of hazard considered in the design

process.

The first aspect is highly influenced by variation in the periods of the frames

on both sides of a joint as well as the damping generated by the ductile

behavior of plastic hinges. This aspect is addressed in terms of

recommendations or limits on periods ratio for frames on both sides of an

expansion joint.

The second aspect is addressed using a static push-over analysis. As shown in

Figure 2-4, the collapse displacement is usually reached when the P-Δ line

intersects the load-displacement curve of the structure, because at this point,

any increment in displacement produces an increment in the P-Δ effect due to

gravity loads that cannot be resisted by the lateral resistant system. It is

important to mention for structures with relatively small gravity loads, a much

larger reduction in component strength can be tolerated without reaching

structural collapse. This is especially relevant to bridge columns carrying axial

loads typically ranging from .05 c gf A′ to .15 c gf A′ maximum. In essence, the

continuity of the superstructure and low axial loads in columns make a typical

bridge more resilient against collapse in a seismic event.


Figure 2-1: Idealized Load – Deflection Curve of a Bridge

Under earthquake ground motions at the supports, the structure or any of its

components can fail under a smaller displacement than the displacement

Δcollapse illustrated in Figure 2-1. This failure is mainly attributed to

nonsymmetric cumulative plastic displacement that is highly depended on the

characteristics of the earthquake ground motions. The reliable displacement

capacity is typically associated with the displacement corresponding to a

limited decrease in strength of 20% to 30% maximum obtained under

monotonically increasing deformation. As shown in Figure 2-1, the

displacement capacity Δcapacity can only be established given the descending

slope following the point of maximum lateral resistance Fmax. Recognizing the


complexity of determining Δcapacity, the Δcapacity/bridge is used as a conservative and

simple measure assuming nominal properties.

In summary, the two aspects described above should be considered in the

practice to justify a reduction in the design hazard and ensure the development

of a simplified methodology that addresses the different sources of

conservatism included in the current state of the practice.

In order to assess the feasibility of a reduction in hazard from the 2% in 50

years hazard level adopted by NCHRP 12-49, a Probabilistic/Deterministic

comparison is conducted on 20 sites. Table 2-2 shows the state, city, dominant

source, latitude and longitude of the selected sites. Table 2-3 shows the short

period and one-second spectral acceleration for the Deterministic Seismic

Hazard Analysis (DSHA), the Deterministic Cap taken at 1.5 times the

(DSHA) value, the Maximum Considered Earthquake, and the Design Spectral

Acceleration SDS and SD1 based on NEHRP 1997 guidelines.

Table 2-4 shows the short period and one-second acceleration based on a

Probabilistic Seismic Hazard Analysis (PSHA) for 10% and 5% exceedance in

50 years. Table 2-4 includes two additional sites to the 20 sites identified in

Table 2-2 and 2-3.

Table 2-5 shows the short period and one-second period acceleration including

Type D soil effect for the proposed 5% exceedance in 50 years Design Spectrum.

Table 2-6 shows a comparison of the one-second acceleration (PSHA) to the

USGS 1996. As seen from Table 2-6, California sites show a decrease of the

acceleration values while other sites show a marginal change or an increase.

Table 2-7 shows the PSHA/DSHA comparison for the one-second acceleration of

the PSHA/DSHA ratio at each of the selected sites. These ratios are shown

graphically in Figure 2-2.


Based on this comparison, the following recommendations are proposed:

1. Adopt the 5% in 50 years hazard level for development of a design

spectrum.

2. Ensure sufficient conservatism (1.5 safety factor) for minimum seat

width requirement. This conservatism is needed to enable to use the

reserve capacity of hinging mechanism of the bridge system. This

conservatism shall be embedded in the specifications to address

unseating vulnerability. It is recommended to embed this safety factor

for sites outside of California.

3. Partition Seismic Performance Categories (SPCs) into four categories

and proceed with the development of analytical bounds using the 5% in

50 years Hazard level.

Table 2-2: Selected Sites for PSHA/DSHA Comparison ST CITY FEATURE DOMINANT SOURCE LATITUDE LONGITUDE CA Daly City Zip Code 94015 San Andreas 37.681240 -122.479000 CA San Francisco City Hall San Andreas 37.779083 -122.417450 CA SFOBB Site from Po/Roy Hayward 37.750000 -122.250000 CA Berkeley Site from Po/Roy Hayward 37.871667 -122.271667


CA Benicia Martinez

Site from Po/Roy Concord 38.000000 -122.116667

CA Los Angeles City Hall Puente Hills blind thrust

34.053700 -118.243183

CA Vincent Thomas Site from Po/Roy Location corrected

Palos Verdes 33.749218 -118.271466

CA Long Beach Zip Code 90810 Newport-Inglewood 33.813890 -118.217000 CA Coronado

Bridge Site from Po/Roy Rose Canyon 32.616667 -117.116667

WA Seattle Space Needle Seattle fault zone 47.621150 -122.348950 WA Tacoma North Site from Po/Roy Seattle fault zone 47.250000 -122.366667 UT Salt Lake City State Capital Wasatch fault, Salt

Lake City section 40.776367 -111.887983

UT Salt Lake City Site from Po/Roy Wasatch fault, Salt Lake City section

40.750000 -111.883333

IN Evansville Zip code 47720 New Madrid fault zone 38.023280 -87.617100 MO St. Louis Zip code 63129 New Madrid fault zone 38.466780 -90.319400 KY Paducah Zip code 42003 New Madrid fault zone 37.034190 -88.603800 TN Union City Zip code 38261 New Madrid fault zone 36.428110 -89.059500 TN Memphis City Hall New Madrid fault zone 35.148750 -90.054700 TN Memphis Zip code 38127 New Madrid fault zone 35.225170 -90.008400

Table 2-3: Design Spectral Acceleration based on NCHRP 1997


CITY Det Ss, g Det S1, g 1.5*Det Ss, g 1.5*Det S1, g MCE Ss, g MCE S1, g SDs, g SD1, gDaly City 1.49 0.85 2.23 1.28 2.23 1.28 1.49 0.85San Francisco 0.88 0.44 1.32 0.67 1.50 0.67 1.00 0.44SFOBB 0.84 0.30 1.26 0.45 1.50 0.60 1.00 0.40Berkeley 1.28 0.49 1.93 0.74 1.93 0.74 1.28 0.49Benicia Martinez 0.96 0.31 1.44 0.47 1.50 0.60 1.00 0.40Los Angeles 1.50 0.57 2.24 0.86 2.19 0.74 1.46 0.49Vincent Thomas 1.41 0.63 2.12 0.95 2.08 0.92 1.38 0.61Long Beach 1.28 0.51 1.91 0.77 1.81 0.70 1.20 0.47Coronado Bridge 1.19 0.47 1.78 0.70 1.37 0.54 0.91 0.36Seattle 1.34 0.48 2.01 0.73 1.41 0.48 0.94 0.32Tacoma North 0.47 0.18 0.71 0.28 1.20 0.41 0.80 0.27Salt Lake City 1.28 0.53 1.92 0.80 1.71 0.69 1.14 0.46Salt Lake City 1.25 0.53 1.88 0.79 1.70 0.69 1.13 0.46Evansville 0.27 0.09 0.41 0.13 0.67 0.19 0.45 0.13St. Louis 0.23 0.08 0.34 0.12 0.61 0.17 0.40 0.12Paducah 0.89 0.24 1.33 0.36 1.50 0.47 1.00 0.31Union City 0.86 0.23 1.29 0.35 1.50 0.57 1.00 0.38Memphis 0.60 0.17 0.91 0.25 1.40 0.38 0.93 0.25Memphis 0.65 0.18 0.98 0.27 1.50 0.41 1.00 0.27

Table 2-4: Probabilistic Spectral Acceleration for 10% and 5% in 50 Years

CITY10%/50 yr

Ss, g10%/50 yr

S1, g5%/50 yr

Ss, g5%/50 yr

S1, gDaly City 1.60 0.78 2.15 1.12San Francisco 1.15 0.53 1.45 0.69SFOBB 1.26 0.50 1.57 0.62Berkeley 1.65 0.63 2.19 0.83Benicia Martinez 1.24 0.43 1.58 0.55Los Angeles 1.20 0.41 1.60 0.54Vincent Thomas 1.02 0.37 1.47 0.56Long Beach 0.96 0.35 1.30 0.49Coronado Bridge 0.60 0.22 0.89 0.34Seattle 0.73 0.24 0.99 0.33Tacoma North 0.68 0.23 0.89 0.30Salt Lake City 0.69 0.24 1.10 0.42Salt Lake City 0.68 0.24 1.09 0.42Evansville 0.25 0.07 0.40 0.11St. Louis 0.23 0.06 0.36 0.10Paducah 0.54 0.11 0.97 0.24Union City 0.53 0.12 1.06 0.27Memphis 0.38 0.09 0.75 0.19Memphis 0.40 0.09 0.80 0.20Charleston 0.31 0.06 0.69 0.15Phoenix 0.09 0.03 0.12 0.04

Table 2-5: Spectral Acceleration (Type B & D Soil) for 5% in 50 Years


CITY

Type B5%/50 yr

Ss, g

Type B5%/50 yr

S1, g

Type D5%/50 yr

, g

Type D5%, 50 yr

,gDaly City 2.15 1.12 2.15 1.67San Francisco 1.45 0.69 1.45 1.04SFOBB 1.57 0.62 1.57 0.93Berkeley 2.19 0.83 2.19 1.24Benicia Martinez 1.58 0.55 1.58 0.83Los Angeles 1.60 0.54 1.60 0.81Vincent Thomas 1.47 0.56 1.47 0.84Long Beach 1.30 0.49 1.30 0.73Coronado Bridge 0.89 0.34 1.02 0.58Seattle 0.99 0.33 1.10 0.58Tacoma North 0.89 0.30 1.01 0.54Salt Lake City 1.10 0.42 1.17 0.67Salt Lake City 1.09 0.42 1.15 0.66Evansville 0.40 0.11 0.60 0.26St. Louis 0.36 0.10 0.55 0.24Paducah 0.97 0.24 1.08 0.46Union City 1.06 0.27 1.15 0.50Memphis 0.75 0.19 0.89 0.39Memphis 0.80 0.20 0.94 0.41Charleston 0.69 0.15 0.86 0.34Phoenix 0.12 0.04 0.19 0.09

DSS1DS

Table 2-6: One-Second Spectral Acceleration Comparison to USGS 1996

CITY10%/50 yr

S1, g5%/50 yr

S1, g

10%/50yr1996S1,g

5%/50yr1996S1,g Ratio 1 Ratio 2

Daly City 0.78 1.12 1.08 1.50 1.38 1.35San Francisco 0.53 0.69 0.64 0.83 1.22 1.20SFOBB 0.50 0.62 0.62 0.79 1.24 1.27Berkeley 0.63 0.83 0.65 0.86 1.03 1.04Benicia Martinez 0.43 0.55 0.55 0.69 1.27 1.25Los Angeles 0.41 0.54 0.42 0.54 1.02 1.00Vincent Thomas 0.37 0.56 0.41 0.58 1.10 1.03Long Beach 0.35 0.49 0.42 0.60 1.20 1.23Coronado Bridge 0.22 0.34 0.21 0.31 0.95 0.92Seattle 0.24 0.33 0.22 0.32 0.92 0.97Tacoma North 0.23 0.30 0.20 0.28 0.88 0.93Salt Lake City 0.24 0.42 0.21 0.42 0.86 1.00Salt Lake City 0.24 0.42 0.21 0.40 0.87 0.96Evansville 0.07 0.11 0.06 0.12 0.92 1.08St. Louis 0.06 0.10 0.05 0.10 0.85 0.99Paducah 0.11 0.24 0.09 0.20 0.81 0.83Union City 0.12 0.27 0.09 0.21 0.78 0.79Memphis 0.09 0.19 0.07 0.16 0.81 0.84Memphis 0.09 0.20 0.07 0.17 0.78 0.84Charleston 0.06 0.15 0.07 0.17 1.17 1.11Phoenix 0.03 0.04 0.03 0.04 1.11 1.02


Table 2-7: Probabilistic to Deterministic Comparison of One-Second Acceleration

CITYDet

S1, g5%/50 yr

S1, g RatioDaly City 0.85 1.12 1.31San Francisco 0.44 0.69 1.56SFOBB 0.30 0.62 2.07Berkeley 0.49 0.83 1.67Benicia Martinez 0.31 0.55 1.76Los Angeles 0.57 0.54 0.94Vincent Thomas 0.63 0.56 0.89Long Beach 0.51 0.49 0.95Coronado Bridge 0.47 0.34 0.72Seattle 0.48 0.33 0.68Tacoma North 0.18 0.30 1.63Salt Lake City 0.53 0.42 0.79Salt Lake City 0.53 0.42 0.79Evansville 0.09 0.11 1.24St. Louis 0.08 0.10 1.28Paducah 0.24 0.24 1.00Union City 0.23 0.27 1.13Memphis 0.17 0.19 1.13Memphis 0.18 0.20 1.12

0.00

0.50

1.00

1.50

2.00

2.50

Daly C

ity

San Fran

cisco

SFOBB

Berkele

y

Benici

a Mart

inez

Los A

ngele

s

Vincen

t Tho

mas

Long

Bea

ch

Corona

do B

ridge

Seattle

Tacom

a Nort

h

Salt La

ke C

ity

Salt La

ke C

ity

Evans

ville

St. Lou

is

Paduc

ah

Union C

ity

Memph

is

Memph

is

5%-50 Yr/Deterministic

Figure 2-2: Probabilistic to Deterministic Ratio at Selected Sites

1325 NCHRP 20-7(193) Task 6 Report.doc 2A-1

TASK 2

APPENDIX 2A


NEHRP 1997 Seismic Hazard Practice FEMA 274 describes the background of NEHRP 1997 provisions as follows:

“The NEHRP Recommended Provisions (BSSC, 1997) update process to the

1994 NEHRP Provisions included the formation of a special Seismic Design

Procedures Group (SDPG), consisting of earth scientists from the USGS and

engineers engaged in the update process. The SDPG was charged with the

responsibility of working with the USGS to produce ground motion maps

incorporating the latest earth science procedures, and with appropriate design

procedures to allow use of these maps in the Recommended Provisions. The

SDPG determined that rather than designing for a nationwide uniform hazard

- such as a 10%/50 year or 2%/50 year hazard- it made more sense to design for

a uniform margin of failure against a somewhat arbitrarily selected maximum

earthquake level”.

“This maximum earthquake level was termed a Maximum Considered

earthquake (MCE) in recognition of the fact that this was not the most severe

earthquake hazard level that could ever affect a site, but it was the most

severe level that it was deemed practical to consider for design purposes. The

SDPG decided to adopt a 2%/50 year exceedance level definition for the MCE in

most regions of the nation, as it was felt that this would capture recurrence of

all of the large-magnitude earthquakes that had occurred in historic times”.

“There was concern, however, that the levels of ground shaking derived for this

exceedance level were not appropriate in zones near major active faults. There

were several reasons for this. First, the predicted ground motions in these

regions were much larger than those that had commonly been recorded by near

field instrumentation in recent magnitude 6 or 7 California events. Second, it

was noted, based on the observed performance of buildings in these

earthquakes, that structures designed ot the code had substantial margin

against collapse for ground shaking that is much larger than that for which the


building had nominally been designed; in the judgment of the SDPG members,

this margin represented a factor of at least 1.5”.

“Consequently, it was decided to adopt a definition of the MCE in zones near

major active faults that consisted of the smaller of the probabilistically

estimated 2%/50 year motion or 150% of the mean ground motion calculated for

a deterministic characteristic earthquake on these major active faults, and to

design all buildings, regardless of location, to provide for protection of occupant

life safety at earthquake ground shaking levels that are 1/1.5 times (2/3) of the

MCE ground motion”.

Following the 1997 NEHRP Provisions, the ratio of the mapped acceleration at

one-second period for return periods of 474, 1000, 1500, 2000 and 2500 years is

normalized against the mapped acceleration at one-second period for a return

period of 474 years. The results of this normalization for California, Pacific,

Intermountain, Central US, and Eastern US are found in Table 2-1. The

California and Pacific Regions are designated with a deterministic cap based

on the description mentioned in the above paragraphs. The normalization is

appropriate for sites where the short period mapped acceleration SS is greater

than 1.5 g (i.e. higher ground shaking).

Table 2-1: Normalized One Second Spectral Acceleration Return Period

Years Region

California Pacific Intermountain Central US

Eastern US

474 1 1 1 1 1 1000 1.2 1.6 1.6 2.3 2.2 1500 1.4 2.2 2.0 3.5 3.4 2000 1.5 2.6 2.4 4.8 4.5 2500 1.6 3.0 2.7 6.1 5.7

Deterministic CAP

Yes Yes No No No

Achieving a national uniform hazard is difficult given the drastic difference

from one region to the other as illustrated in the normalization shown in Table


2-1. Furthermore, the regional difference in the recurrence of large magnitude

earthquakes makes the task of achieving a uniform hazard even more difficult.

Therefore, it is important that the selection of the Design Hazard can be

implicitly made such that collapse prevention is not compromised when

considering historical large earthquakes.

Caltrans Seismic Hazard Practice California Practice is described by Caltrans Commentary as follows:

“Caltrans bridge engineering practice has generally embraced deterministic

ground motion hazards since the 1971 San Fernando earthquake.

Deterministic practice considers the largest expected earthquake.

Deterministic practice considers the largest expected earthquake on any

known fault. Caltrans uses the mean event for standard practice, and refers to

it as the maximum credible earthquake (MCE). The deterministic method does

not take into consideration the recurrence of an MCE. This method assumes

that the MCE could occur at any time. Bridge engineering practice, therefore

should prudently allow for structures to be able to resist the MCE demands

without endangering the traveling public.”

“In recent years, Caltrans has implemented alternative ground motion hazard

site evaluations to address special situations. The alternatives were used with

consideration for: a) remaining life of a particular structure; b) bridge

performance capacity following an earthquake; c) liquefaction potential at an

existing bridge site; d) potential hazard at a bridge in anticipation of a future

retrofit or replacement; and other similar situations.”

Caltrans, with the support of an external Seismic Advisory Board and the

ATC-32 project team, has developed a set of seismic performance criteria for


new bridges. Following ATC-32, all bridges shall be designed to meet the

seismic performance criteria given in Table 2-2. Definitions of the terms in

Table 2-2 are given on the following page.

Table 2-2: Seismic Performance Criteria

Ground Motion

at Site

Ordinary Bridges Important Bridges

Functional-Evaluation

Ground Motion

Service Level-Immediate

Repairable Damage


Minimal Damage

Safety-Evaluation

Ground Motion

Service Level-Limited

Significant Damage


Repairable Damage

Each bridge shall be classified as either Important or Ordinary, as follows:

a) Important Bridge: Any bridge satisfying one or more of the following:

– Required to provide secondary life safety

– Time for restoration of functionality after closure would create a

major economic impact

– Formally designated by a local emergency plan as critical

b) Ordinary Bridge: Any bridge not classified as an Important Bridge.

The Evaluation Levels are defined as follows:

a) Safety-Evaluation Ground Motion: This ground motion may be assessed

either deterministically or probabilistically. The deterministic

assessment corresponds to the maximum credible earthquake (MCE), as

defined by the Division of Mines and Geology Open File Report 92-1


(CDMG, 1992). A probabilistically assessed ground motion is one with a

long return period (approximately 1000-2000 years).

For Important Bridges both methods shall be given consideration; however, the

probabilistic evaluation shall be reviewed by a Caltrans-approved consensus

group. For Ordinary Bridges, the motions shall be based only on the

deterministic evaluation.

b) Functional-Evaluation Ground Motion: This is a probabilistically

assessed ground motion that has a 60 percent probability of not being

exceeded during the useful life of the bridge. The determination of this

event is to be reviewed by a Caltrans-approved consensus group.

The following performance levels, expressed in terms of service levels and

damage levels are defined as follows:

a) Service Levels

– Immediate: Full access to normal traffic is available almost

immediately following the earthquake.

– Limited: Limited access (i.e., reduced lanes, light emergency

traffic) is possible within days of the earthquake. Full service is

restorable within months.

b) Damage Levels

– Minimal Damage: Essentially elastic performance.

– Repairable Damage: Damage that can be repaired with a

minimum risk of losing functionality.

– Significant Damage: A minimum risk of collapse, but damage

that would require closure to repair.


Following recommendations of ATC-32, Caltrans published the Seismic Design

Criteria Version 1.1, July 1999 (SDC 1.1) with focus on Ordinary Bridges as

previously defined in ATC-32. The design spectra (i.e. ARS curves) included in

SDC 1.1 were adopted from the ATC-32 design spectra. However, recognizing

some of the complexities dealing with the roles of probabilistic and

deterministic assessments, it was found that depending on the seismic activity

of a given region, the deterministic and probabilistic assessments may be

different. In the highly seismic zones of the San Francisco Bay region, the

deterministic ground motion assessments using the mean ARS spectra for the

MCE correspond to return periods of about 300 to 400 years. This variation

between the probabilistic and deterministic approaches is still an outstanding

issue in the California Seismic Hazard Practice.

Based on the currently adopted SDC 1.2 released in December 2001, an

ordinary bridge is designed for a standard 5% damped SDC ARS curve, a

modified SDC ARS curve, or a site-specific ARS curve.

For preliminary design, prior to receiving the geotechnical engineer’s

recommendation, a standard SDC ARS curve may be used in conjunction with

the peak rock acceleration from the 1996 Caltrans Seismic Hazard Map. The

standard SDC ARS curves were adopted from ATC-32. If standard SDC ARS

curves are used during preliminary design, they should be adjusted for long

period bridges and bridges in close proximity to a fault as described below.

For preliminary design of structures within 10 miles (15 km) of an active fault,

the modified SDC ARS curve is obtained by magnifying the spectral

acceleration on the SDC ARS curves as follows:

• Spectral acceleration magnification is not required for T ≤ 0.5 seconds

• Increase the spectral accelerations for T ≥ 1.0 seconds by 20%


• Spectral accelerations for 0.5 ≤ T ≤ 1.0 shall be determined by linear

interpolation

For preliminary design of structures with a fundamental period of vibration

T ≥ 1.5 seconds on deep soil sites (depth of alluvium ≥ 250 feet (75 m)) the

modified SDC ARS curve is obtained by magnifying the spectral ordinates of

the standard ARS curve as follows:

• Spectral acceleration magnification is not required for T ≤ 0.5 seconds

• Increase the spectral accelerations for T ≥ 1.5 seconds by 20%

• Spectral accelerations for 0.5 ≤ T ≤ 1.5 shall be determined by linear

interpolation

A site specific response spectrum is typically required when a bridge is located

in the vicinity of a major fault or located on soft or liquefiable soils and the

estimated earthquake moment magnitude Mm > 6.5.

In formally adopting the displacement approach following the release of

Caltrans Seismic Design Criteria Version 1.1, July 1999, the State of Practice

is implicitly calibrated to the Mean Hazard as stated above including a 1.2

magnification factor of spectral acceleration ordinates for a period of one

second or greater for bridges near a fault. It is deemed important to mention

that the normalization shown in Table 1 reveals that California’s one-second

spectral ratio is at 1.6 and that the NEHRP 1997 deterministic cap is set at 1.5

of the mean ground motion as previously mentioned under the NEHRP 1997

Seismic Hazard Practice. When considering locations close to active faults, the

normalization of the one-second spectral ratio shows average value is close to

1.8 but can be as high as 2.2 using the USGS 1997 maps. This value

represents an increase of 12% to 38% in comparison to the 1.6 ratio. This

increase is comparable to the 20% increase that the Caltrans Seismic Design


SDC 1.2 has adopted in its December 2001 release. Recent studies suggest

considerably larger increases that have, however, not been yet endorsed in the

practice

NYCDOT and NYSDOT Seismic Hazard Practice New York State Department of Transportation (NYSDOT) practice continues

in accordance with 1996 AASHTO Standard Specifications for Highway

Bridges, Division 1-A. New York City Department of Transportation

(NYCDOT) has adopted modifications to the 1996 AASHTO Division 1-A that

reflect the findings of the “New York City Seismic Hazard and It’s Engineering

Application” final report prepared by Weidlinger Associates, December 1998.

These modifications are applicable to NYC Metro Region including the

Counties listed in Table 2-3. NYCDOT bridges are classified as Critical,

Essential and Other. Table 2-3 summarizes the relationship of bridge

importance and performance requirements. In all cases, “No Collapse” is

permitted.

The following guidelines are adopted for NYCDOT bridges:

• For Bridges designed by the one level approach (Essential and Other),

Figure 2-1 shows the acceleration response spectra to be used for

different soil types (soil classes). Soil classes are defined in Table 2-4.

• Site specific soil effects for the two earthquake levels approach (i.e.

Critical Bridges) should be obtained from an expert. Soil spectra for

different types of soils, base on NEHRP amplification factors are not

recommended for design of Critical Bridges.

• For Critical bridges, site-specific ground motions shall be computed

using rock motions based on the spectra for hard rock (Soil Class A).


• For Essential bridges where the site condition can be classified as A, B,

C, D, or E, the empirically derived soil spectra shown on Figure 2-1 (2/3

(2% probability of exceedance in 50 years)) shall be used.

• For bridges classified as Other, including single-span bridges, the

spectra shown on Figure 2-1 shall be used for Soil Classes A, B, C, D,

and E.

• For Soil Class F, regardless of bridge Importance Category, site-specific

analysis should be performed.


Table 2-3: Performance Criteria and Seismic Hazard Level for Design and Evaluation of Bridges

(Applicable to NYC Metro Region/Downstate Counties: Bronx, Kings, Nassau, New York, Queens, Richmond, Rockland and Westchester)


Table 2-4: Soil Classification

Figure 2-1: NYCDOT Soil Acceleration Response Spectra for One-

Level Approach (2% in 50 Years Probability of Exceedance) /1.5


A comparison between the panel spectra adopted by NYCDOT and NEHRP

1997 shows the following:

• The NEHRP/97 MCE hard rock is lower than the Panel Hard Rock as

demonstrated in Figure 2-2.

• The short period soil amplification Panel Spectra factors are lower than

the NEHRP 1997 corresponding factors as demonstrated in

Table 2-5.

• The long period soil amplification Panel Spectra factors are essentially

the same in comparison to the NEHRP 1997 corresponding factors as

demonstrated in Table 2-6.

• In Summary, the long period spectral acceleration values for

NEHRP/1997 are lower than the NYCDOT adopted Panel Spectral

values. However, the difference is less pronounced when comparing the

NEHRP 1997 values to the 2/3 NYCDOT values adopted for essential

and other bridges as mentioned earlier in this section.


Figure 2-2: Spectra Comparison – NYC Rock Acceleration Response Spectra – 5% Damping-2500 Year Return Period = 2% in 50 Years

Probability of Exceedance

Table 2-5: Fa = Short Periods Amplification Factor/Normalized For Soil Class B For 2% In 50 Yrs Probability of Exceedance Curves (*)

For NEHRP 94 For NEHRP 97 Panel Based Spectra Soil Class Aa = .16g Ss = .40g Aa = .30g Ss = .72g - .75g

A 0.8 0.8

B 1.0 1.0

C 1.2 1.1

D 1.48 1.2

E 2.02 1.2

(*) VALUES FOR Aa AND Ss ARE FOR SOIL CLASS B


Table 2-6: Fv = Long Periods Amplification Factor/Normalized For Soil Class B For 2% In 50 Yrs Probability of Exceedance Curves (*)

For NEHRP 94 For NEHRP 97 Panel Based Spectra Soil Class Av = .09g S1 = .09g Av = .13g S1 = .13g

A 0.8 0.8

B 1.0 1.0

C 1.7 1.67

D 2.4 2.28

E 3.5 3.41

(*) VALUES FOR Av AND S1 ARE FOR SOIL CLASS B

NCHRP 12-49 Seismic Hazard Proposed Practice The proposed November 2001 NCHRP 12-49 Design Earthquakes and

Performance Objectives are best described in Section 1.3 of the Recommended

LRFD Guidelines for the Seismic Design of Highway Bridges Part I

Specifications. “The USGS probabilistic maps published in 1996 (Frankel et

al., 1996) are used in formulating the design Earthquakes Response Spectrum.

The provisions provide ‘definitive performance objectives and damage states’ for

two design earthquakes with explicit design checks to ensure the performance

objectives are met. The upper-level event, termed the rare earthquake or

Maximum Considered Earthquake (MCE), describes ground motions that, for

most locations, are defined probabilistically and have a probability of

exceedance of 3% in 75 years. However, for locations close to highly active

faults, the MCE ground motions are deterministically bounded so that the

levels of ground motions do not become unreasonably high. Deterministic

bound ground motions are calculated assuming the occurrence of maximum

magnitude earthquakes on the highly active faults and are equal to 1.5 times

median ground motions for the maximum magnitude earthquake but not less

than 1.5 g for the short-period spectral acceleration plateau and 0.6g for 1.0-


second spectra acceleration. On the current MCE maps, deterministic bounds

are applied in high-seismicity portions of California, in local areas along the

California-Nevada border, along coastal Oregon and Washington, and in high-

seismicity portions of Alaska and Hawaii. In areas where deterministic bounds

are imposed, ground motions are lower than ground motions for 3% PE in 75

years. The MCE earthquake governs the limits on the inelastic deformation in

the substructures and the design displacements for the support of the

superstructure.

The lower level design event, termed the Expected Earthquake, has ground

motions corresponding to 50% PE in 75 years. This event ensures that

essentially elastic response is achieved in the substructures for the more

frequent or ‘expected earthquake’.”

According to the proposed Guidelines, “Bridges shall be designed to satisfy the

performance criteria given in Table 2-7. As a minimum, bridge shall be

designed for the life safety level of performance”.


Table 2-7: Design Earthquakes and Seismic Performance Objectives

Performance Level (1)

Probability of Exceedance

For Design Earthquake Ground Motions (4)

Life Safety

Operational

Service (2) Significant Disruption

Immediate Rare Earthquake (MCE)

3% PE in 75 years/1.5 Median Deterministic

Damage (3) Significant Minimal

Service Immediate Immediate Expected Earthquake

50% PE in 75 years Damage Minimal Minimal to None

Notes:

(1) Performance Levels:

These are defined in terms of their anticipated performance objectives in

the upper level earthquake. Life safety in the MCE event means that the

bridge should not collapse but partial or complete replacement may be

required. Since a dual level design is required the Life Safety

performance level will have immediate service and minimal damage for

the expected design earthquake. For the operational performance level

the intent is that there will be immediate service and minimal damage for

both the rare and expected earthquakes.

(2) Service Levels:

• Immediate – Full access to normal traffic shall be available following

an inspection of the bridge.

• Significant Disruption – Limited access (Reduced lanes, light

emergency traffic) may be possible after shoring, however the bridge

may need to be replaced.


(3) Damage Levels:

• None – Evidence of movement may be present but no notable

damage.

• Minimal – Some visible signs of damage. Minor inelastic response

may occur, but post-earthquake damage is limited to narrow flexural

cracking in concrete and the onset of yielding in steel. Permanent

deformations are not apparent, and any repairs could be made under

non-emergency conditions with the exception of superstructure joints.

• Significant – Although there is no collapse, permanent offsets may

occur and damage consisting of cracking, reinforcement yield, and

major spalling of concrete and extensive yielding and local buckling

of steel columns, global and local buckling of steel braces, and

cracking in the bridge deck slab at shear studs on the seismic load

path is possible. These conditions may require closure to repair the

damage. Partial or complete replacement of columns may be

required in some cases. For sites with lateral flow due to

liquefaction, significant inelastic deformation is permitted in the

piles, whereas for all other sites the foundations are capacity-

protected and no damage is anticipated. Partial or complete

replacement of the columns and piles may be necessary if significant

lateral flow occurs. If replacement of columns or other components is

to be avoided, the design approaches producing minimal or moderate

damage, such as seismic isolation or the control and reparability

design concept should be assessed.

(4) The upper-level earthquake considered in these provisions is designated

the Maximum Considered Earthquake, or MCE. In general, the ground

motions on national MCE ground motion maps have a probability of


exceedance (PE) of approximately 3% PE in 75 years. However, adjacent

to highly active faults, ground motions on MCE maps are bounded

deterministically as described above. When bounded deterministically,

MCE ground motions are lower than ground motions having 3% PE in

75 years. The performance objective for the expected earthquake is

either explicitly included as an essentially elastic design for the 50% PE

in 75 year force level or results implicitly from design for the 3% PE in

75 year force level.

The 2001 Guidelines were amended in May 2002 to delete the “Operational”

Performance Objective. The provisions were edited to reflect the consideration

of only the Life Safety Performance Objective. This change was necessary to

address the concern of some stakeholders that having more than one

performance objective as a minimum standard may create undue liability to

stakeholders that choose only a Life Safety Performance Objective with no

explicit consideration for the Operational Performance Objective. The main

changes of interest to the above-mentioned table are shown in

Table 2-8 in the Word Edit format.

Table 2-8: Design Earthquakes and Seismic Performance Objectives

Performance Level Objective(1)

Probability of Exceedance (PE)


For Design Earthquake Ground Motions (4) Life Safety Operational

Service (2) Significant Disruption

Immediate Rare Maximum Considered Earthquake (MCE)

3% PE in 75 years/ or 1.5 Median Deterministic Damage (3) Significant Minimal

Service Immediate Immediate Expected Earthquake (EE)

50% PE in 75 years Damage Minimal Minimal to None

Notes:

(1) Performance Levels Objective:

These are defined in terms of their anticipated performance objectives in

the upper level earthquake. Life safety in the MCE event means that the

bridge should not collapse but partial or complete replacement may be

required. Since a dual level design is required the Life Safety

performance level will have immediate service and minimal damage for

the expected design earthquake. For the operational performance level

the intent is that there will be immediate service and minimal damage for

both the rare and expected earthquakes.

(2) Service Levels:

• Immediate – Full access to normal traffic shall be available following

an inspection of the bridge.

• Significant Disruption – Limited access (Reduced lanes, light

emergency traffic) may be possible after shoring, however the bridge

may need to be replaced.

(3) Damage Levels:

• None – Evidence of movement may be present but no notable

damage.


• Minimal – Some visible signs of damage. Minor inelastic response

may occur, but post-earthquake damage is limited to narrow flexural

cracking in concrete and the onset of yielding in steel. Permanent

deformations are not apparent, and any repairs could be made under

non-emergency conditions with the exception of superstructure joints.

• Significant – Although there is no collapse, permanent offsets may

occur and damage consisting of cracking, reinforcement yield, and

major spalling of concrete and extensive yielding and local buckling

of steel columns, global and local buckling of steel braces, and

cracking in the bridge deck slab at shear studs on the seismic load

path is possible. These conditions may require closure to repair the

damage. Partial or complete replacement of columns may be

required in some cases. For sites with lateral flow due to

liquefaction, significant inelastic deformation is permitted in the

piles, whereas for all other sites the foundations are capacity-

protected and no damage is anticipated. Partial or complete

replacement of the columns and piles may be necessary if significant

lateral flow occurs. If replacement of columns or other components is

to be avoided, the design approaches producing minimal or moderate

damage, such as seismic isolation or the control and repairability

design concept should be assessed.

(4) The upper-level earthquake considered in these provisions is designated

the Maximum Considered Earthquake, or MCE. In general, the ground

motions on national MCE ground motion maps have a probability of

exceedance (PE) of approximately 3% PE in 75 years. However, adjacent

to highly active faults, ground motions on MCE maps are bounded

deterministically as described above. When bounded deterministically,

MCE ground motions are lower than ground motions having 3% PE in


75 years. The performance objective for the expected earthquake is

either explicitly included as an essentially elastic design for the 50% PE

in 75 year force level or results implicitly from design for the 3% PE in

75 year force level.

SCDOT Seismic Hazard Practice The South Carolina Department of Transportation (SCDOT) has initiated the

development and implementation of a bridge seismic design program. A

central feature of the new SCDOT bridge design program is the development of

new seismic bridge design criteria and standards that: 1) incorporate a new

generation U.S. Geological Survey seismic ground shaking hazard maps, 2)

treat certain inadequacies of existing bridge design codes to adequately

address the large earthquake, and 3) address the no collapse bridge criteria

and life safety issues in the central and eastern United States. This section

summarizes the upgraded bridge seismic design provisions and describes

variations in national seismicity that motivated the development of the

SCDOT “Seismic Design Specifications for Highway Bridges”. Basically, the

revised specifications specify that the design of new bridges in South Carolina

directly account for the effects of the large earthquake as done by the State of

California. This is to ensure conformance with the guiding principle used in

the development of AASHTO provisions that the "…exposure to shaking from

the large earthquake should not cause collapse of all or part of the bridge…”

Several of the revisions were adopted from bridge design provisions of the

California Department of Transportation (Caltrans) [2], because of similar high

intensity seismic hazard at the Safety Evaluation Earthquake (SEE) level and

the state-of-practice progress gained due to recent earthquakes that have not

yet been incorporated into AASHTO.


At least two developments of the U.S. Geological Survey during the past

several years have been a major contribution to bridge earthquake

engineering. One development was an assessment of the nature of the seismic

ground shaking hazard as it varies nationally that revealed apparent

inequalities in safety that result when a single level of probability common to

bridge code design is used. The second development was a new generation of

probabilistic ground-motion hazard maps that provide uniform hazard spectra

for exposure times of 500 and 2500 years and make possible the treatment of

the inequality in safety of bridge code design using existing earthquake

engineering design and evaluation provisions and methodology.

The new generation of probabilistic ground motion maps was produced by the

USGS under the National Earthquake Hazard Reduction Program (NEHRP)

with significant input from the committee on Seismic Hazard Maps of the

Building Seismic Safety Council (BSSC) and the Structural Engineers

Association of California (SEAOC). They allow development of uniform hazard

spectra and permit direct definition of the design spectra by mapping the

response spectral ordinates at different periods.

The recommended seismic design procedures were developed to meet current

bridge code objectives, including both serviceability and life safety in the event

of a large earthquake. The primary function of these new provisions is to

provide minimum standards for use in bridge design to maintain public safety

in the extreme earthquake likely to occur within the state of South Carolina.

They are intended to safeguard against major failures and loss of life, to

minimize damage, maintain functions, or to provide for easy repair.

For normal or essential bridges (see Table 2-9), the Single Level Design

Method is adopted by this code. This method consists of applying seismic


design loading calculated based upon the value of the spectral accelerations of

the 2%/50-year earthquake (i.e., the Safety Evaluation Earthquake).

Table 2-9: Seismic Performance Criteria

Ground Motion Level

Performance Level

Normal Bridges

Essential Bridges

Critical Bridges

Service NR* NR* Immediate Functional-Evaluation Damage NR* NR* Minimal

Service Impaired Recoverable Maintained Safety-Evaluation Damage Significant Repairable Repairable

*Functional Evaluation Not Required.

For Critical Bridges, which are designated by SCDOT, the seismic performance

goals are to be achieved by a two-level design approach, one for each of the two

earthquakes (i.e., Two-Level Design Method). In addition to the 2%/50-year

earthquake (Safety Evaluation Earthquake), critical bridges shall also be

designed to provide adequate functionality after the 10%/50-year earthquake

(Functional Evaluation Earthquake). The minimum performance levels for the

design and evaluation of bridges shall be in accordance with the level of service

and damage for the two design earthquakes as shown in Table 2-9. Service

Levels and Damage Levels are defined in these criteria. The Bridge Category

is also defined in these criteria. The SCDOT may specify project-specific or

structure-specific performance requirements different from those defined in the

table. For example, for a Critical or Essential bridge it may be desirable to

have serviceability following a 2%/50-year earthquake. The SCDOT may


require a site-specific design spectrum or a complete hazard study as part of

the design.

This new SCDOT specification establishes design and construction provisions

for bridges in South Carolina to minimize their susceptibility to damage from

earthquakes. This specification is intended to be used in conjunction with

AASHTO Division I [3] and as a replacement to Division I-A, Seismic Design,

of the same specifications. Additionally, the new specifications include

references to the AASHTO Guide Specifications for Seismic Isolation

Design [4].

The principles used for the development of the new SCDOT provisions are:

i. Small to moderate earthquakes should be resisted within the

essentially elastic range of the structural components without

significant damage. The Functional Evaluation Earthquake (FEE) is

adopted to represent seismic ground motion level produced by small

to moderate earthquakes.

ii. State-of-Practice seismic ground motion intensities and forces are

used in the design procedures.

iii. Exposure to shaking from large earthquakes should not cause

collapse of all or part of the bridge. Where possible, damage that

does occur should be readily detectible and accessible for inspection

and repair unless prohibited by the structural configuration. The

Safety Evaluation Earthquake (SEE) is adopted to represent seismic

ground motion level produced by large earthquakes.

The performance levels, expressed in terms of service levels and damage levels,

are:


(a) Service Levels

• Immediate: Full access to normal traffic is available

immediately following the earthquake.

• Maintained: Short period of closure to Public. Immediately Open

to Emergency Vehicles.

• Recoverable: Limited period of closure to Public.

• Impaired: Extended closure to Public.

(b) Damage Levels

• Minimal Damage: No collapse, essentially elastic performance.

• Repairable Damage: No collapse. Concrete cracking, spalling of

concrete cover, and minor yielding of structural steel will occur.

However, the extent of damage should be sufficiently limited that

the structure can be restored essentially to its pre-earthquake

condition without replacement of reinforcement or replacement of

structural members (i.e., ductility demands less than 4). Damage

can be repaired with a minimum risk of losing functionality.

• Significant Damage: Although there is minimum risk of collapse,

permanent offsets may occur in elements other than foundations.

Damage consisting of concrete cracking, reinforcement yielding,

major spalling of concrete, and deformations in minor bridge

components may require closure to repair. Partial or complete

demolition and replacement may be required in some cases.

Bridge structures on the state highway system are classified as “normal

bridges”, “essential bridges” or “critical bridges”. For a bridge to be classified


as an “essential bridge” or a “critical bridge”, one or more of the following items

must be present: (1) bridge is required to provide secondary life safety, (2) time

for restoration of functionality after closure creates a major economic impact,

and (3) the bridge is formally designated as critical by a local emergency plan.

Each bridge is classified as Critical, Essential or Normal as follows:

(a) Critical bridges: Bridges that must be open to all traffic once

inspected after the safety evaluation design earthquake and be

usable by emergency vehicles and for security/defense purposes

immediately after the safety evaluation design earthquake, i.e., a

2,500-year return period event.

(b) Essential bridges: Bridges that will, as a minimum, be open to

emergency vehicles and for security/defense purposes after the safety

evaluation design earthquake, i.e., a 2,500-year return period event

and open to all traffic within days after the SEE event.

(c) Normal Bridges: Any bridge not classified as a Critical or Essential

Bridge.

The SCDOT Specifications aims at including state-of-practice in seismic based

on displacement analysis for reinforced concrete components. Force reduction

factors are used for steel superstructure due to limited use of the displacement

approach in steel design of bridges. In addition, only limited level of ductility

is so far accepted for members of steel superstructure with a plate girder

system.

In September 2003, SCDOT adopted new Seismic Hazard Maps for Bridges.

These new SCDOT Seismic Hazard Maps take into account the sediment

thickness and/or the near surface weathering, updating the State seismic


hazard information that was originally provided in the 2001 Specifications

based on USGS.


TASK 3

3 EXPAND THE EXTENT OF THE “NO ANALYSIS” ZONE

3.1 Introduction In developing the Displacement Based Approach, reference to the analysis can

be separated into two types:

a) Analysis conducted to establish seismic displacement demands on the

structures. This reference is similar to the reference made by AASHTO

Division 1-A for required seismic analysis in regions where PGA > 9% g

to determine forces. This can be referred to as “Seismic Demand

Analysis”.

b) Analysis conducted to establish the displacement capacity of the

structure, a subsystem or a component of the structure. This can be

referred to as “Seismic Capacity Analysis”. This type of analysis is also

commonly referred to as Pushover Analysis. In addition to obtaining the

displacement capacity of the structure, the “Seismic Capacity Analysis”

is used to obtain the load path and force distribution on the members of

the structure based on the hinging mechanism of these members. These

forces are used to design various members such that the developed

hinging mechanism of the overall system is confirmed.

In summary, the “Seismic Capacity Analysis” includes two parts. One is the

“Displacement Capacity” and the second is the “Capacity Design”.


The overall second objective identified in Task F3-4 is to increase the range of

applicability for No Analysis or Limited Analysis. This objective is made in

reference to NCHRP 12-49 Proposed Guidelines where it was found that

considerable amount of analysis was required on a larger number of bridges in

comparison with the AASHTO Division 1-A Practice. This finding is well

documented in the evaluation conducted on NCHRP 12-49 by performing trial

designed in several states. In further examining this objective, several steps

that are required to fulfill this objective are identified:

1) At a minimum, maintain the number of bridges under the “Seismic

Demand Analysis”. This objective is accomplished by comparing

Proposed Guidelines to current requirements in AASHTO Division 1-A.

2) Relative to Proposed NCHRP 12-49 Guidelines, reduce the number of

bridges where “Seismic Capacity Analysis” needs to be performed. This

objective is accomplished by identifying a threshold where implicit

procedures can be used.

3) Identify threshold where “Capacity Design” shall be used. This objective

is achieved in conjunction with the “Seismic Capacity Analysis”

requirements.

In reviewing the current State of the Practice in addressing the Range of

Applicability for No Analysis or Limited Analysis, the following sources are

examined:

1. AASHTO Division 1-A

2. Caltrans Seismic Design Criteria

3. NCHRP 12-49

4. SCDOT Specifications


The review of these four references is documented in Task F3-5 Report

AASHTO T-3 Support and included in Appendix 3A as background

information.

3.2 Proposed Range of Applicability of Analysis In addressing the proposed range of Applicability of Analysis, a key issue is the

selection of the most pertinent parameter indicative of the seismic demands

considered in the design of the bridge structure. The Spectral Acceleration at

1.0 second period, 1a DSS − , for the Design Spectrum is adopted considering the

following:

• 1a DSS − is a good representation of the difference in regional demands

(i.e., 1a DSS − is considerably lower in the Eastern U.S.)

• The choice of high frequency spectral indicator as recommended in

NCHRP 12-49 penalizes the Eastern U.S. for no credible justification

considering that damage to bridges is associated with low frequency

range of bridge period.

• The choice of 1a DSS − fits well with the adopted displacement approach for

bridges considering that ductility is taken into account when assessing

the capacity.

Considering the first objective of the recommended specifications addressing

the selection of a Return Period and Design Spectrum for a Single Hazard

Level pertaining to a No Collapse Criteria of bridges, the Important

Classification (IC) as defined in AASHTO Division 1-A is reduced to one

classification. Furthermore, considering the second objective of the

recommended specifications for defining and increasing the range of


applicability for No Analysis or Limited Analysis, the Seismic Performance

Category definition is changed to include four categories SPC A, B, C, and D

encompassing requirements for:

• Seismic Demand Analysis requirement.

• Seismic Capacity Analysis requirement.

• Capacity Design requirement.

• Level of seismic detailing requirement including four tiers corresponding

to SPC A, B, C and D.

The above-mentioned approach is an extension to the direction taken in

NCHRP 12-49 and SCDOT Specifications. The Seismic Performance

Categories SPC A, B, C and D ranges are partitioned based on the one-second

spectral acceleration 1a DSS − similarly to the SCDOT Specifications except that

the four requirements mentioned above are developed further to achieve the

second objective of the recommended specifications (i.e., reduce number of

bridges requiring analysis).

Table 3-1 shows the partition of the proposed Seismic Performance Categories

A, B, C and D.

Table 3-1: Proposed Partitions for Seismic Performance Categories A, B, C, and D

Value of 1a DSS − Importance Classification (IC)


1a DSS − < 0.15g A

0.15g ≤ 1a DSS − < 0.30g B 0.30g ≤ 1a DSS − < 0.50g C 0.50g ≤ 1a DSS − D

For illustration, the one-second acceleration corresponding to 0.5g, proposed to

be the threshold for SPC D, is pinpointed in Figures 3-1 and 3-2, showing the

acceleration response spectrum for Magnitude 6.5 and 8 (Type D Soil),

respectively. As seen from the figures, SPC D would include sites with Peak

Bedrock Acceleration greater than 0.3g for magnitude 6.5 and greater than

0.2g for magnitude 8. This shows that SPC D is rather conservative in

applying the most stringent criteria on the sites mentioned above.

Tables 3-2 thru 3-4 show the one-second spectral acceleration for Caltrans SDC

Magnitude 6.5, 7 and 8 for Soil Type B thru E. The numbers that are not

shaded represent values greater than the 0.5g threshold considered for SPC D.

Table 3-5 shows the one-second acceleration for (Division 1A) design spectrum

for Soil Type 1 thru 4. The numbers that are not shaded represent values

greater than 0.5g considered the threshold for SPC D.

Table 3-6 shows the one-second acceleration modified for Type B and D soil for

the sites selected in Task 2. Each site is assigned an SPC based on the

proposed partition shown in Table 3-1. Table 3-6 reflects the distribution of

SPC A, B, C and D given Type B and D soils.


Figure 3-1: Elastic Response Spectra Curves (5% Damping) for Soil Profile Type D (M = 6.5 ± 0.25) (Caltrans SDC)


Figure 3-2: Elastic Response Spectra Curve (5% Damping) for Soil Profile Type D (M = 8.0 ± 0.25) (Caltrans SDC)


Table 3-2: 1.0 sec. Spectral Acceleration for Magnitude 6 "A"(g) soil B soil C soil D soil E

0.1 0.08 0.14 0.19 0.280.2 0.16 0.21 0.32 0.520.3 0.24 0.37 0.45 0.690.4 0.35 0.48 0.55 0.800.5 0.41 0.53 0.610.6 0.48 0.61 0.70


0.1 0.10 0.17 0.24 0.360.2 0.21 0.32 0.41 0.690.3 0.29 0.44 0.54 0.810.4 0.42 0.59 0.66 0.990.5 0.52 0.68 0.770.6 0.67 0.86 1.000.7 0.92 1.19 1.38


0.1 0.12 0.21 0.29 0.430.2 0.24 0.38 0.48 0.750.3 0.35 0.53 0.64 0.930.4 0.45 0.63 0.71 1.020.5 0.57 0.73 0.900.6 0.72 0.93 1.120.7 0.96 1.25 1.45

Table 3-5: 1.0 sec. Spectral Acceleration (Division 1A)

"A"(g) soil 1 soil 2 soil 3 soil 40.1 0.12 0.14 0.18 0.240.2 0.24 0.29 0.36 0.480.3 0.36 0.43 0.54 0.720.4 0.48 0.58 0.72 0.960.5 0.60 0.72 0.90 1.20.6 0.72 0.86 1.08 1.440.7 0.84 1.01 1.26 1.68


Table 3-6: Seismic Performance Category for Selected Sites

State CITY

Type B Soil5%/50 yr

Sa1, g SPC

Type D Soil5%/50 yr

Sa1, g SPCCA Daly City 1.12 D 1.67 DCA San Francisco 0.69 D 1.04 DCA SFOBB 0.62 D 0.93 DCA Berkeley 0.83 D 1.24 DCA Benicia Martinez 0.55 D 0.83 DCA Los Angeles 0.54 D 0.81 DCA Vincent Thomas 0.56 D 0.84 DCA Long Beach 0.49 C 0.73 DCA Coronado Bridge 0.34 C 0.58 DWA Seattle 0.33 C 0.58 DWA Tacoma North 0.30 C 0.54 DUT Salt Lake City 0.42 C 0.67 DCA Salt Lake City 0.42 C 0.66 DIN Evansville 0.11 A 0.26 B

MO St. Louis 0.10 A 0.24 BKY Paducah 0.24 B 0.46 CTN Union City 0.27 B 0.50 DTN Memphis 0.19 B 0.39 CTN Memphis 0.20 B 0.41 CSC Charleston 0.15 B 0.34 CAZ Phoenix 0.04 A 0.09 A

The three requirements for each of the proposed Seismic Performance

Categories are as follows:

1. SPC A

a. No Displacement Capacity Check Needed

b. No Capacity Design Required

c. Tier I No Detailing Requirements


2. SPC B

a. Implicit Displacement Capacity Check Required (i.e., use a Closed

Form Solution Formula)

b. No Capacity Design Required

c. Tier II Level of Detailing

3. SPC C

a. Implicit Displacement Capacity Check Required

b. Capacity Design Required

c. Tier III Level of Detailing

4. SPC D

a. Pushover Analysis Required

b. Capacity Design Required

c. Tier IV Level of Detailing

The level of detailing for Tiers I, II, III, and IV will consider at a minimum the

following:

• Column Longitudinal Reinforcement Splicing

• Column Transverse Reinforcement Splicing

• Column Plastic Hinge Zone Identification

• Joint Shear Reinforcement


• Knee Joint Reinforcement

• Bent Cap Continuous Reinforcement

• Superstructure Continuous Reinforcement

• Footing Shear Reinforcement

• Bent Cap Shear Reinforcement

• Plate Girder Bracing/Diaphragm Detailing

The three requirements for each of SPC A, B, C and D will follow the core

flowchart that was presented in Task F3-5 and shown in Figure 3-3.

Figure 3-3: Core Flowchart

SPC "A"

D em an dA n aly sisSPC "B " Im plic it

C apacity

Y es

No

Y es1D

C ≤ T ier IID etailin g

C om pleteYes

SPC "C"

No No

Y es1D

C ≤ C apacityD esig n

T ier IIID etailin g

Yes

SPC "D"

No No

P u sh overC apacityA n alys is

1DC ≤ T ier IV

D etailin g

Yes

No

A d ju st B rid g eC h aracteristicsDep en d s on Ad ju stm en ts

Yes

D em an dA n aly sis

Im plic itC apacity C om plete

C om plete

D em an dA n aly sis

C apacityD esig n C om plete


The major performance measures that govern the design for each of the

Seismic Performance Categories include the following:

1. Column Shear requirement

2. Drift Capacity requirement

3. Seat Width requirement

3.3.1 Column Shear Requirement for SPC B The shear demand for a column, Vd, in SPC B shall be determined based on the

lesser of:

• The force obtained from an elastic linear analysis

• The force corresponding to plastic hinging of the column

The column shear strength capacity shall be calculated based on

n oV Vφ ≥ 0.85φ = (3.1)

n c sV V V= + (3.2)

,v yhs

A f DV

s⎛ ⎞

= ⎜ ⎟⎝ ⎠

where 2v bA n Aπ⎛ ⎞= ∗ ∗⎜ ⎟⎝ ⎠ (3.3)

n = number of individual interlocking spiral or hoop core sections.

For tied columns or pier walls (in the weak direction).


v yhs

A f DV

s⎛ ⎞

= ⎜ ⎟⎝ ⎠

(3.4)

vA = Total area of the shear reinforcement

c c eV v A= × (3.5)

0.8e gA A= × (3.6)

1 3.52000c c c

g

Pv f fA

α⎛ ⎞

′ ′ ′= + ≤⎜ ⎟⎜ ⎟⎝ ⎠

(3.7)

Spirally reinforced columns 0.015 s ytfα ρ′ = (3.8)

Rectangular hoop reinforced columns 0.030 w ytfα ρ′ = (3.9)

Where the spiral reinforcement ratio,

4 sps

ADs

ρ = (3.10)

and the web reinforcement ratio

vw

Abs

ρ = (3.11)


3.3.2 Column Shear Requirement for SPC C The shear demand for a column, Vd, in SPC B shall be determined based on the

force corresponding to plastic hinging of the column including an overstrength

factor

The column shear strength capacity shall be calculated based on

n oV Vφ ≥ 0.85φ = (3.12)

n c sV V V= + (3.13)

,v yhs

A f DV

s⎛ ⎞

= ⎜ ⎟⎝ ⎠

where 2v bA n Aπ⎛ ⎞= ∗ ∗⎜ ⎟⎝ ⎠

(3.14)

n = number of individual interlocking spiral or hoop core sections.

For tied columns or pier walls (in the weak direction).

v yhs

A f DV

s⎛ ⎞

= ⎜ ⎟⎝ ⎠

(3.15)

vA = Total area of the shear reinforcement

c c eV v A= × (3.16)

0.8e gA A= × (3.17)


1 3.52000c c c

g

Pv f fA

α⎛ ⎞

′ ′ ′= + ≤⎜ ⎟⎜ ⎟⎝ ⎠

(3.18)

Spirally reinforced columns 0.010 s ytfα ρ′ = (3.19)

Rectangular hoop reinforced columns 0.020 w ytfα ρ′ = (3.20)

Where the spiral reinforcement ratio,

4 sps

ADs

ρ = (3.21)

and the web reinforcement ratio

vw

Abs

ρ = (3.22)

3.4 Drift Capacity for SPC B and SPC C Columns for bridges in SPC B are targeted for a limited drift corresponding to

minor damage. Columns for bridges in SPC C are targeted for a maximum

drift corresponding to moderate damage. The approach taken to come up with

a closed form solution is to equally weigh in the results of numerical methods

as well as experimental testing of various columns.

Considering the numerical approach as described below, columns with

diameter ranging from 3 feet to 7 feet having 1 to 4% longitudinal

reinforcement and height ranging between 20 to 50 feet are considered. The


different permutations are shown in Table 3-7. A regression analysis is

performed and a lower bound curve is identified in Figure 3-4 for the following:

a. Curve 1, designated as C1, showing drift capacity corresponding to

column yielding

b. Curve 2, designated as C2, showing drift capacity corresponding to

concrete spalling

c. Curve 3, designated as C3, showing drift capacity corresponding to a

column ductility of 4.

The drift capacity for all three curves are shown as a function of the

slenderness ratio FbL where:

F = Flixity Factor ranging from 1 to 2

b = column width or diameter

L = column clear height

Experimental results are considered based on the statistical study adopted by

Pacific Earthquake Engineering Research Center (PEER) and reported by

Berry and Eberhand August 2003. The recommended equation by Berry and

Eberhard at the onset of spalling is:

11.6 1 110( )g c

PDH A f H

⎛ ⎞⎛ ⎞Δ ⎜ ⎟= − +⎜ ⎟⎜ ⎟′ ⎜ ⎟⎝ ⎠⎝ ⎠ (3.23)

Considering a typical bridge column axial load corresponding to 0.1 g cA f ′ , Curve

4, designated as C4, shows graphically the recommended equation by Berry

and Eberhand (PEER).


The recommended maximum drift capacity for SPC B is further defined as

Curve 5, designated as C5, and shown in Figure 3-4.

Curve 2 + Curve 4Curve 52

= (3.24)

The recommended maximum drift capacity for SPC C is further defined as

Curve 6, designated as C6, and shown in Figure 3-4.

Curve 3 + Curve 4Curve 62

= (3.25)

Table 3-7: Column Parameters

Column Diameter D (ft) ρ

(%) Column Height L (ft)

3 1,2,3,4 20 4 1,2,3,4 20,30 5 1,2,3,4 20,30,40 6 1,2,3,4 30,40,50 7 1,2,3,4 30,40,50


0.00

1.00

2.00

3.00

4.00

5.00

6.00

0.1 0.15 0.2 0.25 0.3

Fb/L

Drif

t Cap

acity

(%) Yield (C1)

Spalling (C2)Ductility 4 (C3)Experimental (C4)SPC B (C5)SPC C (C6)

Figure 3-4: Proposed Drift Capacity for SPC B and C

3.5 Hinge Seat Requirement The calculation for a hinge seat width involves four components:

a. Minimum edge distance

b. Other movement attributed to prestress shortening, creep, shrinkage,

and thermal expansion or contraction

c. Skew effect

d. Relative hinge displacement

3.5.1 Minimum Edge Distance The minimum edge distance set by Division IA and NCHRP 12-49 is set at 4

inches. It is recommended to retain this value.

SPC C

SPC B


3.5.2 Other Movement Division IA currently has a movement rating of 2 inches per 100 feet for SPC B

and a movement rating of 3 inches per 100 feet for SPC C & D.

The seat width based on NCHRP 12-49 is calculated as:

21(1 1.25 )0.10 0.0017 0.007 0.05 1 2 vF SBN L H H

L cos α

⎡ ⎤ +⎛ ⎞⎢ ⎥= + + + × + ⎜ ⎟⎢ ⎥⎝ ⎠⎣ ⎦ (3.26)

L = the distance between joints in meters

H = the tallest pier between the joints in meters

B = the width of the superstructure in meters

α = the skew angle

The term .0017L equates to 100.0017 1003.3m

ft⎛ ⎞⎜ ⎟⎝ ⎠

or .05 100m

ft equal to 2 inches

per 100 feet.

Three alternatives are considered for including “other” movement in the seat

width equation:

a. The first alternative considers the temperature movement tΔ and other

movements as calculated for various states based on their extreme

temperature range, in addition to prestress and shortening, and thermal

expansion or contraction.

b. The second alternative is a 2-inch movement per 100 feet, which is quite

conservative.


c. The last alternative has a 1-inch movement rating per 100 feet

considered an average nominal value in Practice especially in

combination with seismic movement.

It is recommended for clarity and transparency to adopt Alternative (a) stated

above.

3.5.3 Skew Effect A comparison of Equation 6.3.1 adopted in NCHRP 12-49 to Division I-A seat

width magnification for various skew angles is shown in Figure 3-5. As seen

from Figure 3-5, NCHRP 12-49 magnification is larger than Division I-A.

Doubling the magnification set in Division 1A as shown in the upper bound

curve as ( )21 4000

S+ is recommended. This recommendation is based on the

failures observed in past earthquakes for bridges with skewed bents.

1

1.05

1.1

1.15

1.2

1.25

1.3

1.35

1.4

1.45

0 5 10 15 20 25 30 35 40 45

Skew Angle

Am

plifi

catio

n Fa

ctor NCHRP 12-49

Division 1AProposed

Figure 3-5: Skew Effect Seat Width Amplification

Factor for Various Skew Angles


3.5.4 Relative Hinge Displacement The relative hinge displacement, Deq, is determined following guidelines by

Desroches & Fenves adopted by the recently published FHWA “Seismic

Retrofitting Manual for Highway Structures, 2004.”

( )2 2min max 12 min max2eqD D D D Dρ= + − (3.27)

where, Dmin = Displacement of the short period frame

Dmax = Displacement of the long period frame.

The correlation coefficient 12ρ is calculated as:

( )( )( ) ( ) ( )

3/ 22

12 2 22 2

8 1

1 4 1

ε β βρ

β ε β β

+=

− + + (3.28)

where 2

1

TT

β = T2 and T1 being the first and second modes of the structure

system.

The damping ε is calculated as:

( )15% 1 0.95 0.05ε μ μπ

= + − − (3.29)

where μ is the ductility factor

Consider the displacement ratio α :

min

max

DD

α = (3.30)


( )2 2 2 2max max 12 max2eqD D D Dα ρ α= + − (3.31)

( )2121 2α αρ= + − (3.32)

In the long period range, α , is also equal to the ratio of the short period frame

over the long period frame.

short

long

TT

α = (3.33)

Figure 3-6 shows Dmax vs. the ratio α for the following:

a. Deq for a target ductility of 2 shown as Curve 1

b. Deq for a target ductility of 4 shown as Curve 2

c. Caltrans SDC shown as Curve 3

d. Relative hinge displacement based on (Trocholak is et. al. 1997) shown

as Curve 4

Considering that a variation from the design plans of the structure cannot be

avoided during the life of the structure and that a substantial drop in the

required seat width is only achieved for an α greater than 0.8, it is

recommended that:

• Deq is equal to 1.1 Dmax the peak value of Curves 1 and 2.

Furthermore, a safety factor of 1.5 is proposed for regions other than California

as described in Task 2.

• Deq is equal to 1.1 Dmax for California


• Deq is equal to 1.65 Dmax for states other than California

0.00

0.20

0.40

0.60

0.80

1.00

1.20

1.40

1.60

0 0.2 0.4 0.6 0.8 1 1.2

Ratio of Tshort/Tlong

Rat

io o

f Deq

/Dm

ax

Curve 1Curve 2Curve 3Curve 4

Figure 3-6: Relative Seismic Displacement vs. Period Ratio

The proposed seat width requirement is compared to NCHRP 12-49 Equation

3.26 (shown in Curves 1 and 2 of Figure 3-7) and Division 1A seat width

requirement (shown in Curves 3 and 4 of Figure 3-7). The following is

considered for Equation 3.26:

a) substitution of SDR 2 “FvS1” by the maximum value of 0.25.

b) substitution of “B/L” ratio by the maximum value of 3/8.

c) Substitution of SDR 3 “FvS1” by the maximum value of 0.40.

The proposed seat width requirement is shown in Figures 3-7 and 3-8 for

H = 20 ft and 30 ft, respectively independent of any skew effect. The proposed

seat width requirement is illustrated with four cases identified in four curves:


1. Curve 5 for FvS1 = 0.15g corresponding to a period of 1 second for the

flexible frame with Deq equal to 1.65 Dmax.







The calculation for seat width requirement of the four cases above considers a

1-inch displacement per 100 feet for displacement other than seismic. The

choice of one inch per 100 feet leads to shallower slope of lines 5 thru 8 and

reinforces the choice of a realistic TΔ rather than the conservative 2 inches per

100 feet adopted in NCHRP 12-49. It is expected that the choice of alternative

(a) identified in Section 3.5.2 would yield a movement not exceeding one inch

per 100 feet of bridge length. The choice of a realistic eqΔ is important for the

design of hinges within-a span and the selection of a reasonable dimension for

the bent cap width.


0

5

10

15

20

25

30

35

40

45

0 200 400 600 800 1000

Bridge Length (ft.)

Seat

Wid

th (i

n.)

Curve 1 SDR 2Curve 2 SDR 3Curve 3 SPC BCurve 4 SPC C&DCurve 5 .15g, 1 secCurve 6 .5g, 1 secCurve 7 .15g, 2 secCurve 8 .50g, 2 sec

Figure 3-7: Proposed Seat Width Compared to NCHRP 12-49

and DIV 1A (H=20ft)

05

1015

2025

3035

40

45

50

0 200 400 600 800 1000Bridge Length (ft.)

Seat

Wid

th (i

n.)

Curve 1 SDR 2Curve 2 SDR 3Curve 3 SPC BCurve 4 SPC C&DCurve 5 .15g, 1 secCurve 6 .5g, 1 secCurve 7 .15g, 2 secCurve 8 .50g, 2 sec

Figure 3-8: Proposed Seat Width Compared to NCHRP 12-49

and DIV 1A (H=30ft)


REFERENCE

Berry, Michael and Eberhand, Marc, Pacific Earthquake Engineering Research

Center (PEER), “Estimating Flexural Damage in Reinforced Concrete

Columns,” University of California, Berkeley, August 2003.

DesRoches, Reginald, and Gregory Fenves, New design and analysis

procedures for intermediate hinges in multiple-frame bridges. Berkeley, Calif.:

Earthquake Engineering Research Center, University of California. 1997.

202p. (UCB/EERC 97/12).

Aschheim, Mark, and Jack P. Moehle, Shear Strength and Deformability of RC

Bridge Columns Subjected to Inelastic Cyclic Displacements, Berkeley, Calif.:

Earthquake Engineering Research Center, University of California. March

1992. (UCB/EERC 92/04).


TASK 3

APPENDIX 3A


AASHTO Division 1-A Range of Applicability of Analysis The analysis requirements based on AASHTO Division 1-A are derived based

on the Seismic Performance Category (SPC) and the regularity or irregularity

of a given bridge. These requirements are relevant to the “Seismic Demand

Analysis” mentioned above.

Each bridge is assigned to one of four Seismic Performance Categories (SPC), A

through D, based on the Acceleration Coefficient (A) and the Importance

Classification (IC), as shown in Table 3-1. Minimum analysis and design

requirements are governed by the SPC.

Table 3-1: Seismic Performance Category (SPC)

Acceleration Coefficient

Importance Classification (IC)

A I II A ≤ 0.09 A A 0.09 < A ≤ 0.19 B B 019 < A ≤ 0.29 C C 0.29 < A D C

An Importance Classification (IC) is assigned for all bridges with an

Acceleration Coefficient greater than 0.09 for the purpose of determining the

Seismic Performance Category (SPC) as follows:

1. Essential bridges – IC = I

2. Other Bridges – IC = II

Bridges shall be classified on the basis of Social/Survival and Security/Defense

requirements, guidelines for which are given in the Commentary of AASHTO

Division 1-A.


Minimum requirements for the selection of an analysis method for a particular

bridge type are given in Table 3-2. Applicability is determined by the

“regularity” of a bridge which is a function of the number of spans and the

distribution of weight and stiffness. Regular bridges have less than seven

spans, no abrupt or unusual changes in weight, stiffness, or geometry and no

large changes in these parameters from span-to-span or support-to-support

(abutments excluded). They are defined in Table 3-3. Any bridge not

satisfying the requirements of Table 3-3 is considered to be “not regular”. A

more rigorous, generally accepted analysis procedure may be used in lieu of the

recommended minimum such as the Time History Method (Procedure 4).

Table 3-2: Minimum Analysis Requirements

Seismic Performance

Category

Regular Bridges with

2 Through 6 Spans

Not Regular Bridges with

2 or More Spans A Not Required Not Required

Use Procedure Use Procedure B, C, D 1 or 2 3

Procedure 1. Uniform Load Method

Procedure 2. Single-Mode Spectral Method

Procedure 3. Multimode Spectral Method

Procedure 4. Time History Method


Table 3-3: Regular Bridge Requirements Parameters

Value Number of Spans 2

3 4 5 6

Maximum subtended angle (curved bridge)

90º 90º 90º 90º 90º

Maximum span length ratio from span-to-span

3 2 1.5 1.5

Maximum bent/pier stiffness ratio from span-to-span (excluding abutments)

- 4 4 3 2

Note: All ratios expressed in terms of the smaller value.

Curved bridges comprised of multiple simple spans shall be considered to be

“not regular” bridges if the subtended angle in plan is greater than 20º; such

bridges shall be analyzed by either Procedure 3 or 4.

Caltrans Range of Applicability of Analysis The Caltrans Seismic Design Criteria (SDC) specify the minimum seismic

design requirements that are necessary to meet the performance goals

established for Ordinary Standard bridges.

A structure must meet all of the following requirements to be classified as an

Ordinary Standard bridge:

• Span lengths less than 300 feet (90 m).

• Constructed with normal weight concrete girder, and column or pier

elements.


• Horizontal members either rigidly connected, pin connected, or

supported on conventional bearings by the substructure, isolation

bearings and dampers are considered nonstandard components.

• Dropped bent caps or integral bent caps terminating inside the exterior

girder, C-bents, outrigger bents, and offset columns are nonstandard

components.

• Foundations supported on spread footing, pile cap with piles, or pile

shafts.

• Soil that is not susceptible to liquefaction, lateral spreading, or scour.

Ordinary Nonstandard bridges require project specific criteria to address their

non-standard features.

Based on Caltrans SDC, each bridge presents a unique set of design

challenges. The designer is given the latitude to determine the appropriate

methods and level of refinement necessary to design and analyze each bridge

on a case-by-case basis. Situations may arise that warrant detailed attention

beyond what is provided in the SDC. The designer is referred to other

resources to establish the correct course of action. The Senior Seismic

Specialists, the Earthquake Committee, and the Earthquake Engineering

Branch of the Office of Earthquake Engineering and Design Support should be

consulted for recommendations.

The global displacement demand estimate for Ordinary Standard bridges is

determined by linear elastic response spectrum analysis utilizing effective

section properties.

Equivalent Static Analysis is used to determine the displacement demand if a

dynamic analysis will not add significantly more insight into behavior. The


Equivalent Static Analysis is best suited for bridges or individual frames with

the following characteristics:

• Response primarily captured by the fundamental mode of vibration with

uniform translation

• Simply defined lateral force distribution (e.g., balanced spans,

approximately equal bent stiffness)

• Low skew

Elastic Dynamic Analysis is used to determine the displacement demand for all

other Ordinary Standard bridges.

The global displacement demand estimate shall include the effects of

soil/foundation flexibility if they are significant.

Following the Caltrans Seismic Design Criteria V1.2 the Inelastic Static

Analysis commonly referred to as “push over” analysis is to be used to

determine the reliable displacement capacities of a structure or frame as it

reaches its limit of structural stability.

The two-dimensional plane frame “push over” analysis of a bent or frame can

be simplified to a column model (fixed-fixed or fixed-pinned) if it does not cause

a significant loss in accuracy in estimating the displacement demands or the

displacement capacities. The effect of overturning on the column axial load

and associated member capacities must be considered in the simplified model.

Simplifying the demand and capacity models is not permitted if the structure

does not meet the following stiffness and period requirements:

a) Balanced Stiffness


Constant Width Frame Variable Width Frame

Stiffness ratio between any two bents within a frame or between any two columns within a bent.

0.5ei

ej

kk ≥

0.5

ei

iej

j

km

km

≥

Stiffness ratio between adjacent bents within a frame or between adjacent columns within a bent.

0.75ei

ej

kk ≥

0.75

ei

iej

j

km

km

≥

eik = The smaller effective bent or column stiffness im = Tributary mass of column or bent i

ejk = The larger effective bent or column stiffness jm = Tributary mass of column bent j

b) Balanced Frame Geometry

The ratio of fundamental periods of vibration for adjacent frames in the

longitudinal and transverse direction shall satisfy:

0.7i

j

TT ≥ where

iT = Natural period of the less flexible frame

jT = Natural period of the more flexible frame

In addition to the global analysis conducted on the overall structure to

determine displacement demands, a Stand-Alone analysis (i.e., shake down) is

performed in both the transverse and longitudinal directions. This analysis is

performed on individual frames that are separated by a superstructure

expansion joint.


In summary, Caltrans SDC v1.2 gives some latitude to the bridge engineer to

decide the type and amount of analysis to be conducted. This latitude is offset

by a quality control mechanism that is established in Caltrans and may not

exist in all other agencies nationwide. Furthermore, as Caltrans uses one set

of standard details for the whole bridge inventory, bridges in lower seismic

zones may end up with more stringent requirements and detailing that are not

needed in lower seismic zones. Therefore, in examining the Caltrans Practice,

it is deemed important to recognize that this Practice needs to be selectively

replicated for use by other states or agencies.

NCHRP 12-49 Range of Applicability of Analysis Each bridge is assigned a Seismic Hazard Level that is the highest level

determined by the valued of FvS1 or FaSs from Table 3-4 for the MCE event.

The spectral acceleration ordinates FvS1 and FaSs are illustrated in Figure 3-1.

Table 3-4: Seismic Hazard Levels Seismic Hazard Level

Value of FvS1 Value of FaSs

I FvS1≤0.15 FaSs≤0.15 II 0.15 < FvS1≤0.25 0.15 < FaSs≤0.35 III 0.25 < FvS1≤0.40 0.35 < FaSs≤0.60 IV 0.40 < FvS1 0.60 < FaSs

Notes:

1. For the purposes of determining the Seismic Hazard Level for Site Class

E Soils, the value of Fv and Fa need not be taken larger than 2.4 and 1.6

respectively when S1 is less than or equal to 0.10 and Ss is less than

0.25.

2. For the purposes of determining the Seismic Hazard Level for Site Class

F Soils, Fv and Fa values for Site Class E soils may be used with the

adjustment described in Note 1 above.


Figure 3-1: Design Spectrum

Each bridge is designed, analyzed and detailed for seismic effects in accordance

with Table 3-5. Seismic Design and Analysis Procedures (SDAP) are described

in Section 4 of the NCHRP 12-49 document. Minimum seismic design

requirements (SDR) for SDR 1 and 2, SDR 3 and SDR 4 are given in Sections 6,

7 and 8, respectively of NCHRP 12-49.

Table 3-5: Seismic Design and Analysis Procedures (SDAP) and Seismic Design Requirements (SDR)

Life Safety Seismic Hazard Level SDAP SDR

I A1 1 II A2 2 III B/C/D/E 3 IV C/D/E 4

SDAP A1 and A2 do not have dynamic analysis requirements. Bridges

qualifying for SDAP B do not require a seismic demand analysis but capacity

design principles and minimum design details are required. SDAP C is the

Capacity Spectrum Design Method. SDAP C combines a demand and capacity

Site Class Spectrum


analysis. The procedure applies only to bridges that behave essentially as a

single degree-of-freedom system. SDAP C is restricted to bridges with a very

regular configuration provided the abutments are not considered part of the

Earthquake Resistant System.

SDAP D is the Elastic Response Spectrum Method. SDAP D is a one-step

design procedure using an elastic (cracked section properties) analysis. Either

the Uniform Load or Multimode method of analysis may be used. The analysis

shall be performed for the 3% PE in 75-year/1.5 mean deterministic and the R-

Factors given in Tables 3-6 and 3-7. Capacity design principles shall be used

for column shear design and the design of all column connections and

foundation design. If sacrificial elements are part of the design (i.e., shear

keys) they shall be sized to resist the 50% PE in 75-year forces and the bridge

shall be capable of resisting the 3% PE in 75-year/1.5 mean deterministic

forces without the sacrificial elements (i.e., two analyses are required if

sacrificial elements exist in a bridge).

SDAP E is the Elastic Response Spectrum Method with Displacement Capacity

Verification. SDAP E requires an elastic (cracked section properties) response

spectrum analysis for the governing design spectra (50% PE in 75-year or 3%

PE in 75-year/1.5 mean deterministic) and P-Δ design check. The results of

these analyses shall be used to perform preliminary flexural design of plastic

hinges in columns and to determine the displacement of the structure. To take

advantage of the higher R-Factors in Table 3-6, displacement capacities shall

be verified using two-dimensional nonlinear static (pushover) analyses in the

principal structural directions. Design forces on substructure elements may be

reduced below those obtained for the 3% PE in 75-year event/1.5 mean

deterministic divided by the R-Factor. If sufficient displacement capacity

exists, the substructure design forces may be further reduced an additional

30% for a new sizing of the substructure members provided a second


displacement capacity is performed. Capacity design principles shall be used

to design the connection of the columns to the superstructure and foundation

and for column shear design.

Table 3-6: Base Response Modification Factors, RB, for Substructure Performance

Objective Life Safety Substructure Element

SDAP D SDAP E Wall Piers – larger dimension 2 3 Columns – Single and Multiple 4 6 Pile Bents and Drilled Shafts – Vertical Piles – above ground 4 6

Pile Bents and Drilled Shafts – Vertical Piles – 2 diameters below ground level – No owners approval required

1 1.5

Pile Bents and Drilled Shafts – Vertical Piles – in ground – Owners approval required. N/A 2.5

Pile Bents with Batter Piles N/A 2 Seismically Isolated Structures 1.5 1.5 Steel Braced Frame – Ductile Components 3 4.5 Steel Braced Frame – Nominally Ductile Components 1.5 2 All Elements for Expected Earthquake 1.3 1.3

Table 3-7: Response Modification Factors, R – Connections

Connection All Performance Objectives

Superstructure to abutment 0.8 Expansion joints within a span of the superstructure 0.8 Columns, piers, or pile bents to cap beam or superstructure 0.8

Columns or piers to foundations 0.8

Following the NCHRP 12-49 specifications, the displacement capacity

verification analysis shall be applied to individual piers or bents to determine


the lateral load-displacement behavior of the pier or bent. The capacity

evaluation shall be performed for individual piers or bents in the longitudinal

and transverse direction separately.

The capacity evaluation shall identify the component in the pier or bent that

first reaches its inelastic deformation capacity. The displacement at which the

first component reaches its maximum permitted deformation capacity defines

the maximum displacement capacity, Δcapacity for the pier or bent and this shall

exceed the factored displacement demand, Δ, according to the following

requirement:

1.5Δ ≤ Δ capacity

The model for the displacement capacity verification is based on nominal

capacities of the inelastic components. Stiffness and strength degradation of

inelastic components and effects of loads acting through the lateral

displacement shall be considered.

In examining SDAP E, which is based on a force reduction approach with

higher Response Modification factors RB provided a displacement verification is

performed, it is deemed important to reiterate the following:

a) NCHRP 12-49 recognizes a 30% further reduction of substructure design

forces provided a displacement capacity is performed. This statement is

in tune with current state of the practice highlighting the advantages of

using a displacement approach.

b) The displacement capacity is established based on the weakest

component; therefore no strength loss or degradation is considered

acceptable. Even though this practice can be adopted for simplicity, it is

considered extremely conservative when it is associated with the

following:


• The use of 1.5 factor for displacement demands established based

on the 3% PE in 75-year/1.5 mean deterministic event.

• The use of nominal properties for establishing capacities of

inelastic components.

As seen above, the NCHRP 12-49 recognizes the advantage of using a

displacement approach (a) but then retracts or offsets this advantage by

placing a 1.5 factor on the displacement demand.

In retrospect, the adoption of displacement capacity determination based on

the weakest component (i.e. no strength loss) is consistent with the state of the

practice aiming for some degree of simplicity in performing the push over

analysis. Furthermore, the use of nominal properties is also consistent with

current state of the practice.

In summary, the use of a 1.5 factor for displacement demand is considered

excessive and unwarranted considering the inherent conservatism in

establishing the displacement capacity.

SCDOT Specifications Range of Applicability of Analysis Similar to AASHTO, Division 1-A, the “Seismic Demand Analysis”

requirements in the SCDOT Specifications are derived based on the Seismic

Performance Category (SPC) and the regularity or irregularity of a given

bridge. The regularity requirements in the SCDOT Specifications are

identical to those from AASHTO Division 1-A Specifications.

The seismic hazard varies form very small to high across the State of South

Carolina. Therefore, for purposes of design, four Seismic Performance

Categories (SPC) are defined on the basis of the spectral acceleration for the


one second period of the 2%/50-year earthquake, SD1-SEE, and the Importance

Classification (IC) as shown in Table 3-6. The design response spectral

acceleration at 1.0-second period D1-SEES is shown in Figure 3-2. Different

degrees of complexity and sophistication of seismic analysis and design are

specified for each of the four Seismic Performance Categories.

Table 3-6: Seismic Performance Category (SPC)

Importance Classification (IC) Value of Spectral Acceleration, SD1-

SEE I II III

SD1-SEE<0.30g B B A 0.3g≤SD1-SEE<0.45g C C B 0.45g≤ SD1-SEE<0.6g D C C 0.6g≤ SD1-SEE D D C


Figure 3-2: Design response spectrum curve

The design spectrum for the FEE (10% in 50 years) and the SEE (2% in 50

years) were developed using the 1997 NEHRP Maps. The curves are anchored

to the 0.2 second mapped design spectral accelerations for Site Class B rock site. As shown in Figures 3-3 and 3-4 the following discrete points for SDS are

considered:

• SDS = 0.25g, 0.3g, and 0.35g for the FEE level.

• SDS = 0.4g, 0.5g, 0.6g, 0.8g, 1.0g, 1.25g, 1.5g, and 1.66g for the SEE

level.

Ss=0.60g, SEE(2%/50years)

0.00.10.20.30.40.50.60.70.80.91.0

0 1 2 3 4

SD_6ASD_6BSD_6CSD_6DSD_6E

Periods T (sec)

Site Class A B C D E

SD1-SEE


Figure 3-3: Design spectral response acceleration map short period – SDS–FEE for site class B.

Figure 3-4: Design spectral response acceleration map short period – SDS-SEE for site class B.


A family of curves for Soil Site Class A thru E referenced to the short period mapped design spectral acceleration SDS =1.0g is shown in Figure 3-5. The

curves were developed using both the short period and the one-second period

maps.

Ss=1.00g, SEE(2%/50years)

0.0

0.2

0.4

0.6

0.8

1.0

1.2

0 1 2 3 4

SD_4ASD_4BSD_4CSD_4DSD_4E

Periods T (sec)

Site Class A B C D E

Figure 3-5: Design spectra for site class A, B, C, D and E, 5% damping.

The Seismic Performance Category (SPC) definition in the SCDOT

Specifications differs from the AASHTO Division 1-A as follows:

1. The Seismic Performance Category (SPC) is based on the one-second

spectral acceleration at the SEE level Earthquake having a 2%

probability of exceedance in 50 years.

2. The Importance Classification (IC) in the SCDOT Specifications include

three categories of bridges, Critical, Essential, and Normal associated

with IC I, II, and III respectively while AASHTO Division 1-A has two

classifications, IC “I” for Essential bridges and IC “II” for other bridges.

The Specifications are for the design and construction of new bridges to resist

the effects of earthquake motions. The provisions apply to bridges of

conventional slab, beam girder and box girder superstructure construction with


spans not exceeding 500 ft (150 m). For other types of construction (suspension

bridges, cable-stayed bridges, arch type and movable bridges) and spans

exceeding 500 ft, the SCDOT shall specify and/or approve appropriate

provisions.

Seismic effects for box culverts and buried structures need not be considered,

except when they are subject to unstable ground conditions (e.g., liquefaction,

landslides, and fault displacements) or large ground deformations (e.g., in very

soft ground).

The provisions specified in the specifications are minimum requirements.

Additional provisions are needed to achieve higher performance criteria for

essential or critical bridges. Those provisions are site/project specific and are

tailored based on structure type.

No detailed seismic analysis is required for any single span bridge or for any

bridge in Seismic Performance Category A. For both single span bridges and

bridges classified as SPC A the connections must be designed for specified

forces and must also meet minimum support length requirements.

For SPC B, the displacement demand is checked implicitly against the capacity

without performing an elaborate pushover analysis to determine the

displacement capacity.

For SPC B the displacement capacity, cΔ , is easily obtained for each column

using the following expression:

X( ) 5.3 (.0013)100cHftΔ = ∗ ∗

where,

X DH= Λ


Λ is a fixity factor for the column equal to: a. Λ = 1 for fixed-free (pinned on one end). b. Λ = 2 for fixed top and bottom.

D = Column Diameter (ft.). H = Height from top of footing to C.G. of superstructure (ft.).

In summary, the objective in developing the new SCDOT Seismic Design

Specification is to balance the required numerical computations to the severity

of the seismic hazard established in SPC A, B, C and D.

Figure 3-6: SPC B Drift Criteria (SCDOT)

Range of Applicability of “Seismic Demand Analysis” Seismic Analysis is conducted in regions where PGA > 9%g following AASHTO

Division 1-A. For illustration of difference in the extent of regions requiring

“Seismic Demand Analysis” following AASHTO Division 1-A and the

recommended specifications, a comparison is performed on the area

surrounding the New Madrid fault and South Carolina. These two areas are

% c

HΔ

DH


considered since they represent the largest increase in seismic demands when

considering larger return period up to the proposed return period of 2500 years

adopted in NCHRP 12-49. The reference to the increase in seismic demands is

made in relation to AASHTO Division 1-A State of the Practice. Figure 3-7

shows the AASHTO region of required seismic analysis.

Figure 3-7: AASHTO Region of Required Seismic Analysis PGA > 9%

With the selection of the one-second spectral design acceleration spectrum

1a DSS − , the regions of required “Seismic Demand Analysis” vary depending on

the site class (i.e., type of soil) as established in NEHRP 1997 and adopted in

the NCHRP 12-49 document.

Considering a Site Class B for the New Madrid/South Carolina area, the

contour shown in Figure 3-8 in bold black establishes the region of required

“Seismic Demand Analysis” corresponding to the proposed target design

hazard. Based on preliminary selection, the target design hazard is calibrated


at 2/3 of the spectrum established based on the 2002 USGS hazard maps for a

probability of 2% exceedance in 50 years. The proposed region for Site Class B

of required “Seismic Demand Analysis” is substantially smaller than the

corresponding AASHTO Division 1-A region.

For comparison, the region of required “Seismic Demand Analysis” for Site

Class D is shown in Figure 3-9 for the same area. The proposed region for Site

Class D shows relatively small reduction to the corresponding AASHTO

Division 1-A region.

Figure 3-8: Region of Required “Seismic Demand Analysis” for the Target Design Hazard, Site Class B


Figure 3-9: Region of Required “Seismic Demand Analysis” for the Target Design Hazard, Site Class D

In comparing the proposed Guidelines to current requirements in AASHTO

Division 1-A, the proposed guidelines fulfill the objective of maintaining to

reducing the number of bridges subjected to “Seismic Demand Analysis.”

Range of Applicability of “Seismic Capacity Analysis” “Seismic Capacity Analysis” is performed for SPC B, C, and D. This analysis is

incremental as follows:

1. SPC B

Implicit displacement capacity check is required similar to SCDOT

Specifications. No Capacity Design is required. This category is

associated with small displacement demand and drifts. Given the

relatively small demands and based on a minimum level of detailing


identified as Tier II, the bridge structure is expected to perform well,

with its members targeted to remain essentially elastic at ductility level

less than two.

2. SPC C

Implicit displacement capacity deck is required similar to SPC B;

however, setting the acceptance capacity criteria to a higher level of

ductility relative to SPC B. Given the moderate displacement demands

on the bridge structure, a Capacity Design analysis is required in order

to ensure adequate force distribution and proper design for hinging

mechanism. Considering the moderate acceptance criteria, an

incremental Tier III level of detailing is required. An elaborate

pushover analysis is not warranted.

3. SPC D

A pushover analysis is required for this category as a high level of

ductility is expected. Proper distribution of forces and Capacity design

requirements need to be satisfied to ensure a reliable comparison of the

structure displacement capacity against the displacement demands. A

Tier IV level of detailing is required in SPC D.

The contours presented in Section 3.3.1 coincide with SPC B contours. The

same area identified in Section 3.3.1 is used to show the region of required

pushover analysis. By illustrating the region of minimum “Seismic Capacity

Analysis” associated with SPC B and the region of maximum “Seismic Capacity

Analysis” associated with SPC D, the reader can appreciate the incremental

approach proposed for the Guidelines.


Figures 3-10 and 3-11 shows the Region of required Maximum “Seismic

Capacity Analysis” for the target design hazard for Site Class B and Site Class

D, respectively. A pushover analysis is required in this region.

Figure 3-10: Region of Required Maximum “Seismic Capacity

Analysis” for the Target Design Hazard, Site Class B


Figure 3-11: Region of Required Maximum “Seismic Capacity

Analysis” for the Target Design Hazard, Site Class D

As shown in Figures 3-10 and 3-11, the region where a pushover analysis is

required is chosen very selectively and is tuned to displacement demands on

the bridge structure. The proposed guidelines aim at fulfilling Task F3-4

objective No. 2 identifying range of applicability for NO Analysis or Limited

Analysis. This approach is a by-product of the steps taken in the NCHRP 12-

49 proposed guidelines and the SCDOT Specifications combined with practical

applications developed and gained in the seismic design practice over the last

decade.


TASK 4

4 SELECT THE MOST APPROPRIATE DESIGN PROCEDURE FOR STEEL

4.1 General The objective of this task is to select the most appropriate design procedure

(i.e., displacement or force based) for a bridge with a steel superstructure and

to examine both the NCHRP 12-49 and SCDOT using a trial design.

This task emphasis is to address analysis and design requirements for a bridge

with steel girders. The seismic design of a bridge system and components

needs to encompass two categories:

a. System with a restrained connection between the superstructure and the

substructure.

b. System with an unrestrained connection between the superstructure

and the substructure.

Emphasis on the load path and design of various components must be

established recognizing that a lack of consensus may still be present on some

issues.

The 2nd Edition of the LRFD Specifications included for the first time a new

section about the seismic lateral load distribution that discusses the seismic

load path. The focus for these criteria is steel bridges since they normally do

not have monolithic connections as the structural concrete box girder bridges.


The specifications require that a clear and a straightforward load path from

the superstructure to the substructure should exist. All elements that lie in

the load path are primary seismic members and should be designed to stay

elastic during severe ground motions. Diaphragms and cross-frames, lateral

bracing and bearings should be part of the seismic load path. The

specifications suggest that if these members were designed to respond in a

ductile manner or allow some movements, the damage will be limited.

However, the specifications require that the cross frames and end diaphragms

to stay elastic during earthquakes.

On the contrary, NCHRP 12-49 and SCDOT seismic specifications allow for

ductility (i.e., inelastic action) in the superstructure. None of the specifications

contains a uniform and a complete list of primary members identification for

the seismic load path.

4.2 Design Examples Two design examples were selected from the work done by Itani and Sedarat in

2000 entitled “Seismic Analysis and Design of the AISI LRFD Design

Examples of Steel Highway Bridges”. This effort was a continuation to the

1996 AISI published Vol. II Chapter 1B of the Highway Structures, Design

Handbook, “Four LRFD Design Examples of Steel Highway Bridges.” In 1996

these design examples covered the gravity design of the superstructure

according to the AASHTO LRFD Bridge Specifications. The main two

purposes in examining this report is to:

1. Identify the performance objective for seismic design of steel girder

structures.


2. Identify the specifications utilized for proper completion of the design

process.

Appendix 4A includes the portion of AISI-LRFD report used in this task. This

appendix contains the design calculations as well as the drawing showing

details of each of the two bridges.

Example 1 is a Simple-Span Composite I Girder. The design process shown in

the report includes:

1. Calculation of lateral load at the end cross-frame.

2. The design of the top strut.

3. The design of the diagonal member.

4. The design of the bottom strut.

Two important aspects of the design process are identified:

a. The end cross-frame is designed for the full seismic force with no

reduction of this force assuming a restrained condition of the bridge (i.e.,

shear keys capable of sustaining the full seismic force).

b. A single angle bracing is used for the diagonal member of the end-cross-

frame. As this practice is typical and favored for ease of construction,

the design process for a single angle bracing needs to be referenced or

included for clarity of use by the bridge engineer. AISC has a stand

alone document on “LRFD Design Specification for Single-Angle

Members” that can be included or referenced in the Specifications. This

document is attached in Appendix 4B.

Example 2 is a Two-Span Continuous Composite I Girder. The design process

shown in the report includes:


1. Calculation of the lateral load at the bent cross-frame.

2. The design of the plate girder connections to the R/C Deck.

3. Design of the top strut.

4. Design of the diagonal member.

5. Design of the bottom strut.

6. Calculation of superstructure lateral capacity.

Three important aspects of the design process are identified:

a. The bent cross-frame is designed to ensure column hinging mechanism

assuming a restrained condition of the superstructure to the bent.

b. The load path from the deck to the girders or the top strut is checked.

c. Double angles with stitches are used for the top strut and the diagonal

member due to the higher seismic demand on this bridge located in

seismic zone 4. AISC LRFD Specifications Chapter E applies to compact

and non-compact prismatic members subject to axial compression

through the centroidal axis. The design process for members with

stitches is also included. The inclusion or reference of the specifications

is needed for clarity and consistency of use by the bridge engineer.

4.3 Load Path and Performance Criteria Specifications regarding the load path for a slab-on-girder bridge are examined

using SCDOT and NCHRP 12-49 documents. SCDOT specifications has a

general section on load path while NCHRP 12-49 has a section only on “Ductile

End-Diaphragm in Slab-on-Girder Bridge.” The section from SCDOT


specifications is included in Appendix 4D while the section from NCHRP 12-49

is included in Appendix 4E.

As seen from examining both of these documents, it is important to

differentiate between ordinary bracing referred to in the SCDOT specifications

and specially detailed diaphragms referred to in the NCHRP 12-49. The AISC

provisions limit the force reduction factor R to 3 for ordinary bracing that is a

part of a seismic resisting system not satisfying the special seismic provisions.

It is proposed to adopt the AISC limit for an R reduction factor of 3. Special

end-diaphragm addressed in NCHRP 12-49 will be considered for bracing

system with a reduction factor, R, greater than 3 as stipulated in the AISC

provisions.

Section 7.1 and 7.2 of SCDOT specifications will be enhanced for general

treatment of load path and Performance Criteria. The following is a

duplication of these two sections.

General

“The Engineer shall demonstrate that a clear, straight-forward load

path to the substructure exists and that all components and connections

are capable of resisting the imposed seismic load effects consistent with

the chosen load path.

The flow of forces (see Figure 4.1) in the assumed load path must be

accommodated through all affected components and details including,

but not limited to, flanges and webs of main beams or girders, cross-

frames, steel-to-steel connections, slab-to-steel interfaces, and all

components of the bearing assembly from bottom flange interface

through the confinement of anchor bolts or similar devices in the


substructure. The substructure shall also be designed to transmit the

imposed force effects into the ground.

a) Pile Footing b) Drilled Shaft

Figure 4-1: Seismic Load Path and Affected Components

The design of end diaphragms and cross-frames shall include analysis

cases with horizontal supports at an appropriate number of bearings,

consistent with Section 7.7.2 of SCDOT Specifications.

A viable load path shall be established to transmit the inertial loads to

the foundation based on the stiffness characteristics of the deck,

diaphragms, cross-frames, and lateral bracing. Unless a more refined

analysis is made, an approximate load path shall be assumed as follows:

The following requirements apply to bridges with either: Formatted: Bullets and Numbering


• a concrete deck that can provide horizontal diaphragm action or

• a horizontal bracing system in the plane of the top flange.

The seismic loads in the deck shall be assumed to be transmitted

directly to the bearings through end diaphragms or cross-frames. The

development and analysis of the load path through the deck or through

the top lateral bracing, if present, shall utilize assumed structural

actions analogous to those used for the analysis of wind loadings.”

Criteria

“This section is intended for design of superstructure steel components.

Those components are classified into two categories: Ductile and

Essentially Elastic. Based on the characteristics of the bridge structure,

the designer has one of three choices:

• Type 1 – Design a ductile substructure with an essentially elastic

superstructure.

• Type 2 – Design an essentially elastic substructure with a ductile

superstructure.

• Type 3 – Design an elastic superstructure and substructure with

a fusing mechanism at the interface between the superstructure

and the substructure.

For Type 1 choice, the designer shall refer back to Section 8 of this

document on designing for a ductile substructure. For Type 2 choice, the

design of the superstructure is accomplished using a force reduction

approach. Those factors are used for the design of transverse bracing

members, top laterals and bottom laterals. The reduction factors shown

in Table 7.1 shall be used.


Table 4-1: Reduction Factors for Steel Superstructure Bracings

Essential or

Critical Bridges Normal Bridges

Functional Evaluation 1 2 Safety Evaluation 2 4

For Type 3 choice, the designer shall assess the overstrength capacity for

the fusing interface including shear keys and bearings, then design for

an essentially elastic superstructure and substructure. The minimum

overstrength lateral design force shall be calculated using an

acceleration of 0.4 g or the elastic seismic force whichever is smaller. If

isolation devices are used, the superstructure shall be designed as

essentially elastic (see Section 7.6 of SCDOT Specifications).

In this section, reference to an essentially elastic component is used

where the force demand to capacity ratio of any member in the

superstructure is less than 1.3.”

4.4 Summary In reviewing the SCDOT specifications, the NCHRP 12-49, and the AISI LRFD

examples, the following recommendations are proposed:

1. Adopt AISC LRFD Specifications for design of single angle members and

members with stitches.

2. Allow for three types of a bridge structural system as adopted in SCDOT

Specifications.

3. Adopt a force reduction factor of 3 for design of normal end cross-frame.

3


4. Adopt NCHRP 12-49 for design of “Ductile End-Diaphragm” where a

force reduction factor greater than 3 is desired.


TASK 4

APPENDIX 4A

CENTER FOR CIVIL ENGINEERING AND EARTHQUAKE RESEARCH

Report No. CCEER 00-8

Seismic Analysis and Design of the

AISI LRFD Design Examples

of Steel Highway Bridges

Ahmad M. Itani

Hassan Sedarat

• Reno

Engineering Research and Development Center

College of Engineering

University of Nevada, Reno

1325 NCHRP 20-7(193) Task 6 Report.doc 4B-1

TASK 4

APPENDIX 4B

LOAD AND RESISTANCE FACTOR DESIGN SPECIFICATION FOR

SINGLE-ANGLE MEMBERS

1325 NCHRP 20-7(193) Task 6 Report.doc 4C-1

TASK 4

APPENDIX 4C

CHAPTER E – COLUMNS AND OTHER COMPRESSION MEMBERS

1325 NCHRP 20-7(193) Task 6 Report.doc 4D-1

TASK 4

APPENDIX 4D

SOUTH CAROLINA DEPARTMENT OF TRANSPORTATION

SPECIFICATIONS FOR HIGHWAY BRIDGES

1325 NCHRP 20-7(193) Task 6 Report.doc 4E-1

TASK 4

APPENDIX 4E

NCHRP 12-49


7.7 Structural Steel Design Requirements

7.7.8.2 Ductile End-Diaphragm in Slab-on-Girder Bridge

Ductile end-diaphragms in slab-on-girder bridges can be designed to be the

ductile energy dissipating elements for seismic excitations in the transverse

directions of straight bridges provided that:

a. specially detailed diaphragms capable of dissipating energy in a stable

manner and without strength degradation upon repeated cyclic testing

are used;

b. only ductile energy dissipating systems whose adequate seismic

performance has been proven through cycling inelastic testing are used;

c. the design considers the combined and relative stiffness and strength of

end-diaphragms and girders (together with their bearing stiffeners) in

establishing the diaphragms strength and design forces to consider for

the capacity protected elements;

d. the response modification factor to be considered in design of the ductile

diaphragm is given by:

1

DED

SUB

DED

SUB

KKR KK

μ⎛ ⎞+⎜ ⎟⎜ ⎟=⎜ ⎟+⎜ ⎟⎝ ⎠

(7.7.8.2-1)

where μ is the ductility capacity of the end-diaphragm itself, and

KDED/KSUB is the ratio of the stiffness of the ductile end-diaphragms and

substructure (unless the designer can demonstrate otherwise, μ should

not be taken greater than 4);

e. all details/connections of the ductile end-diaphragms are welded;


f. the bridge does not have horizontal wind-bracing connecting the bottom

flanges of girders, unless the last wind bracing panel before each support

is designed as a ductile panel equivalent and in parallel to its adjacent

vertical end-diaphragm; and

g. an effective mechanism is present to ensure transfer of the inertia-

induced transverse horizontal seismic forces from the slab to the

diaphragm.

Overstrength factors to be used to design the Capacity Protected Elements

depend on the type of ductile diaphragm used, and shall be based on available

experimental research results.

8.7 Structural Steel Design Requirements

8.7.8.2 Ductile End-Diaphragm in Slab-on-Girder Bridge

Ductile end-diaphragms in slab-on-girder bridges can be designed to be the

ductile energy dissipating elements for seismic excitations in the transverse

directions of straight bridges provided that:

a. Specially detailed diaphragms capable of dissipating energy in a stable

manner and without strength degradation upon repeated cyclic testing

are used;

b. Only ductile energy dissipating systems whose adequate seismic

performance has been proven through cycling inelastic testing are used;

c. Design considers the combined and relative stiffness and strength of

end-diaphragms and girders (together with their bearing stiffeners) in

establishing the diaphragms strength and design forces to consider for

the capacity protected elements;


d. The response modification factor to be considered in design of the ductile

diaphragm is given by:

1

DED

SUB

DED

SUB

KKRKK

μ⎛ ⎞+⎜ ⎟⎜ ⎟=⎜ ⎟+⎜ ⎟⎝ ⎠

(8.7.8.2-1)

where μ is the ductility capacity of the end-diaphragm itself, and

KDED/KSUB is the ratio of the stiffness of the ductile end-diaphragms and

substructure; unless the engineer can demonstrated otherwise, μ should

not be taken greater than 4;

e. All details/connections of the ductile end-diaphragms are welded.

f. The bridge does not have horizontal wind-bracing connecting the bottom

flanges of girders, unless the last wind bracing panel before each support

is designed as a ductile panel equivalent and in parallel to its adjacent

vertical end-diaphragm.

g. An effective mechanism is present to ensure transfer of the inertia-

induced transverse horizontal seismic forces from the slab to the

diaphragm.

Overstrength factors to be used to design the capacity-protected elements

depend on the type of ductile diaphragm used, and shall be based on available

experimental research results.


TASK 5

5 RECOMMEND LIQUEFACTION DESIGN PROCEDURE

5.1 Objective The objective of this task is to review applicable recent research and

information currently available on liquefaction and to recommend design

procedures consistent with the “Displacement Approach” adopted for the

proposed specifications. The proposed approach is to streamline the provisions

provided by NCHRP 12-49 in one separate section or appendix. The extent of

the provisions are established in light of the overall methodology and

framework established in the tasks:

a. Task 2 – Finalize Seismic Hazard Level

b. Task 3 – Expand the Extent of the “No Analysis” Zone

5.2 NCHRP 12-49 Liquefaction Design Requirements

NCHRP 12-49 added considerable amount of information for the provisions on

liquefaction. The general design approach outlined in NCHRP 12-49 consists

of the following:

1. Specific design requirements for piled foundations, drilled shafts and

spread footing exposed to liquefaction with no lateral flow.


2. For the above mentioned types of foundations subjected to lateral flow,

proceed with the following steps:

a. Design the piles or spread footings to resist the forces generated

by the lateral spreading.

b. If the structure cannot be designed to resist the forces, assess

whether the structure is able to tolerate the anticipated

movements and meet the geometric and structural constraints of

the provisions.

c. If the structure cannot meet the performance requirements of the

provisions, assess the costs and benefits of various mitigation

measures to minimize the movements to a tolerable level to meet

the desired performance objective.

Appendix 5A contains NCHRP 12-49 requirements for Foundation Design and

Liquefaction Design for SDR 3 (Chapter 7 of NCHRP 12-49) and SDR 4

(Chapter 8 of NCHRP 12-49).

In adopting a “Displacement” Approach for the new specifications and

considering a No Collapse Criteria, the new specifications will be altered in

determining the adequacy of the structure based strictly on the displacement

demands. Minimum strength requirements would be introduced to minimize

the effects of any geometric non-linearities. Provisions related to steps a) and

c) mentioned above and related to a “Force Based Approach” will be eliminated

for consistency with the overall approach.


5.3 Damage Severity in Past Earthquakes In order to gain insight on the damage severity on bridges during past

earthquakes, the catalog on the seismic performance of bridges in the presence

of liquefaction-induced ground displacement authored by Stephen A.

Dickenson, Nason J. McCullough, Mark G. Barkau, and Bryan J. Wavra is

used.

Each bridge in this catalog has been assigned a damage severity rating DSR

according to the classification scheme outlined in Table 5-1. A summary of this

catalog is shown in Table 5-2.

Table 5-1: Damage Severity Description

DAMAGE SEVERITY RATING (DSR) DAMAGE DESCRIPTION

DSR = 3 Severe Damage: Abutments moved streamward and/or markedly subsided;

piers shifted, tilted, settled, or fell over. Large movements of foundation units. Substructure rendered unsalvable.

DSR = 2 Moderate Damage: Distinct and measurable net displacements as in

previous category but to a lesser degree, so that the substructure could perhaps be repaired and used to support a new superstructure.

DSR = 1 Minor Damage: Evidence of foundation movements such as cracked

backwalls, split piles, and closed expansion devices, but net displacements small and substructure serviceable. Minor abutment slumping.

DSR = 0 Nil Damage: No evidence of foundation displacements.


Table 5-2: Damage Severity Rating vs. Earthquake Magnitude

Earthquake Mw DSR Minimum DSR Maximum1995 Manzanillo, Mexico 7.5 1 11995 Hyogo-ken-Nanbu (Kobe), Japan 6.9 0 31994 Northridge 6.7 0 01994 Mindoro Island, Phillipines 7.1 3 31993 Island of Guam 8.4 1 11993 Hokkaido Nansei-oki, Japan 7.8 0 21992 Erzincan, Turkey 6.7 1 21991 Costa Rica 7.4 0 31990 Luzon, Phillipines 7.9 1 31989 Loma Prieta 6.9 0 11983 Nihonkai-Chuba 7.7 0 11980 El-Asnam, Algeria 7.2 1 11979 Imperial Valley, California 6.5 1 11978 Miyagi-Ken-oki, Japan 7.3 1 21976 Mindanao, Phillipines 7.9 1 11976 Tangshan, China 7.8 3 31975 Haicheng, China 7.2 3 31968 Ebino 6.1 1 11964 Alaska 9.21964 Niigata, Japan 7.3 3 31948 Fukui, Japan 6.9 2 31923 Kanto, Japan 6.9 2 31906 San Francisco 7.9 0 31886 Charleston, South Carolina ? 3 3

The full catalog is included in Appendix 5B. As seen from Table 5-2 a DSR

equal to 2 corresponding to moderate damage is associated with an earthquake

magnitude Mw of 6.7 or higher while a DSR equal to 3 corresponding to severe

damage is associated with an earthquake magnitude Mw of 6.9 or higher.

5.4 Proposed Liquefaction Design Requirements An evaluation of the potential for and consequences of liquefaction within near

surface soil shall be made in accordance with the following requirements:

Liquefaction is required for a bridge in SPC D unless one of the following

conditions is met:


a. The mean magnitude for the 5% PE in 50-year event is less than 6.5.

b. The mean magnitude for the 5% PE in 50-year event is less than 6.7

and the normalized Standard Penetration Test (SPT) blow count

[(N1)60] is greater than 20.

Procedures given in Appendix D of NCHRP 12-49 and adopted from California

DMG Special Publication 117 shall be used to evaluate the potential for

liquefaction.

If it is determined that liquefaction can occur at a bridge site then the bridge

shall be supported on deep foundations or the ground improved so that

liquefaction does not occur. If liquefaction occurs then the bridge shall be

designed and analyzed in two configurations as follows:

1. Nonliquefied Configuration: The structure shall be analyzed and

designed, assuming no liquefaction occurs using the ground response

spectrum appropriate for the site soil conditions.

2. Liquefaction Configuration: The structure as designed in Nonliquefied

Configuration above shall be reanalyzed and redesigned, if necessary,

assuming that the layer has liquefied and the liquefied soil provides

whatever residual resistance is appropriate (i.e., “p-y curves” or modulus

of sub-grade reaction values for lateral pile response analyses consistent

with liquefied soil conditions). The design spectra shall be the same as

that used in Nonliquefied Configuration unless a site-specific response

spectra has been developed using nonlinear, effective stress methods

(e.g., computer program DESRA or equivalent) that properly account for

the buildup in pore-water pressure and stiffness degradation in

liquefiable layers. The reduced response spectra resulting from the site-


specific nonlinear, effective stress analyses shall not be less than 2/3’s of

that used in Nonliquefied Configuration.

The Designer shall cover explicit detailing of plastic hinging zones for both

cases mentioned above since it is likely that locations of plastic hinges for the

Liquefied Configuration are different than locations of plastic hinges for the

Non-Liquefied Configuration. Design requirements of SPC “D” including shear

reinforcement shall be met for the Liquefied and Non-Liquefied Configuration.

5.5 Summary The following list highlights the main proposed liquefaction design

requirements:

a. Liquefaction design requirements are applicable to SPC “D”.

b. Liquefaction design requirements are dependent on the mean magnitude

for the 5% PE in 50-year event and the normalized Standard

Penetration Test (SPT) blow count [(N1)60].

c. If liquefaction occurs, then the bridge shall be designed and analyzed for

the Liquefied and Non-Liquefied configurations.

Design requirements for lateral flow are still debatable and have not reached a

consensus worth comfortably adopting. The IAI geotechnical team is preparing

a task to address this topic and complement the effort produced in the NCHRP

12-49 document.


TASK 5

APPENDIX 5A

NCHRP 12-49

PART I: SPECIFICATIONS 2003 GUIDELINES FOR THE SEISMIC DESIGN OF HIGHWAY BRIDGES

SECTION 7 72 MCEER/ATC-49

7.4 FOUNDATION DESIGN REQUIREMENTS

7.4.1 Foundation Investigation

7.4.1.1 General

A subsurface investigation, including borings and laboratory soil tests, shall be conducted in accordance with the provisions of Appendix B to provide pertinent and sufficient information for the determination of the Site Class of Article 3.4.2.1. The type and cost of foundations should be considered in the economic, environmental, and aesthetic studies for location and bridge type selection.

7.4.1.2 Subsurface Investigation

Subsurface explorations shall be made at pier and abutment locations, sufficient in number and depth, to establish a reliable longitudinal and transverse substrata profile. Samples of material encountered shall be taken and preserved for future reference and/or testing. Boring logs shall be prepared in detail sufficient to locate material strata, results of penetration tests, groundwater, any artesian action, and where samples were taken. Special attention shall be paid to the detection of narrow, soft seams that may be located at stratum boundaries.

7.4.1.3 Laboratory Testing

Laboratory tests shall be performed to determine the strength, deformation, and flow characteristics of soils and/or rocks and their suitability for the foundation selected. In areas of higher seismicity (e.g., SDR 3, 4, 5, and 6), it may be appropriate to conduct special dynamic or cyclic tests to establish the liquefaction potential or stiffness and material

damping properties of the soil at some sites, if unusual soils exist or if the foundation is supporting a critical bridge.

7.4.2 Spread Footings

Spread footing foundations for SDR 3 shall be designed using column loads developed by capacity design principles or elastic seismic loads, in accordance with Strength Limit State requirements given in Article 10.6.3 of the LRFD Bridge Design Specifications (AASHTO, 1998a, and subsequent amendments), hereinafter referred to as the AASHTO LRFD provisions. It will not normally be necessary to define spring constants for displacement evaluations or moment-rotation and horizontal force-displacement behavior of the footing-soil system (Article 5.3.4). Checks shall also be made to confirm that flow slides and loss of bearing support from liquefaction do not occur (Article 7.6).

7.4.2.1 Moment and Shear Capacity

The overturning capacity of the spread footings shall be evaluated using 1.0 times the nominal moment capacity of the column (Article 4.8) or the elastic seismic design force within the column, whichever is less. Procedures for Strength Limit State Design given in Article 10.6.3 of the AASHTO LRFD provisions shall be used when performing this evaluation.

A triangular elastic stress distribution within the soil shall be used. The peak bearing soil pressure for the triangular distribution shall not exceed the ultimate bearing capacity of the soil at the toe of the footing. The width of maximum liftoff shall be no greater than 1/2 of the footing width for moment loading in each of the two directions treated separately.



If a non-triangular stress distribution occurs or if the liftoff is greater 1/2 of the footing, either the footing shall be re-sized to meet the above criteria or special studies shall be conducted to demonstrate that non-triangular stress pressure distribution or larger amounts of liftoff will not result in excessive permanent settlement during seismic loading. The special studies shall include push-over analyses with nonlinear foundation springs for SDAP E conditions.

No shear capacity evaluation of the footing will normally be required for SDR 3.

7.4.2.2 Liquefaction Check

An evaluation of the potential for liquefaction within near-surface soil shall be made in accordance with requirements given in Article 7.6 and Appendix D of these Specifications. If liquefaction is predicted to occur for the design earthquake, the following additional requirements shall be satisfied: Liquefaction without Lateral Flow or Spreading

For sites that liquefy but do not undergo lateral flow or spreading, the bottom of the spread footing shall be located either below the liquefiable layer or at least twice the minimum foundation width above the liquefiable layer. If liquefaction occurs below the footing, settlements resulting from the dissipation of excess porewater pressures shall be established in accordance with procedures given in Appendix D.

If the depth of the liquefiable layer is less than twice the minimum foundation width, spread footing foundations shall not be used, unless • ground improvement is performed to mitigate

the occurrence of liquefaction, or • special studies are conducted to demonstrate

that the occurrence of liquefaction will not be

detrimental to the performance of the bridge support system.

Before initiating any evaluations of ground improvement alternatives or before conducting special studies, the potential applicability of deep foundations as an alternative to spread footings shall be discussed with the owner. Liquefaction with Lateral Flow or Spreading

If lateral flow or lateral spreading is predicted to occur, the amount of displacement associated with lateral flow or lateral spreading shall be established in accordance with procedures given in Appendix D. Once the deformation has been quantified, the following design approach shall be used. • Determine whether the spread footings can be

designed to resist the forces generated by the lateral spreading without unusual size or design requirements.

• If the footing cannot resist forces from lateral spreading or flow, assess whether the structure is able to tolerate the anticipated movements and meet the geometric and structural constraints of Table C3.2-1. The maximum plastic rotation shall be as defined in Article 7.7.9 and 7.8.6.

• If the structure cannot meet the performance requirements of Table 3.2-1, assess the costs and benefits of various mitigation measures to minimize the movements to a level that will meet the desired performance objective. If a higher performance is desired so that the spread footings will not have to be replaced, the allowable plastic rotations for concrete columns given in Article 7.7.9 and 7.8.6 shall be met.

The owner shall be apprised of and concur with the approach used for the design of spread footing foundations for lateral flow or lateral spreading conditions.



7.4.3 Driven Piles

7.4.3.1 General

Resistance factors for pile capacities shall be as specified in Table 10.5.4-2 of the AASHTO LRFD provisions, with the exception that resistance factors of 1.0 shall be used for seismic loads.

For the effect of settling ground and downdrag loads, unfactored load and resistance factors (γ = 1.0; φ = 1.0) shall be used, unless required otherwise by the owner.

Batter piles shall not be used where downdrag loads are expected unless special studies are performed.

For seismic loading the groundwater table location shall be the average groundwater location, unless the owner approves otherwise.

7.4.3.2 Design Requirements

Driven pile foundations subject to SDR 3 shall be designed for column moments and shears developed in accordance with the principles of capacity design (Article 4.8) or the elastic design forces, whichever is smaller. The Strength Limit State requirements given in Article 10.7.3 of the AASHTO LRFD provisions shall apply for design.

With the exception of pile bents, it will not normally be necessary to define spring constants for displacement evaluations or moment-rotation and horizontal force-displacement analyses for SDR 3 (Article 5.3.4). For pile bents, the estimated depth of fixity shall be used in evaluating response.

If liquefaction is predicted at the site, the potential effects of liquefaction on the capacity of the driven pile foundation system

shall be evaluated in accordance with procedures given in Article 7.4.3.4.

7.4.3.3 Moment and Shear Design

The capacity of the geotechnical elements of driven pile foundations shall be designed using 1.0 times the nominal moment capacity of the column or the elastic design force within the column (Article 4.8), whichever is smaller. Unfactored resistance (φ = 1.0) shall be used in performing the geotechnical capacity check. The loads on the leading row of piles during overturning shall not exceed the plunging capacity of the piles. Separation between the pile tip and the soil (i.e. gapping) shall be allowed only in the most distant row of piles in the direction of loading. Forces on all other rows of piles shall either be compressive or not exceed the nominal tension capacity of the piles.

If the plunging capacity of the leading pile is exceeded or if uplift of other than the trailing rows of piles occurs (see Figure C3.3.1-2), special studies shall be conducted to show that performance of the pile system is acceptable. These studies shall be performed only with the prior consent of the owner and SDAP E is required.

Structural elements of pile foundations shall be designed using the overstrength moment capacity of the column or the elastic design force within the column (Article 4.8), whichever is smaller.

The maximum shear force on the pile(s) shall be less than the structural shear capacity of the piles.

7.4.3.4 Liquefaction Check

An evaluation of the potential for liquefaction shall be made in accordance with requirements given in Article 7.6 and



Appendix D of these Specifications. If liquefaction is predicted to occur for the design earthquake, the following additional requirements shall be satisfied: Liquefaction without Lateral Flow or Spreading

• The pile shall penetrate beyond the bottom of the liquefied layer by at least 3 pile diameters or to a depth that axial and lateral pile capacity are not affected by liquefaction of the overlying layer, whichever is deeper.

• The shear reinforcement in a concrete or pre-stressed concrete pile shall meet the requirements of Sec 7.8.2.3 from the pile or bent cap to a depth of 3 diameters below the lowest liquefiable layer.

• Effects of downdrag on the pile settlements shall be determined in accordance with procedures given in Appendix D.

Liquefaction with Lateral Flow or Lateral Spreading

• Design the piles to resist the forces generated by the lateral spreading.

• If the forces cannot be resisted, assess whether the structure is able to tolerate the anticipated movements and meet the geometric and structural constraints of Table C3.2-1. The maximum plastic rotation of the piles shall be as defined in Article 7.7.9 and Article 7.8.6.

• If the structure cannot meet the performance requirements of Table 3.2-1, assess the costs and benefits of various mitigation measures to reduce the movements to a tolerable level to meet the desired performance objective. If a higher performance is desired so that the piles will not have to be replaced, the allowable plastic rotations of Articles 7.7.9.2 and 7.8.6.2 shall be met.

7.4.4 Drilled Shafts

Procedures identified in Article 7.4.3.2, including those for liquefaction and dynamic settlement, shall be applied with the exception that the ultimate capacity in compression or uplift loading for single shaft foundations in SDR 3 shall not be exceeded during maximum seismic loading without special design studies and the owner’s approval. The flexibility of

the drilled shaft shall also be represented in the design using either the estimated depth of fixity or soil springs in a lateral pile analysis.

Diameter adjustments shall be considered during lateral load analyses of shafts with a diameter greater than 600 mm if the shaft is free to rotate, as in the case of a column extension (i.e., no pile cap). Contributions from base shear shall also be considered.


MCEER/ATC-49 79 SECTION 7

7.6 LIQUEFACTION DESIGN REQUIREMENTS

7.6.1 General

An evaluation of the potential for and consequences of liquefaction within near-surface soil shall be made in accordance with the following requirements. A liquefaction assessment is required unless one of the following conditions is met or as directed otherwise by the owner. • Mean magnitude for the Maximum

Considered Earthquake (MCE) is less than 6.0 (Figures 7.6.1-1 to 7.6.1-4);

• Mean magnitude of the MCE is less than 6.4 and equal to or greater than 6.0, and the normalized Standard Penetration Test (SPT) blow count [(N1)60] is greater than 20;

• Mean magnitude for the MCE is less than 6.4 and equal to or greater than 6.0, (N1)60 is greater than 15, and FaSs is between 0.25 and 0.375.

If the mean magnitude shown in Figures 7.6.1-1 to 7.6.1-4 is greater than or equal to 6.4, or if the above requirements are not met for magnitudes between 6.0 and 6.4, or if for the Expected Earthquake, FaSs is greater than 0.375, evaluations of liquefaction and associated phenomena such as lateral flow, lateral spreading, and dynamic settlement shall be evaluated in accordance with these Specifications. 7.6.2 Evaluation of Liquefaction Potential

Procedures given in Appendix D shall be used to evaluate the potential for liquefaction.

7.6.3 Evaluation of the Effects of Liquefaction and Lateral Ground Movement

Procedures given in Appendix D shall be used to evaluate the potential for and effects of liquefaction and liquefaction-related permanent ground movement (i.e., lateral spreading, lateral flow, and dynamic settlement). If both liquefaction and ground movement occur, they shall be treated as separate and independent load cases, unless agreed to or directed otherwise by the owner.

7.6.4 Design Requirements if Liquefaction and Ground Movement Occurs

If it is determined from Appendix D that liquefaction can occur at a bridge site, then one or more of the following approaches shall be implemented in the design.

If liquefaction and no lateral flow occurs, then the bridge shall be designed by conventional procedures including the following requirements: a. Piled Foundations, Drilled Shafts and Pile

Bents: The pile or shaft shall penetrate beyond the bottom of the liquefied layer by at least 3 pile diameters or to a depth that is not affected by liquefaction of the overlying layer or by partial build-up in pore-water pressure, whichever is deeper. In addition the shear reinforcement in a concrete or pre-stressed concrete pile shall meet the requirements of Sec 7.8.2.3 from the pile or bent cap to a depth of 3 diameters below the lowest liquefiable layer.

b. Spread Footings: The bottom of the spread footing shall either be below the liquefiable layer or it shall be at least twice the minimum foundation width of the footing above the liquefiable layer. If liquefaction occurs beneath the base of the footing, the magnitude of settlement caused by liquefaction shall be estimated, and its effects on bridge performance assessed.



If lateral flow or lateral spreading is predicted to occur, the following options shall be considered as detailed in Appendix D. 1. Design the piles or spread footings to resist the

forces generated by the lateral spreading. 2. If the structure cannot be designed to resist the

forces, assess whether the structure is able to tolerate the anticipated movements and meet the geometric and structural constraints of Table C3.2-1. The maximum plastic rotation

of the piles shall be as defined in Article 7.7.9 and 7.8.6.

3. If the structure cannot meet the performance requirements of Table 3.2-1, assess the costs and benefits of various mitigation measures to minimize the movements to a tolerable level to meet the desired performance objective. If a higher performance is desired so that the spread footings or piles will not have to be replaced, the allowable plastic rotations of Articles 7.7.9.2 and 7.8.6.2 shall be met.



Figure 7.6.1-1 Mean Earthquake Magnitude Map for Western United States



Figure 7.6.1-2 Mean Earthquake Magnitude Map for Central and Eastern United States



Figure 7.6.1-3 Mean Earthquake Magnitude Map for Northwest Alaska



Figure 7.6.1-4 Mean Earthquake Magnitude Map for Southeast Alaska



7.6.5 Detailed Foundation Design Requirements

Article 7.4 contains detailed design requirements for each of the different foundation types.

7.6.6 Other Collateral Hazards

The potential occurrence of collateral hazards resulting from fault rupture, landsliding, differential ground compaction, and flooding and inundation shall be evaluated. Procedures for making these evaluations are summarized in Appendix D.



8.4 FOUNDATION DESIGN REQUIREMENTS

8.4.1 Foundation Investigation

8.4.1.1 General

A subsurface investigation, including borings and laboratory soil tests, shall be conducted in accordance with the provisions of Appendix B to provide pertinent and sufficient information for the determination of the Site Class of Article 3.4.2.1. The type and cost of foundations should be considered in the economic, environmental, and aesthetic studies for location and bridge type selection.

8.4.1.2 Subsurface Investigation

Subsurface explorations shall be made at pier and abutment locations, sufficient in number and depth, to establish a reliable longitudinal and transverse substrata profile. Samples of material encountered shall be taken and preserved for future reference and/or testing. Boring logs shall be prepared in detail sufficient to locate material strata, results of penetration tests, groundwater, any artesian action, and where samples were taken. Special attention shall be paid to the detection of narrow, soft seams that may be located at stratum boundaries.

8.4.1.3 Laboratory Testing

Laboratory tests shall be performed to determine the strength, deformation, and flow characteristics of soils and/or rocks and their suitability for the foundation selected. In areas of higher seismicity (e.g., where SDR 4, 5, and 6 apply), it may be appropriate to conduct special dynamic or cyclic tests to establish the liquefaction potential or stiffness and material

damping properties of the soil at some sites, if unusual soils exist or if the foundation is supporting a critical bridge.

8.4.2 Spread Footings

The design of spread footing foundations located in SDR 4, 5, and 6 shall be based on column moments and shears developed using capacity design principles as described in Section 4.8.

Foundation flexibility (Article 5.3.4) shall be modeled for Soil Types C, D, and E if foundation flexibility results in more than a 20% change in response (see Article C5.3.4). For Soil Types A and B, soil flexibility does not need to be considered because of the stiffness of the soil or rock. The potential for and effects of liquefaction and dynamic settlement shall also be determined for spread footing foundations subject to SDR 4 and above. Normally, spread footings shall not be located at SDR 4, 5, and 6 sites where liquefaction is predicted to occur, unless: • the foundation is located below the liquefiable

layer.

• it can be demonstrated by special studies that liquefaction and its effects are very limited, or

• the ground will be improved such that liquefaction will not occur.

Owner approval shall be obtained before proceeding with a spread footing design at a site where liquefaction is predicted to occur.

8.4.2.1 Spring Constants for Footing (Nonliquefiable Sites)

When required to represent foundation flexibility, spring constants shall be developed for spread footing using equations given in Tables 8.4.2.1-1 and 8.4.2.1-2. Alternative procedures given in the FEMA 273 Guidelines for the Seismic Rehabilitation of Buildings



(ATC/BSSC, 1997) are also suitable for estimating spring constants. These computational methods are appropriate for sites that do not liquefy or lose strength during earthquake loading. See Article 8.4.2.3 for sites that are predicted to liquefy.

The shear modulus (G) used to compute the stiffness values in Table 8.4.2.1-1 shall be determined by adjusting the low-strain shear modulus (Gmax) for the level of shearing strain using the following strain adjustment factors, unless other methods are approved by the owner. For FvS1 ≤ 0.40:

• G/Gmax = 0.50 for Expected Earthquake ground motions

• G/Gmax = 0.25 for Maximum Considered Earthquake (MCE) ground motions

For FvS1 > 0.40: • G/Gmax = 0.25 for Expected Earthquake

ground motions • G/Gmax = 0.10 for MCE ground motions

Uplift shall be allowed for footings subject to SDR 4, 5, and 6. The following area adjustment factors (Ra) shall be applied to the equivalent area to account for geometric

nonlinearity introduced by uplift, unless the Owner approves otherwise. For FvS1 ≤ 0.40: • Ra = 1.0 for Expected Earthquake ground

motions

• Ra = 0.75 for MCE ground motions

For FvS1 > 0.40: • Ra = 0.75 for Expected Earthquake ground

motions

• Ra = 0.5 for MCE ground motions

Values of Gmax shall be determined by seismic methods (e.g., crosshole, downhole, or SASW), by laboratory testing methods (e.g., resonant column with adjustments for time), or by empirical equations (Kramer, 1996). The uncertainty in determination of Gmax shall be considered when establishing strain adjustment factors.

No special computations are required to determine the geometric or radiation damping of the foundation system. Five percent system damping shall be used for design, unless special studies are performed and approved by the owner.



Table 8.4.2.1-1 Surface Stiffnesses for a Rigid Plate on a Semi-Infinite Homogeneous Elastic

Half-Space (Adapted from Gazetas, 1991)1

Stiffness Parameter Rigid Plate Stiffness at Surface, Ki'

Vertical Translation, Kz'

( )ν

⎡ ⎤+⎢ ⎥− ⎣ ⎦

0.75

0.73 1.541GL B

L

Horizontal Translation, Ky' (toward long side) ( )ν

⎡ ⎤+⎢ ⎥− ⎣ ⎦

0.85

2 2.52GL B

L

Horizontal Translation, Kx' (toward short side) ( )ν ν

⎡ ⎤⎡ ⎤ ⎛ ⎞+ − −⎢ ⎥⎜ ⎟⎢ ⎥− − ⎝ ⎠⎣ ⎦ ⎣ ⎦

0.85

2 2.5 0.1 12 0.75GL GL BB

L L

Rotation, Kθx' (about x axis) ν

⎛ ⎞ ⎛ ⎞+⎜ ⎟ ⎜ ⎟− ⎝ ⎠ ⎝ ⎠

0.250.75 2.4 0.5

1 XG L BI

B L

Rotation, Kθy' (about y axis) ν

⎡ ⎤⎛ ⎞⎢ ⎥⎜ ⎟− ⎝ ⎠⎢ ⎥⎣ ⎦

0.150.75 3

1 YG LI

B

Table note: 1. See Figure 8.4.2.1-1** for definitions of terms

Table 8.4.2.1-2 Stiffness Embedment Factors for a Rigid Plate on a Semi-Infinite Homogeneous

Elastic Half-Space (Adapted from Gazetas, 1991)1

Stiffness Parameter Embedment Factors, ei

Vertical Translation, ez

( )⎡ ⎤⎛ ⎞+⎡ ⎤⎛ ⎞ ⎢ ⎥+ + + ⎜ ⎟⎢ ⎥⎜ ⎟ ⎜ ⎟⎢ ⎥⎝ ⎠⎣ ⎦ ⎝ ⎠⎣ ⎦

0.672 2

1 0.095 1 1.3 1 0.2L BD B d

B L LB

Horizontal Translation, ey (toward long side)

( )⎧ ⎫⎡ ⎤⎛ ⎞− +⎪ ⎪⎜ ⎟⎢ ⎥⎡ ⎤⎛ ⎞ ⎪ ⎪⎝ ⎠⎢ ⎥+ +⎢ ⎥ ⎨ ⎬⎜ ⎟ ⎢ ⎥⎝ ⎠⎢ ⎥ ⎪ ⎪⎣ ⎦ ⎢ ⎥⎪ ⎪⎣ ⎦⎩ ⎭

0.4

0.5

2

162 21 0.15 1 0.52

dD L B dDB B L

Horizontal Translation, ex (toward short side)

( )⎧ ⎫⎡ ⎤⎛ ⎞− +⎪ ⎪⎜ ⎟⎢ ⎥⎡ ⎤⎛ ⎞ ⎪ ⎪⎝ ⎠⎢ ⎥+ +⎢ ⎥ ⎨ ⎬⎜ ⎟ ⎢ ⎥⎝ ⎠⎢ ⎥ ⎪ ⎪⎣ ⎦ ⎢ ⎥⎪ ⎪⎣ ⎦⎩ ⎭

0.4

0.5

2

162 21 0.15 1 0.52

dD L B dDL L B

Rotation, eθx (about x axis)

−⎛ ⎞⎛ ⎞ ⎛ ⎞+ +⎜ ⎟⎜ ⎟ ⎜ ⎟⎜ ⎟⎝ ⎠ ⎝ ⎠⎝ ⎠

0.20 0.5021 2.52 1d d d BB B D L

Rotation, eθy (about y axis)

−⎛ ⎞⎛ ⎞ ⎛ ⎞ ⎛ ⎞+ +⎜ ⎟⎜ ⎟ ⎜ ⎟ ⎜ ⎟⎜ ⎟⎝ ⎠ ⎝ ⎠ ⎝ ⎠⎝ ⎠

0.60 1.9 0.602 21 0.92 1.5d d dL L D

Table note: Embedment factors multiplied by spring



y

x

y

x

z

z

Plan

Section

L (length)

B (width)

d (thickness)

D (depth)

Homogeneous Soil Properties G (shearing modulus)

ν ( Poisson's ratio)

Figure 8.4.2.1-1 Properties of a Rigid Plate on a Semi-Infinite Homogeneous Elastic Half-Space

for Stiffness Calculations

8.4.2.2 Moment-Rotation and Shear-Displacement Relationships for Footing (Nonliquefiable Sites)

The moment and shear capacity of the foundation shall be confirmed for design loads given in Article 4.8. Moment-rotation and shear force-displacement relationships shall be developed as required by Article 5.3.4. Unless approved otherwise by the owner, the moment-rotation curve for SDAP E shall be represented by a bilinear, moment-rotation curve. The initial slope of the bi-linear curve shall be defined by the rotational spring constant given in Article 8.4.2.1.

The maximum resisting force (i.e., plastic capacity) on the force-deformation curve shall be defined for the best-estimate case. The footing liftoff shall be no more than 50% of the footing area at peak displacement during the push-over analysis, unless special studies are performed and approved by the owner. A bilinear force displacement relationship shall

also be developed for the shear component of resistance.

This approach shall not be used at sites that will liquefy during seismic loading. See Article 8.4.2.3 for sites that liquefy.

8.4.2.3 Liquefaction and Dynamic

Settlement

An evaluation of the potential for liquefaction within near-surface soil shall be made in accordance with requirements given in Article 8.6 and Appendix D of these specifications. If liquefaction is predicted to occur under the design ground motion, spread footings foundations shall not be used unless

• the footing is located below the liquefiable layer,

• ground improvement is performed to mitigate the occurrence of liquefaction, or



• special studies are conducted to demonstrate that the occurrence of liquefaction will not be detrimental to the performance of the bridge support system.

The owner’s approval shall be obtained before initiating ground improvement or special studies.

8.4.3 Driven Piles

8.4.3.1 General

Resistance factors for pile capacities shall be as specified in Table 10.5.4-2 of the AASHTO LRFD provisions, with the exception that resistance factors of 1.0 shall be used for seismic loads.

For the effect of settling ground and downdrag loads, unfactored load and resistance factors (γ = 1.0; φ = 1.0) shall be used, unless required otherwise by the owner.

Batter piles shall not be used where downdrag loads are expected unless special studies are performed.

For seismic loading the groundwater table location shall be the average groundwater location, unless the owner approves otherwise.

8.4.3.2 Design Requirements

The design of driven pile foundations shall be based column loads determined by capacity design principles (Article 4.8) or elastic seismic forces, whichever is smaller. Both the structural and geotechnical elements of the foundation shall be designed for the capacity design forces of Article 4.8.

Foundation flexibility (Article 5.3.4) shall be incorporated into design for Soil Profile Types

C, D, and E, if the effects of foundation flexibility contribute more than 20% to the displacement of the system. For SDAP E foundations flexibility shall be included in the push-over analysis whenever it is included in the dynamic analysis.

Liquefaction shall be considered when applicable during the development of spring constants and capacity values for these seismic design and analysis procedures.

8.4.3.3 Axial and Rocking Stiffness for Driven Pile/Pile Cap Foundations (Nonliquefiable Sites)

The axial stiffness of the driven pile foundations shall be determined for design cases in which foundation flexibility is included. For many applications, the axial stiffness of a group of piles can be estimated within sufficient accuracy using the following equation: Ksv = Σ 1.25AE/L (8.4.3.2-1)

where

A = cross-sectional area of the pile E = modulus of elasticity of the piles L = length of the piles N = number of piles in group and is

represented by the summation symbol in the above equations.

The rocking spring stiffness values about each horizontal pile cap axis can be computed assuming each axial pile spring acts as a discrete Winkler spring. The rotational spring constant (i.e., moment per unit rotation) is then given by Ksrv = Σ kvn Sn

2 (8.4.3.2-2)

where



kvn = axial stiffness of the nth pile Sn = distance between the nth pile and

the axis of rotation

The effects of group action on the determination of stiffness shall be considered if the center-to-center spacing of piles for the group in the direction of loading is closer than 3 pile diameters.

8.4.3.4 Lateral Stiffness Parameters for Driven Pile/Pile Cap Foundations (Nonliquefiable Sites)

The lateral stiffness parameters of driven pile foundations shall be estimated for design cases in which foundation flexibility is included. Lateral response of a pile foundation system depends on the stiffness of the piles and, very often, the stiffness of the pile cap. Procedures for defining the stiffness of the pile component of the foundation system are covered in this article. Methods for introducing the pile cap stiffness are addressed in Article 8.4.3.5.

For preliminary analyses involving an estimate of the elastic displacements of the bridge, pile stiffness values can be obtained by using a series of charts prepared by Lam and Martin (1986). These charts are reproduced in Figures 8.4.3.4-1 through 8.4.3.4-6. The charts are applicable for mildly nonlinear response, where the elastic response of the pile dominates the nonlinear soil stiffness.

For push-over analyses the lateral load displacement relationship must be extended into the nonlinear range of response. It is usually necessary to use computer methods to develop the load-displacement relationship in

this range, as both the nonlinearity of the pile and the soil must be considered. Programs such as LPILE (Reese and Wang, 1997), COM 624 (Wang and Reese, 1991), and FLPIER (Hoit and McVay, 1996) are used for this purpose. These programs use nonlinear "p-y" curves to represent the load-displacement response of the soil; they also can accommodate different types of pile-head fixity. Procedures for determining the "p-y" curves are discussed by Lam and

Martin (1986) and more recently by Reese et al. (1997).

The effects of group action on lateral stiffness shall be considered if the center-to-center spacing of the piles is closer than 3 pile diameters.

8.4.3.5 Pile Cap Stiffness and Capacity

The stiffness and capacity of the pile cap shall be considered in the design of the pile foundation. The pile cap provides horizontal resistance to the shear loading in the column. Procedures for evaluating the stiffness and the capacity of the footing in shear shall follow procedures given in Article C8.4.2.2 for spread footings, except that the base shear resistance of the cap shall be neglected.

When considering a system comprised of a pile and pile cap, the stiffness of each shall be considered as two springs in parallel. The composite spring shall be developed by adding the reaction for each spring at equal displacements.



Figure 8.4.3.4-1 Recommendations for Coefficient of Variation in

Subgrade Modulus with Depth for Sand (ATC, 1996)



Figure 8.4.3.4-2 Recommendations for Coefficient of Variation in Subgrade Modulus with Depth for Clay (ATC, 1996)



Figure 8.4.3.4-3 Coefficient of Lateral Pile Head Stiffness for Free-Head Pile Lateral Stiffness

(ATC, 1996)



Figure 8.4.3.4-4 Coefficient for Lateral Pile-Head Stiffness for Fixed-Head Pile Lateral Stiffness

(ATC, 1996)



Figure 8.4.3.4-5 Coefficient for Pile Head Rotation (ATC, 1996)



Figure 8.4.3.4-6 Coefficient for Cross-Coupling Stiffness Term (ATC, 1996)



8.4.3.6 Moment and Shear Design (Nonliquefiable Sites)

The capacity of the structural elements of driven pile foundations shall be designed to resist the capacity design forces of Article 4.8 or the elastic design force within the column, whichever is smaller. Unfactored resistance (φ = 1.0) shall be used in performing the geotechnical capacity check. The load on the leading row of piles during overturning shall not exceed the plunging capacity of the piles. Separation between the pile tip and the soil (i.e. gapping) shall be allowed only in the most distant row of trailing piles. Forces on all other rows of piles shall either be compressive or not exceed the nominal tension capacity of the piles. The maximum shear force on the pile(s) shall be less than the structural shear capacity of the piles.

If the plunging capacity is exceeded or gapping of other than the trailing row of piles occurs, special studies shall be conducted to show that performance of the pile system is acceptable. Special studies shall be performed only with the prior consent of the owner and require SDAP E.

8.4.3.7 Liquefaction and Dynamic Settlement Evaluations

If liquefaction is predicted to occur at the site, effects of liquefaction on the bridge foundation shall be evaluated. This evaluation shall consider the potential for loss in lateral bearing support, flow and lateral spreading of the soil, settlement below the toe of the pile, and settlement from drag loads on the pile as excess porewater pressures in liquefied soil dissipate. Procedures given in Appendix D shall be followed when making these evaluations.

If liquefaction causes unacceptable bridge performance, consideration should be given to

the use of ground improvement methods to meet design requirements. In light of the potential costs of ground improvement, the owner shall be consulted before proceeding with a design for ground improvement to review the risks associated with liquefaction relative to the costs for remediating the liquefaction potential.

8.4.4 Drilled Shafts

Procedures identified in Article 8.4.3, including those for liquefaction and dynamic settlement, generally apply with the exceptions that, (1) the ultimate capacity of single shaft foundations in compression and uplift shall not be exceeded under maximum seismic loads and (2) the flexibility of the drilled shaft shall be represented using either the estimated depth of fixity or soil springs in a lateral pile analysis.

Checks shall be conducted to confirm that minimum shaft lengths occur. The stable length can be determined by conducting nonlinear computer modeling or by using a length (L) > πλ where λ = [EIp/Es]0.25 for cohesive soils, and

λ = [EIp/f] 0.20 for cohesionless soils

where E = Young’s modulus of the shaft

Ip = moment of inertia of the shaft

F = coefficient of variation of subgrade modulus

Es = subgrade modulus of soil

Z = embedded depth of the shaft

The nonlinear properties of the shaft shall be considered in evaluating the lateral response of the pile to lateral loads during a seismic event. Diameter adjustments shall be



considered during lateral analyses of shafts with a diameter greater than 600 mm if the shaft is free to rotate, as in the case of a column extension (i.e., no pile cap). Contributions from base shear shall also be considered.



8.6 LIQUEFACTION DESIGN REQUIREMENTS

8.6.1 General

An evaluation of the potential for and consequences of liquefaction within near-surface soil shall be made in accordance with the following requirements: A liquefaction assessment is required unless one of the following conditions is met or as directed otherwise by the owner. • Mean magnitude for the MCE event is less

than 6.0 (Figures 8.6.1-1 to 8.6.1-4); • Mean magnitude of the MCE event is less than

6.4 and equal to or greater than 6.0, and the normalized Standard Penetration Test (SPT) blow count [(N1)60] is greater than 20;

• Mean magnitude for the MCE event is less than 6.4 and equal to or greater than 6.0, (N1)60 is greater than 15, and FaSs is between 0.25 and 0.375.

If the mean magnitude shown in Figures 8.6.1-1 to 8.6.1-4 is greater than or equal to 6.4, or if the above requirements are not met for magnitudes between 6.0 and 6.4 or if for the Expected Earthquake, FaSs is greater than 0.375, evaluations of liquefaction and associated phenomena such as lateral flow, lateral spreading, and dynamic settlement shall be evaluated in accordance with these Specifications.

8.6.2 Evaluation of Liquefaction Potential

Procedures given in Appendix D shall be used to evaluate the potential for liquefaction.

8.6.3 Evaluation of the Effects of Liquefaction and Lateral Ground Movement

Procedures given in Appendix D shall be used to evaluate the potential for and effects of liquefaction and liquefaction-related permanent ground movement (i.e., lateral spreading, lateral flow, and dynamic settlement). If both liquefaction and ground movement occur, they shall be treated as separate and independent load cases, unless agreed to or directed otherwise by the owner.

8.6.4 Design Requirements if Liquefaction and Ground Movement Occurs

If it is determined from Appendix D that liquefaction can occur at a bridge site, then one or more of the following approaches shall be implemented in the design.

Bridges shall be supported on deep foundations unless (1) the footing is located below the liquefiable layer, (2) special design studies are conducted to demonstrate that the footing will tolerate liquefaction, or (3) the ground is improved so that liquefaction does not occur. If spread footings are being considered for use at a liquefiable site, owner approval shall be obtained before beginning the design process.

If liquefaction occurs, then the bridge shall be designed and analyzed in two configurations as follows: 1. Nonliquefied Configuration: The structure

shall be analyzed and designed, assuming no liquefaction occurs using the ground response spectrum appropriate for the site soil conditions.



Figure 8.6.1-1 Mean Earthquake Magnitude Map for Western United States



Figure 8.6.1-2 Mean Earthquake Magnitude Map for Eastern United States



Figure 8.6.1-3 Mean Earthquake Magnitude Map for Alaska



Figure 8.6.1-4 Mean Earthquake Magnitude Map for Southeast Alaska



2. Liquefied Configuration: The structure as designed in Nonliquefied Configuration above shall be reanalyzed and redesigned, if necessary, assuming that the layer has liquefied and the liquefied soil provides whatever residual resistance is appropriate (i.e., “p-y curves” or modulus of subgrade reaction values for lateral pile response analyses consistent with liquefied soil conditions). The design spectra shall be the same as that used in Nonliquefied Configuration unless a site-specific response spectra has been developed using nonlinear, effective stress methods (e.g., computer program DESRA or equivalent) that properly account for the buildup in pore-water pressure and stiffness degradation in liquefiable layers. The reduced response spectra resulting from the site-specific nonlinear, effective stress analyses shall not be less than 2/3’s of that used in Nonliquefied Configuration. The Designer shall provide a drawing of the load path and energy dissipation mechanisms in this condition as required by Article 3.3 since it is likely that plastic hinges will occur in different locations than for the non-liquefied case. Shear reinforcement given in Article 8.8.2.3 shall be used in all concrete and prestressed concrete piles to a depth of 3 pile diameters below the liquefied layer.

If lateral flow or lateral spreading occurs, the following options shall be considered. 1. Design the piles to resist the forces

generated by the lateral spreading.

2. If the structure cannot be designed to resist the forces, assess whether the structure is able to tolerate the anticipated movements and meet the geometric and structural constraints of Table C3.2-1. The maximum plastic rotation of the piles is 0.05 radians as per Article 8.7.9 and 8.8.6.

3. If the structure cannot meet the performance requirements of Table 3.2-1, assess the costs and benefits of various mitigation measures to minimize the movements to a tolerable level to meet

the desired performance objective. If a higher performance is desired so that the piles will not have to be replaced, the allowable plastic rotations in-ground hinges of Article 8.7.9.2 and 8.8.6.2 shall be met.

8.6.5 Detailed Foundation Design Requirements

Article 8.4 contains detailed design requirements for each of the different foundation types.

8.6.6 Other Collateral Hazards

The potential occurrence of collateral hazards resulting from fault rupture, landsliding, differential ground compaction, and flooding and inundation shall be evaluated. Procedures for making these evaluations are summarized in Appendix D.

1325 NCHRP 20-7(193) Task 6 Report.doc 5B-1

TASK 5

APPENDIX 5B

ASSESSMENT AND MITIGATION OF LIQUEFACTION HAZARDS TO BRIDGE

APPROACH EMBANKMENTS IN OREGON

1325 NCHRP 20-7(193) Task 6 Report.doc

ASSESSMENT AND MITIGATION OF

LIQUEFACTION HAZARDS TO

BRIDGE APPROACH EMBANKMENTS

IN OREGON

Final Report

SPR 361

by

Dr. Stephen E. Dickenson Associate Professor

and Nason J. McCullough

Mark G. Barkau Bryan J. Wavra

Graduate Research Assistants Dept. of Civil Construction and Environmental Engineering

Oregon State University Corvallis, OR 97331

for

Oregon Department of Transportation

Research Group 200 Hawthorne Ave. SE Salem, OR 97301-5192

And

Federal Highway Administration

Washington, D.C. 20590

November 2002