dg04 - extended end-plate moment connections

Steel Design Guide Series

Extended End-PlateMoment Connections

Extended End-PlateMoment Connections

Design Guide for Extended End-Plate Moment ConnectionsThomas M. Murray, PhD, RE.Montague-Betts Professor of Structural Steel DesignVirginia Polytechnic Institute and State UniversityBlacksburg, Virginia

A M E R I C A N I N S T I T U T E O F S T E E L C O N S T R U C T I O N

Steel Design Guide Series

© 2003 by American Institute of Steel Construction, Inc. All rights reserved.This publication or any part thereof must not be reproduced in any form without permission of the publisher.

Copyright 1990

by

American Institute of Steel Construction, Inc.

All rights reserved. This book or any part thereofmust not be reproduced in any form without the

written permission of the publisher.

The information presented in this publication has been prepared in accordance with rec-ognized engineering principles and is for general information only. While it is believedto be accurate, this information should not be used or relied upon for any specific appli-cation without competent professional examination and verification of its accuracy,suitablility, and applicability by a licensed professional engineer, designer, or architect.The publication of the material contained herein is not intended as a representationor warranty on the part of the American Institute of Steel Construction or of any otherperson named herein, that this information is suitable for any general or particular useor of freedom from infringement of any patent or patents. Anyone making use of thisinformation assumes all liability arising from such use.

Caution must be exercised when relying upon other specifications and codes developedby other bodies and incorporated by reference herein since such material may be mod-ified or amended from time to time subsequent to the printing of this edition. TheInstitute bears no responsibility for such material other than to refer to it and incorporateit by reference at the time of the initial publication of this edition.

Printed in the United States of America

Second Printing: October 2003


TABLE OF CONTENTS

1. I N T R O D U C T I O N . . . . . . . . . . . . . . . . . . . . . . . . . . .1.1 Background . . . . . . . . . . . . . . . . . . . . . . . . . . . . . .1.2 Overview of Design Guide . . . . . . . . . . . . . . . . . .1.3 Brief Literature O v e r v i e w . . . . . . . . . . . . . . . . . . .

2. RECOMMENDED DESIGN PROCEDURES . . .2.1 Basis of Design Recommendations . . . . . . . . . . .2.2 Limit States Check L i s t . . . . . . . . . . . . . . . . . . . .

3. UNSTIFFENED, EXTENDED END-PLATECONNECTION D E S I G N . . . . . . . . . . . . . . . . . . . . .3.1 The Four-Bolt Configuration Design

Procedures and E x a m p l e s . . . . . . . . . . . . . . . . . . .3.1.1 Design Procedures . . . . . . . . . . . . . . . . . . . .3.1.2 Allowable Stress Design Examples . . . . . . .3.1.3 Load and Resistance Design Example . . . .

3.2 Eight-Bolt Design Procedures and AllowableStress Design Example . . . . . . . . . . . . . . . . . . . . .

4. STIFFENED, EXTENDED END-PLATECONNECTION DESIGN . . . . . . . . . . . . . . . . . . . .

4.1 Design P r o c e d u r e s . . . . . . . . . . . . . . . . . . . . . . . . .

4.2 Design Examples . . . . . . . . . . . . . . . . . . . . . . . . .4.2.1 Allowable Stress Design Examples . . . . . .4.2.2 Load and Resistance Factor Design

Examples . . . . . . . . . . . . . . . . . . . . . . . . . . .

BIBLIOGRAPHY

APPENDIX A—ASD NOMENCLATURE,DESIGN AIDS AND QUICK REFERENCEEXAMPLES . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . .A.1 ASD Nomenclature . . . . . . . . . . . . . . . . . . . . . . .A.2 ASD Design Aids . . . . . . . . . . . . . . . . . . . . . . . .A.3 ASD Quick Reference Examples . . . . . . . . . . . .

APPENDIX B—LRFD NOMENCLATURE,DESIGN AIDS AND QUICK REFERENCEEXAMPLES . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . .B.1 LRFD Nomenclature . . . . . . . . . . . . . . . . . . . . . .B.2 LRFD Design Aids . . . . . . . . . . . . . . . . . . . . . . .B.3 LRFD Quick Reference Examples . . . . . . . . . . .

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PREFACE

This booklet was prepared under the direction of the Com-mittee on Research of the American Institute of Steel Con-struction, Inc. as part of a series of publications on specialtopics related to fabricated structural steel. Its purpose is toserve as a supplemental reference to the AISC Manual ofSteel Construction to assist practicing engineers engaged inbuilding design.

The design guidelines suggested by the authors that are out-side the scope of the AISC Specifications or Code do notrepresent an official position of the Institute and are not in-tended to exclude other design methods and procedures. Itis recognized that the design of structures is within the scopeof expertise of a competent licensed structural engineer, ar-chitect or other licensed professional for the application ofprinciples to a particular structure.

The sponsorship of this publication by the American Ironand Steel Institute is gratefully acknowledged.

The information presented in this publication has been prepared in accordance with recognized engineer-ing principles and is for general information only. While it is believed to be accurate, this information shouldnot be used or relied upon for any specific application without competent professional examination and verifi-cation of its accuracy, suitability, and applicability by a licensed professional engineer, designer or archi-tect. The publication of the material contained herein is not intended as a representation or warranty onthe part of the American Institute of Steel Construction, Inc. or the American Iron and Steel Institute, orof any other person named herein, that this information is suitable for any general or particular use or offreedom infringement of any patent or patents. Anyone making use of this information assumes all liabilityarising from such use.


Chapter 1INTRODUCTION

1.1 BACKGROUND

The use of moment end-plate connections in multi-story,moment resistant frame construction is becoming more com-mon because of advancements in design methods and fabri-cation techniques, both of which have resulted in decreasedcosts. A typical moment end-plate connection is composedof a steel plate welded to the end of a beam section withattachment to an adjacent member using rows of fully ten-sioned high-strength bolts. The connection may be betweentwo beams (splice plate connection) or between a beam anda column. End-plate moment connections are classified aseither flush or extended with or without stiffeners and fur-ther classified depending on the number of bolts at the ten-sion flange. A flush connection is detailed such that the end-plate does not appreciably extend beyond the beam flangesand all bolts are located between the beam flanges. Anextended end-plate is one which extends beyond the tensionflange a sufficient distance to allow the location of bolts otherthan between the beam flanges. Extended end-plates maybe used with or without a stiffener between the end-plateand the beam flange in the plane of the beam web. Flushend-plate connections are typically used in frames subjectto light lateral loadings or near inflection points of gableframes. Extended end-plates are used for beam-to-columnmoment connections. Only extended end-plates are consid-ered in this design guide.

Four extended end-plate configurations are shown in Fig.1.1. The four-bolt unstiffened configuration shown in Fig.1.1(a) is probably the most commonly used in multi-storyframe construction. An allowable stress design (ASD) pro-cedure for this connection is found in the 8th and 9th edi-tions, American Institute of Steel Construction (AISC) Man-ual of Steel Construction (1980, 1989a) and a load andresistance factor design (LRFD) procedure is found in theAISC Load and Resistance Factor Design Manual of SteelConstruction (1986a). Assuming the full beam momentcapacity is to be resisted, A325 bolts and a maximum boltdiameter of 1½ in. (maximum practical size because of tight-ening considerations), this connection is limited because ofbolt capacity to use with less than one-half of the availablebeam sections. The connection strength can be increased byadding a stiffener, Fig. 1.1(b), or increasing the number ofbolts per row to four, Fig. 1.1(c). Formal design proceduresare not available for the former, and the latter requires a widecolumn flange. The stiffened A325 eight-bolt connectionshown in Fig. 1.1(d) is capable of developing the full momentcapacity of most of the available beam sections even if boltdiameter is limited to 1½ in. Design procedures for this con-

figuration are found in the 9th edition AISC Manual of SteelConstruction (1989a).

As with any connection, end-plate connections have cer-tain advantages and disadvantages. The principal advantagesare:

(a) The connection is suitable for winter erection in thatonly field bolting is required.

(b) All welding is done in the shop, eliminating field weld-ing associated problems.

(c) Without the need for field welding, the erection pro-cess is relatively fast.

(d) If fabrication is accurate, it is easy to maintain plumb-ness of the frame.

(e) Lower total installed cost for many cases.The principal disadvantages are:

(a) The fabrication techniques are somewhat more strin-gent because of the need for accurate beam length and"squareness" of the beam end.

(b) Column out-of-squareness can cause erection difficul-ties but can be controlled by fabricating the beams ¼in. to in. short and providing "finger" shims.

(c) End plates often warp due to the heat of welding.

Fig. 1.1. Extended end-plate configurations.

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(d) End-plates are subject to lamellar tearing in the regionof the top flange tension weld.

(e) The bolts are in tension, which can result in pryingforces.

A number of designers and fabricators in the United Stateshave successfully used moment end-plate connections forbuilding frames up to 30 stories in height. It is believed that,in spite of the several disadvantages, moment end-plate con-nections can provide economic solutions for rigid frame con-struction. Because very little research has been conductedon the low cycle fatigue strength of end-plate connections,their use is not presently recommended in areas of high seis-mic activity.

1.2 OVERVIEW OF DESIGN GUIDE

The intent of this guide is to present complete design proce-dures and examples for extended moment end-plate connec-tions suitable for fully restrained (or continuous frame) con-struction. Chapter 2 presents the basic design procedures forthe end-plate configurations shown in Figs, 1.1(a), (c) and(d). Chapter 3 contains ASD and LRFD design examplesfor the four-bolt unstiffened configuration shown in Fig. 1.1(a) and the eight-bolt unstiffened configuration shown in Fig.1.1(c). Chapter 4 contains ASD and LRFD examples for theeight-bolt stiffened configuration shown in Fig. 1.1 (d).Appendix A includes allowable stress design (ASD) nomen-clature, several design aids and quick reference examples.Appendix B is similar to Appendix A except it is for loadand resistance factor design (LRFD). The quick referenceexamples serve as a guide for designers who are thoroughlyfamiliar with moment end-plate design. The following sec-tion is a brief review of available literature for backgroundpurposes.

1.3 BRIEF LITERATURE OVERVIEW

End Plate Design. Research starting in the early 1950s andcontinuing to the present has resulted in refined design pro-cedures for both flush and extended end-plate connections.The earlier design methods were based on statics and sim-ple assumptions concerning prying forces. These methodsresulted in thick end-plates and large diameter bolts. Otherstudies have been based on yield-line theory. The more recentstudies have used the finite element method and regressionanalysis to develop design equations. Accurate solutions canbe developed using the latter technique; however, the pro-cedure is time consuming and the resulting design equationsusually involve terms to odd powers which virtuallyeliminates "structural feel" from the design.

Early attempts (prior to about 1975) to develop designcriteria for moment end-plate connections were based on the

"tee-stub" analogy. All of these methods resulted in designprocedures which predict a high degree of "prying action"resulting in large end-plate thicknesses and large boltdiameters. One such method for the four-bolt, extended con-figuration (Fig. 1.1(a)) is found in the 7th edition of the AISCManual of Steel Construction (1969).

More recently, methods based on refined yield-lineanalyses have been suggested. A number of configurationshave been studied in Europe (Zoetermeijer, 1974, 1981;Packer and Morris, 1977; Mann and Morris, 1979) as wellas in the United States (Srouji, 1983; Hendrick et al., 1985;Morrison, 1986). Most of this work has involved flush end-plate configurations.

Finite element methodology for the analysis of end-plateswas first developed by Krishnamurthy (1978, 1981). Hisexhaustive analytical study of four-bolt, unstiffened, extendedend plates (Fig. 1.1(a)), along with a series of experimentalinvestigations, led to the development of a design procedurefirst published in the 8th edition of the AISC Manual of SteelConstruction.

More recently, Ahuja (1982) and Ghassemieh (1983) haveinvestigated the stiffened configuration with two rows of twobolts on each side of the tension flange (Fig 1.1(d)). Theyused regression analysis to develop design equations. Mur-ray and Kukreti (1988) have developed a simplified designprocedure using their regression results which appears in the9th edition AISC Manual of Steel Construction.

Bolt Design. Early end-plate design procedures (Douty andMcGuire, 1965; Nair et al., 1969; Kato and McGuire, 1973)all involved the calculation of bolt prying forces based onvarious assumptions. The assumed location of the pryingforce was at or near the edge of the end-plate. Packer andMorris (1977), Phillips and Packer (1981), Mann and Morris(1979), and Zoetermeijer (1974, 1981) have all included pry-ing action forces in their yield-line based design procedures.The various recommendations range from rather complicatedanalytical procedures to a simple increase in bolt force overthe applied tension (Mann and Morris, 1979).

Krishnamurthy (1978a) argues that even though pryingaction is present, it is overly conservative to assume it to beacting at the edge of the plate as this normally results in thickerthan necessary end-plates. His studies describe prying forceas a pressure bulb which is formed under the bolt head dueto the tensioning of the bolt and shifts towards the edge asthe beam flange force increases. For any given loading, thepressure bulb is located somewhere between the edge of theend plate and the bolt head. He states, for service load con-ditions when the beam flange loads are small, the pressurebulb is closer to the bolt head than to the plate edge, andthe plate moments are much smaller than those predictedby prying force formulas. Consequently, in his design pro-cedure for four-bolt, extended, unstiffened end plates (Fig.1.1(a)), prying forces are ignored, that is, the bolt size is deter-mined directly from the force delivered by the beam flange.

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Kennedy et al. (1981) have presented a method for calcu-lation of prying forces as a function of plate "thickness" rela-tive to applied load. They identified three types of end-platebehavior. The first type is characterized by the absence ofplastic hinges in the end plate. These end-plates are said tobe "thick." Under low loading conditions all end plates fallinto this category. The upper limit of this behavior occursat a load which causes flexural yielding in the end-plate atthe beam flange. Once this load is exceeded, a plastic hingeis formed at the flange and the end-plate is said to be of"intermediate" thickness. As the load is increased, a sec-ond plastic hinge forms at the bolt lines. At this load, theend-plate is considered to be a "thin" plate. Further, theyconsider bolt force to be the sum of a portion of the flangeforce plus prying force and identify three stages of pryingaction corresponding to the three phases of end-platebehavior. For "thick" plates, the prying force is assumedto be zero. When the end plate is considered as "thin," theprying force is at its maximum. For "intermediate" plates,the prying force is somewhere between zero and the maxi-mum value. They suggest that for ideal design, the end-plateshould be "thick" under service loads, "intermediate" underfactored loads and function as a "thin" plate at ultimateloads.

Srouji (1983), Hendrick et al. (1985) and Morrison (1986)have modified the Kennedy et al. (1981) approach for usewith two- and four-bolt flush end plates; four-bolt, stiffenedextended end plates; and extended end plates with multiplebolt rows below the tension flange. Each researcher haspresented experimental evidence to verify the predictionequations.

Ahuja (1982) and Ghassemieh (1983) have presented finiteelement/regression analysis equations to predict bolt forcesabove the pretension level for eight-bolt, stiffened, extendedend plates. Ahuja's results are based on elastic material prop-erties, but Ghassemieh's results include inelastic materialproperties. Both authors limit the use of their results to A36steel and A325 bolts.

Beam-to-End-Plate Weld Design. Griffiths (1984) suggeststhat either full penetration welds or fillet welds sufficientto develop the beam flange in tension be used to connectthe end plate to the beam. This recommendation holds evenif the full capacity of the beam is not being utilized becauseof the large local deformations that occur along the end plate.

Column Side Design. Relative to end-plate research, theamount of effort devoted to the column side of end-platemoment connections is quite limited. Only a few papers havebeen published which suggest design guidelines for the threecolumn side failure modes: column web yielding, columnweb buckling and column flange bending failure.

The critical section for column web yielding is at the toeof the column web fillet. For design of welded connections,the present AISC Manual (1989a) criteria is based on a loadpath which is assumed to vary linearly on a 2½:1 slope from

the beam flange through the column flange and fillet. If thestress at this critical section exceeds the yield stress of thecolumn material, a column web stiffener is required oppo-site the beam tension and compression flanges.

For the case of end-plate moment connections, the widthof the stress pattern at the critical section may be consider-ably wider due to the insertion of the end plate into the loadpath. Hendrick and Murray (1983) conducted a number ofcolumn compression region tests using both stiffened andunstiffened end plates and concluded that the slope of thestress path through the end plate can be taken as 1:1 and thatin the column as 3:1. This recommendation is also found inHendrick and Murray (1984) and in the AISC LRFD man-ual (1986a). Hendrick's recommendations, except for the 3:1slope, are also found in AISC Engineering for Steel Con-struction (1984), where 2½:1 is used.

Newlin and Chen (1971) recommend that an interactionequation be used to check combined web yielding strengthand web buckling. Possibly anticipating resistance to suchform, they also provided a simple check for web buckling.This latter provision was adopted by AISC in their 1978 spec-ification revision.

Witteveen et al. (1982) found three modes of failure forbending of the column flange. The first mode prevails whenthe column flange is thick when compared with bolt diameter.The second failure mode is when the stiffnesses of the boltsand flange are such that prying forces can develop becauseyield lines form in the flange near the fillet, causing boththe flange and the bolts to fail. The third failure mode occurswhen yield lines form in the flange near both the bolts andthe fillet. Design procedures for each failure mode arepresented as well as test results to verify the analytical work.

Mann and Morris (1979) present complete design proce-dures for the column side of end-plate connections. Therecommendations are based primarily on the work of Packerand Morris (1977). However, only the case when the col-umn flange is much less stiff than the end plate is consid-ered. Three possible failure modes were found to exist. Ifthe flange is very stiff, there are no prying forces and thefailure occurs when the bolts rupture. The second failuremode occurs when the column flange is less stiff, whichresults in a combination of bolt fracture and flange yieldingnear the column web. The third failure mode is character-ized by yield lines forming and causing double curvature inthe flange plate. Provisions to estimate the column flangecapacity for each of the failure modes are provided. If thefirst failure mode governs, the total bolt force is equal tothe applied flange force. For the second failure mode, pry-ing forces are accounted for by limiting bolt capacity to 80%of tensile capacity. Mann and Morris do not provide methodsto estimate prying forces if the third failure mode governs.

Granstrom (1980) extended tee-hanger results to includecolumn flanges. The procedure to determine the required col-umn flange thickness is the same as that used for tee-hanger

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flange thickness except that an effective column flange lengthis used. Granstrom does not consider prying action effects.

Hendrick and Murray (1983) conducted a limited seriesof tests to evaluate the methods suggested by Mann andMorris (1979), Granstrom (1980) and Witteveen et al. (1982)for use with North American rolled sections. They concludedthat the method proposed by Mann and Morris (1979) is themost suitable for the evaluation of unstiffened column flangesin the tension region of four-bolt, unstiffened end-plate con-nections. They also modified the Krishnamurthy (1978a) pro-cedure for end plates by introducing an effective columnflange length to obtain the same results as found with theMann and Morris equations. Finally, they developed the"rule of thumb" found in the AISC Engineering for SteelConstruction manual (1984) which states that, under certainlimitations, the column flange is adequate if its thickness isgreater than the required bolt diameter from the Krishnamur-thy end-plate design procedure. All of his work applies onlyto A36 steel.

Curtis (1985) has proposed design rules for column flangestrength in the tension region of eight-bolt, stiffened end-plate connections. His method is based on the Ghassemieh

(1983) end-plate design procedure with an effective columnflange length and is therefore limited to A36 steel.

Curtis and Murray (1989) have modified both the Hen-drick and Murray (1983) and Curtis (1985) recommendationsto ensure adequate column flange stiffness for use in fullyrestrained (continuous) construction.

Procedures for the design of column web stiffeners to pre-vent web yielding or buckling have been suggested by Hen-drick and Murray (1984) and have the same form as forwelded beam-to-column connections in the 1989 AISC ASDSpecification.

Mann and Morris (1979) have presented methods to esti-mate the resistance of column flanges stiffened using stan-dard web stiffeners. Zoetemeijer (1974) and Moore and Sims(1986) have recommended the use of "flange washer platestiffeners." They have also provided design rules for the four-bolt unstiffened end-plate configuration. Curtis (1985)reported extensive analytical (yield-line) and experimentalwork on washer flange stiffening at both four-bolt unstiffenedand eight-bolt stiffened, extended end plates.

Some of the literature cited was used to develop the designprocedures presented in the following chapter.

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Chapter 2

RECOMMENDED DESIGN PROCEDURES

2.1 BASIS OF DESIGN RECOMMENDATIONS

The recommended design procedures in Chapter 3 for thefour- and eight-bolt unstiffened end-plate configurations,Figs, 1.1(a) and (c), are based on the work of Krishnamur-thy (1978a), "A Fresh Look at Bolted End-Plate Behavior andDesign," and the procedures in the ASD and LRFD AISCmanuals (1980, 1986a, 1989a). Column side design for thefour-bolt configuration is based on the work of Hendrick andMurray (1984), "Column Web Compression Strength at End-Plate Connections," and Curtis and Murray (1989), "Col-umn Flange Strength at Moment End-Plate Connections."

The eight-bolt stiffened end plate, Fig. 1.1(d), design pro-cedures in Chapter 4 are based on the works of Ghassemeih(1983), "Inelastic Finite Element Analysis of Stiffened End-Plate Moment Connections," and Murray and Kukreti (1988),"Design of 8-bolt Stiffened Moment End Plates," and theprocedures in the 9th edition ASD AISC Manual of SteelConstruction (1989a). Column side design procedures for thisconfiguration are based on the previously cited works of Hen-drick and Murray (1984) and Curtis and Murray (1989).

In addition, the following assumptions or conditions areinherent to the design procedures:

1. All bolts are tightened to a tension not less than thatgiven in the AISC ASD and LRFD specifications.

2. The design procedures for the 8-bolt, stiffened con-figuration (Fig. 1.1(d)) are valid for use with A325bolts. A490 bolts should not be used in this configu-ration.

3. Only static loading is permitted. Temperature, windand snow loadings are considered static loadings(AISC, 1986, 1989). The design procedures should notbe used, pending further research, when seismic load-ing is a major design consideration.

4. The smallest possible bolt pitch (distance from faceof beam flange to centerline of nearer bolt) generallyresults in the most economical connection. The recom-mended minimum pitch dimension is bolt diameterplus ½ in. However, many fabricators prefer to usea standard pitch dimension, usually 2 in., for all boltdiameters.

5. End-plate connections can be designed to resist shearforce at the interface of the end-plate and columnflange using either "slip critical" or "bearing" assump-tions. If slip critical (type "SC") criteria are used, allbolts at the interface can be assumed to resist the shearforce and shear/tension interaction can be ignored asexplained in the Commentary on "Specification forStructural Joints Using ASTM A325 or A490 Bolts"

(RCSC, 1985). This Commentary states: "Connectionsof the type.. . in which some of the bolts lose a partof their clamping force due to applied tension sufferno overall loss of frictional resistance. The bolt ten-sion produced by the moment is coupled with a com-pensating compressive force on the other side of theaxis of bending." Thus, the frictional resistance of theconnection remains unchanged.

If very high shear forces exist, a bearing type con-nection may be necessary. In this case, the tensionbolts must be designed with a shear-tension interac-tion equation.

It is noted that shear is rarely a major concern inthe design of moment end-plate connections.

6. It is assumed that the width of the end plate whichis effective in resisting the applied beam moment isnot greater than the beam flange width plus 1 in. Thisassumption is based on engineering judgment and isnot part of any of the referenced end-plate design pro-cedures. Further, the writer is unaware of any end-plate connection tests conducted with end-plates sub-stantially greater in width than the connected beamflange.

7. The gage of the tension bolts (horizontal distancebetween vertical bolt lines) should not exceed the beamtension flange width, again based on engineeringjudgment.

8. Beam web to end-plate welds in the vicinity of the ten-sion bolts are designed to develop 0.6 of the beamweb. This weld strength is recommended even if thefull moment capacity of the beam is not required forframe strength.

9. Only the web to end-plate weld between the mid-depthof the beam and the inside side face of the beam com-pression flange or between the inner row of tensionbolts plus two bolt diameters and the inside face ofthe beam compression flange, whichever is smaller,may be used to resist the beam shear. This assump-tion is based on the author's opinion. Literature wasnot found to substantiate or contradict this assumption.

Column web stiffeners are expensive to fabricate and caninterfere with weak axis column framing. Therefore, it isrecommended that they be avoided whenever possible. If theneed for a stiffener is marginal, it may be more economicalto increase the column size rather than install stiffeners. Ifcolumn web stiffeners are required because of inadequatecolumn flange bending strength or stiffness, increasing theeffective length of the column flange may eliminate the needfor stiffening. This can be accomplished by increasing the

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tension bolt pitch or by switching from a two row configu-ration, Figs, 1.1(a), (b) or (c), to a four row configuration,Fig. 1.1(d). Alternately, column flange washer plates (looseplates with holes, placed on the column flange opposite theend-plate and connected with the end-plate connection ten-sion bolts) may be used. This approach is widely used inEurope (Mann and Morris, 1979; Zoetemeijer, 1981; Mooreand Sims, 1986) and has been studied in the United States(Curtis, 1985), but final design recommendations have notbeen formulated at this writing.

2.2 LIMIT STATES CHECK LISTLimit states (or failure modes) for moment end-plate beam-to-column connections are:

1. Flexural yielding of the end-plate material near thetension flange bolts. This state in itself is not limit-ing, but yielding results in rapid increases in tensionbolt forces and excessive rotation.

2. Shear yielding of the end-plate material. This limitstate is not usually observed, but shear in combina-tion with bending can result in reduced flexural capac-ity and stiffness.

3. Bolt rupture because of direct load and prying forceeffects. This limit state is obviously a brittle failuremode and is the most critical limit state in an end-plate connection.

4. Failure of bolt, or slip of bolt in slip critical connec-tions, due to shear at the interface between the endplate and column flange.

5. Plate bearing failure of end-plate or column flangeat bolts.

6. Rupture of beam tension flange to end-plate welds orbeam web tension region to end-plate welds.

7. Shear yielding of beam web to end-plate weld or ofbeam web base metal.

8. Column web yielding opposite either the tension orcompression flanges of the connected beam.

9. Column web buckling opposite the compression flangeof the connected beam.

10. Column flange yielding in the vicinity of the tensionbolts. As with flexural yielding of the end plate, thisstate in itself is not limiting but results in rapidincreases in tension bolt forces and excessive rotation.

11. Column web stiffener failure due to yielding, localbuckling or weld failure.

12. Column flange stiffener failure due to yielding or weldfailure.

13. Excessive rotation (flexibility) at the connection dueto end-plate and/or flange bending.

14. Column panel zone failure due to yielding or web platebuckling.

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Chapter 3UNSTIFFENED, EXTENDED END-PLATE CONNECTION DESIGN

3.1 THE FOUR-BOLT CONFIGURATIONDESIGN PROCEDURES AND EXAMPLES

3.1.1 Design proceduresThe following design recommendations for the extended,four-bolt, unstiffened, beam-to-column, end-plate connec-tion shown in Fig. 3.1 are based on the works of Krishnamur-thy (1978a), "A Fresh Look at Bolted End-Plate Behavior andDesign"; Hendrick and Murray (1984), "Column Web Com-pression Strength at End-Plate Connections"; and Curtis andMurray (1989), "Column Flange Strength at Moment End-Plate Connections." The basic procedures for end-plate andbolt design are also found in the AISC ASD Manual of SteelConstruction (1989a) and the LRFD Manual of Steel Con-struction (1986a).

In Krishnamurthy's design procedure, prying action forcesare considered negligible and the tension flange force is con-sidered to be distributed equally to the four tension bolts.Possible local yielding of the tension flange and tension areaof the web is neglected. The required end-plate thicknessis determined using the tee-stub analogy with the effectivecritical moment in the end plate given by

in ASD (3.1a)or

in LRFD (3.1b)

withunfactored beam flange force, kipsfactored beam flange force, kips

a constant depending on the plate material yieldstress, the bolt material and the design method(ASD or LRFD)

beam flange width, in.effective end-plate width, in. (not more than1 in.)area of beam tension flange, in.2

web area, clear of flanges, in.2

effective pitch, in.

distance from center line of bolt to nearer surfaceof the tension flange, in. + ½ in. is generallyenough to provide wrench clearance; 2 in. is a com-mon fabricator standard)fillet weld throat size or reinforcement of grooveweld, in.nominal bolt diameter, in.

The term was originally defined and values tabulatedin the AISC ASD manual. The same values were printed inthe AISC LRFD manual. However, to account for the differ-ences in weak axis bending strength between the AISC ASDand LRFD specifications, the original values of must beincreased by (0.90/0.75) = 1.20 for use in LRFD. Further,the values printed in both manuals are for cases where theend-plate and beam material have the same yield strengths,which is generally not the case except for A36 steel. Valuesof for various combinations of beam and end-platematerial are found in Tables A.2 and A.3 for ASD use andin Tables B.2 and B.3 for LRFD use. Tables A.2 and B.2are for A325 bolts and Tables A.3 and B.3 are for A490 bolts.

Values of for hot-rolled beam sections are found inTable A.4.

Fig. 3.1. Four-bolt unstiffened end-plate connection geometry.

7


The required end-plate thickness, , is then determinedfrom

in ASD (3.2a)or

in LRFD (3.2b)

with

the allowable bending stress for the end-plate mate-rial (0.75 times the specified yield stress), ksispecified yield stress of the end-plate material, ksi

The column side limit states are to be checked as follows:1. To prevent column web yielding at either the beam ten-

sion or compression flanges

in ASD (3.3a)or

in LRFD (3.3b)

with

factored beam flange force equal to times thebeam flange force when the flange force is due tolive and dead loads only, or by when the flangeforce is due to live and dead loads in conjunctionwith wind force, kipsspecified yield stress of column material, ksicolumn web thickness, in.beam flange thickness, in.distance from outer face of flange to web toe ofcolumn fillet, in.end-plate thickness, in.leg size of fillet weld or reinforcement weld, in.1.0

8

with

column flange thickness, in.required column flange thickness, in.

The required column flange thickness is determined usingEquation 3.2 modified as follows:

in ASD (3.5a)or

in LRFD (3.5b)

with

effective column flange length, in.2.5cvertical spacing between rows of tension bolts, in.

and or are calculated using Equations 3.1a or3.1b with for ASD and 1.36 for LRFD;

and

with

the column section distance, in.

If the selected criterion is not satisfied, standard col-umn flange to web stiffeners or flange washer platestiffeners can be used to increase the flexural strength ofthe column flanges.

4. To prevent column web shear yielding within the connec-tion, column web reinforcement is required if

in ASD (3.6a)or

in LRFD (3.6b)

with connected beam end moments, ft-kips,connected beam factored end moments, ft-

kips, and planar area of the column connection,in.2 In the above equations, the effect of column shearhas been conservatively ignored.

The following examples illustrate the above design proce-dures for four-bolt, unstiffened extended end-plate connec-tions. Examples 3.1 and 3.2 use the ASD format and Exam-ple 3.3 uses the LRFD format. For these examples, the beamtop flange is in tension and moment reversal is not aconsideration.

3.1.2 Allowable stress design examplesEXAMPLE 3.1. Use ASD procedures to design a beam-to-column end-plate connection for a moment of 200 ft-kipsand a shear of 40 kips. The beam is a W24x55 and the col-umn is a W14x159. A36 steel is used for all members and

If inequality 3.3 is not satisfied, column web stiffeners,capable of resisting a force equal to the difference betweenthe left and right sides of the inequality, must be provided.

2. To prevent column web buckling at the beam compres-sion flange

in ASD (3.4a)or

in LRFD (3.4b)

with

column web depth clear of fillets, in.0.90

If inequality 3.4 is not satisfied, column web stiffenersare required at the beam compression flange.

3. To prevent column flange yielding in the tension regionof the connection, the following must be satisfied assum-ing A36 material even if the column material yield stressis higher:

1.0;


plates. Bolts are ASTM A325. The end plate is to be shopwelded to the beam using E70XX electrodes.

W24x55

W14x159

A. Bolt design, A325-SC boltsThe beam tension flange force, is

The tension force per bolt, B, is then

From Table A.1, try diameter bolts (allowable capac-ity is 26.5 kips). Assuming A325-SC bolts, the single shearcapacity from Table A.1 is 10.5 kips. The number of boltsrequired to resist the applied shear is then

Bolt Selection

Use A325-SC boltsfully tightened, 4 at the ten-sion beam flange and 2 atthe compression beam flange.

B. End-plate design, A36 steelTry edge distance = 1¼ in.

gage, g = 5½ in.pitch,

Required end-plate width is 1¼ + 5½ + 1¼ = 8 in. Effec-tive end-plate width must be less than beam flange widthplus 1 in.

Determine from Equation 3.1a:


Check bolt bearing on end plate since it is less thick thanthe column flange. Assume, conservatively, that the com-pression side bolts resist all of the shear.

End-Plate Selection

C. Weld design, E70XX electrode

i. Beam flanges to end-plate welds:The flange weld must develop the force in the beam flange.For E70XX electrodes the capacity of a 1-in. longfillet weld is

Use ½-in. fillet welds at both beam flanges. Note minimumweld size from the AISC ASD Specification is ¼ in., whichcould be used at the beam compression flange if desired.

ii. Beam web to end-plate weld:Minimum size fillet weld is ¼ in.

Required weld to develop the bending stress in the beamweb near the tension bolts is

Use fillet weld both sides of beam web from insideface of beam flange to centerline of inside bolt holes plustwo bolt diameters.

The applied shear (40 kips) is to be resisted by weldbetween mid-depth of the beam and the inside face of thecompression flange or between the inner row of tension boltsplus two bolt diameters and the inside face of the compres-sion flange, whichever is minimum. By inspection the formergoverns for this example.

9

Check end-plate shear


Use ¼-in. fillet weld (minimum size for ¾-in thick plate)both sides of beam web below tension bolt region.

iii. Check beam web yielding

D. Check column side limit states and design stiffeners if nec-essary, A36 steel.

i. Check column web yielding using inequality 3.3a:

Therefore, stiffeners are not required opposite the beam ten-sion and compression flanges to prevent column webyielding.

ii. Check column web buckling using inequality 3.4a:

Therefore, neither column web or column flange stiffenersare required for this design.

Therefore, column web reinforcement is not required.Final design details are shown in Fig. 3.2.

EXAMPLE 3.2. Using the data, bolt design and end platefrom Example 3.1, determine if stiffeners are required if the

column is a W14x90 A572 Gr50 steel. Only the column sidelimit states need to be checked. ASD procedures apply.

Therefore, stiffeners are not required to prevent column webyielding.

10

Therefore, web stiffeners are not required opposite the beamcompression flange to prevent column web buckling.

iii. Check column flange bending:The required column flange thickness is determined usingEquation 3.2(a) with the previously discussed modifications.

iv. Check column web yielding using inequality 3.6a,50 ksi:

i. Check column web yielding using inequality 3.3a,50 ksi:

Fig. 3.2. Final design details, Example 3.1.


ii. Check column web buckling using inequality 3.4a, =50 ksi:

Therefore, stiffeners are not required to prevent column webbuckling.

iii. Check column flange bending:From Example 3.1, = 8.14 in., and from similar calcu-

lations = 1.66 in., =1.326, = 57.29 in.-kips and= 1.25 in. Note that this check is made assuming the

column material is A36 steel. Since is greater than0.710 in., a stiffener is required opposite the beam tensionflange. Because of the expense and possibility of interfer-ence with weak axis framing, the use of column web stif-feners is not recommended. Possible solutions for thisexample are: (1) to use an 8-bolt stiffened connection (Chap-ter 4) which increases the effective column flange length,(2) to increase the column flange thickness by using a heav-ier column or (3) to increase the bolt pitch which alsoincreases the effective column flange length. If the thirdchange is made, a thicker end-plate may be required. Obvi-ously, the suggested changes require additional expense; how-ever, the resulting connection may be more economicalbecause column web stiffeners are eliminated. If changes arenot practical, the following procedure can be used to deter-mine stiffener size.

Curtis and Murray (1989) do not provide recommendationsfor designing stiffeners when the column flange is inadequate.Assuming that only force in excess of what the unstiffenedcolumn flange can resist need be resisted by the stiffener,the capacity of the unstiffened column flange is first com-puted by rearranging Equation 3.2a and then 3.1a:

Thus, the stiffeners will be designed for the unfactored beamflange force less the capacity of the column flange:

With an allowable stress of

Stiffeners do not need to be full depth of the column web ifonly one beam is connected to the column at a given elevation.

Since the stiffener is in tension, local buckling is not a limitstate and AISC ASD Specification width and thickness rulesdo not apply; however, good engineering practice requiresthe stiffener to be proportioned to match the beam flange.

Try 2PL ½ x 4 x 0'-7

Use ¾-in. x ¾-in. clips to clear column web fillets.

Column flange to stiffener weld:

Minimum weld is ¼ in. Use fillet weld both sides.Column web to stiffener weld:

Minimum weld is To simplify detailing, use fil-let weld both sides.

Check shear stress in stiffener base metal.

Stiffener Selection

Use 2PL ½x4x0'-7 withfillet welds all around.

iv. Check column web yielding using inequality 3.6a,50 ksi:


3.1.3 Load and resistance factor design exampleEXAMPLE 3.3. Using LRFD procedures, design a beam-to-column end-plate connection for a factored moment of260 ft-kips, an unfactored shear of 40 kips and a factoredshear of 52 kips. The beam is a W24x55 and the columnis a W14x90. A36 steel is to be used for all members andplates. Bolts are A325. The end plate is to be shop weldedto the beam using E70XX electrodes.

11

A. Bolt design, A325-SC boltsThe factored beam tension flange force, is


The factored tension force per bolt, is then

From Table B.1, try diameter bolts (design strength is40.6 kips). Assuming A325-SC bolts, the single shear designstrength from Table B.1 is 10.2 kips. The number of boltsrequired to resist the applied shear (unfactored) is then

Bolt Selection

Use A325-SC boltsfully tightened, 4 at the ten-sion beam flange and 2 atthe compression beam flange.


B. End-plate design, A36 steelTry edge distance = 1¼ in.

gage = 5½ in.pitch,

Required end-plate width is 1¼ + 5½ + 1¼ = 8 in. Effec-tive end-plate width must be less than beam flange widthplus 1 in.

Determine from Equation 3.1b:

Determine from Equation 3.2b:

Check bolt bearing on end-plate (note column flange thick-ness is larger and, conservatively, only the compression sidebolts are considered).

in. Check end-plate shear:

End-Plate Selection


i. Beam flanges to end-plate welds:Flange weld must develop the force in beam flange. ForE70XX electrodes the capacity of a 1-in. long filletweld is

Use fillet welds at beam tension flange and minimumweld size at beam compression flange. From the AISC LRFDSpecification minimum weld size is ¼ in.

12


ii. Beam web to end-plate weld:Minimum size fillet weld is ¼ in.

Required weld to develop maximum bending stressin web near tension bolts is

Use fillet weld both sides of beam web from insideface of beam flange to centerline of inside bolt holes plustwo bolt diameters.

The factored shear (52 kips) is to be resisted by weldbetween mid-depth of the beam and the inside face of thecompression flange or between the inner row of tension boltsplus two bolt diameters and the inside face of compressionflange, whichever is minimum. By inspection the former gov-erns for this example.

Use ¼-in. fillet weld (minimum size for ¾-in. thick plate)both sides of beam web below tension bolt region.

iii. Check beam web yielding:

D. Check column side limit states and design stiffeners if nec-essary, A36 steel

i. Check column web yielding using Inequality 3.3b:


ii. Check column web buckling using Inequality 3.4b:

Therefore, web stiffeners are not required opposite the beamcompression flange to prevent column web buckling.

iii. Check column flange bending:The required column flange thickness is determined usingEquation 3.2b with the modifications that resulted in Equa-tion 3.5b.

Therefore, a stiffener is required opposite the beam tensionflange. As was previously discussed, because of the expenseand possibility of interference with weak axis framing, theuse of column web stiffeners is not recommended. Possiblesolutions for this example are: (1) to use an 8-bolt, stiffenedend-plate (Chapter 4) which increases the effective columnflange length, (2) to increase the column flange thicknessby using a heavier column or (3) to increase the bolt pitchwhich increases the effective column flange length anddecreases the required column flange thickness. If the thirdchange is made, a thicker end-plate may be required. Obvi-ously, any change requires additional expense; however, theresulting connection may be more economical if the columnweb stiffeners are eliminated. If changes are not practical,the following procedure can be used to determine stiffenersize.

Assuming only force in excess of what the unstiffened col-umn flange can resist need be resisted by the stiffener, thecapacity of the unstiffened column flange is first computed.

Stiffeners do not need to be full depth of the column webif only one beam is connected to the column at a givenelevation.

Since the stiffener is in tension, local buckling is not a limitstate and AISC LRFD specification width and thickness rulesdo not apply; however, good engineering practice requiresthe stiffener to be proportioned to be compatible with thebeam flange. Assume ¾-in. "clip" to clear column webfillets.

13

Thus, the stiffener will be designed for

The required stiffener area is then

Column flange to stiffener weld:

Minimum weld is ¼ in. Use fillet weld both sides.


Column web to stiffener weld:

Minimum weld is To simplify detailing, usefillet weld both sides.

Check shear force in stiffener base metal using AISC spec-ification Equation J5-3 (length along flange governs).

Stiffener Selection

Use 2PL ½x4x0'-7 withfillet welds.

iv. Check column web yielding using inequality 3.6b:


3.2 EIGHT-BOLT DESIGN PROCEDURESAND ALLOWABLE STRESS DESIGNEXAMPLE

The design procedures for unstiffened extended end-platesin the AISC Manuals imply that the end-plate configurationshown in Fig. 1.1(c) can be designed using the work of Krish-namurthy (1978a). The work of Hendrick and Murray (1984)can be used to evaluate column web yielding and buckling.Column flange bending strength requires special considera-tion. A suggested approach is given in the following ASDexample. Only slight modifications are required for LRFDdesign (see Example 3.3).

EXAMPLE 3.4. Design a beam-to-column end-plate connec-tion for a moment of 700 ft-kips and a shear of 90 kips usingASD procedures. The beam is a W33x118 and the columnis a W14x311. All material is A36. Bolts are A325 and arelimited to 1-in. diameter. E70XX electrodes will be used forall welding. The beam top flange is in tension and momentreversal is not a consideration.


By inspection 8 bolts are required. The force per bolt, B,is then

From Table A.1, try 1-in. diameter bolts (allowable capac-ity is 34.6 kips). Assuming A325-SC bolts, the single shearcapacity from Table A.1 is 13.7 kips. The number of boltsrequired to resist the applied shear is then

Bolt Selection

Use 12 1-in. diameterA325-SC bolts fully tight-ened, 8 at beam tensionflange and 4 at beam com-pression flange.


14


B. End-plate design, A36 steel

Try edge distance = 1¼ in.gages = 6 in. and 12 in. (inside and outside bolts)

pitch,

Required end-plate width is 1¼ + 3 + 6 + 3 + 1¼ = 14½in. (Note column flange width is 16¼ in.) Effective end-platewidth must be less than beam flange width plus 1 in.


Check bolt bearing on end-plate (note column flange thick-ness is larger and, conservatively, only the compression sidebolts are considered).

End-Plate Selection


i. Beam flanges to end-plate welds:By inspection, fillet welds will be impractical; therefore, usefull penetration groove weld with reinforcement atbeam tension flange. Use minimum weld at beamcompression flange.

ii. Beam web to end-plate weldMinimum size of fillet weld is

The required weld to develop the bending stress in thebeam web near the tension bolts is

Use fillet weld both sides of beam web from insideof beam flange to centerline of bolt holes plus two boltdiameters.

The applied shear is to be resisted by weld between mid-depth of the beam and the inside face of the compressionflange or between the inner row of tension bolts plus twobolt diameters and the inside face compression flange, which-ever is minimum. By inspection the former governs for thisexample.

Use fillet weld (minimum size for thick plate)both sides of beam web below tension bolt region.


D. Check column side limit states and design stiffeners if nec-essary, A36 steel

i. Check column web yielding using inequality 3.3a:

Therefore, stiffeners are not required opposite the beam ten-sion and compression flanges to prevent column web yielding.

ii. Check column web buckling using inequality 3.4a:

iii. Check column flange bendingDesign procedures are not available to assess the columnflange bending strength for this bolt pattern. However, thestrength can be evaluated if a small triangular stiffenerbetween the column flange and the column web is used sincethis pattern is similar to that of the eight-bolt stiffened end-plate discussed in Chapter 4. When this approach is usedthe column web is equivalent to the beam flange and the col-umn flange is equivalent to the end-plate. Since test data isnot available, it is recommended that the effective columnflange length (equivalent to the end-plate width) be takenas that recommended for the four-bolt configuration (Curtisand Murray, 1989), e.g., 2.5c. With reference to Chapter 4,for details of the design procedure, the column flange forthis example is now checked. (See ASD nomenclature fordefinition of terms.) Details are shown in Fig. 3.5.

15


The equivalent gage and pitch are

and the equivalent beam flange and end-plate thicknesses are

The effective end-plate width is equal to 2.5c or

The column flange-to-web stiffener should be approximatelyequal to the beam flange thickness (0.740 in.) and extendbeyond the outside row of bolts, thus use a rectangular plate¾ in. x 7 in. x 7 in. Since all limitations given in Chapter4 are satisfied, the simplified method, Equation 4.4a, canbe used to determine the adequacy of the stiffened columnflange. From Equation 4.7a

with


16

Since only 6 bolts are assumed effective, the capacity of thestiffened flange is 6 x 106.4 = 638.4 kips which is greaterthan the applied beam flange force of 261.5 kips and thestiffened column flange is adequate.

Conservatively, the stiffener to flange and web welds willbe designed for the applied beam flange force. Assuming a1½ in. "clip" to clear the column fillet, the required filletweld size is

Use fillet welds both sides of stiffener. (Full penetra-tion groove welds are not practical at this location.)

Stiffener Selection

Use 2PL ¾ x 7 x 0'-7with fillet welds.

iv. Check column web yielding using inequality 3.6a:


From Equation 4.4a

and from Equation 4.5a


Chapter 4

STIFFENED, EXTENDED END-PLATE CONNECTION DESIGN

4.1 DESIGN PROCEDURES

The following ASD design recommendations for theextended, four-bolt, stiffened, beam-to-column, end-plateconnection shown in Fig. 4.1 are based on the works of Mur-ray and Kukreti (1988), "End-Plate Moment Connections—Their Use and Misuse," Hendrick and Murray (1984), "Col-umn Web Compression Strength at End-Plate Connections,"and Curtis and Murray (1989), "Column Flange Strength atMoment End-Plate Connections." The basic procedures forend-plate and bolt design are also found in the 9th ed. AISCASD Manual of Steel Construction.

Murray and Kukreti (1988) present two methods for deter-mining end-plate thickness and bolt diameter. Both methodsare limited to use for A36 end-plate steel and A325 boltsand both include bolt prying action effects. The first methodis a series of equations developed from regression analysesof data generated by the finite element method. The finiteelement model included both second order geometry effectsand inelastic plate and bolt material properties. With thismethod, the required end-plate thickness is the larger ofand determined from (see Fig. 4.1 for definition ofterms):

in ASD (4.1a)

in ASD (4.2a)

or

in LRFD (4.1b)

in LRFD (4.2b)

The regression-based Equations 4.1 are stiffness criteriawhich control end-plate flexibility for use in Type I construc-tion. Equations 4.2 are strength criteria which limit maxi-mum strain on the end-plate. Both ASD equations includea factor of safety of 1.67 and both LRFD equations includea resistance factor of 0.9.

Ultimate bolt force including prying action effects is esti-mated from

with = minimum bolt tension as given in AISC specifi-cations and reproduced here for A325 bolts in Tables A.1and B.1. Equation 4.3a includes a factor of safety of 2.0.Equation 4.3b does not include a resistance factor, thus thespecified minimum tensile strength of the bolt material mustbe used to determine the required bolt diameter.

In the application of Equations 4.1, 4.2 and 4.3, a prelimi-nary bolt diameter is selected assuming that 6.8 of the 8 ten-sion bolts are effective. This ratio must often be decreased

Fig. 4.1. Eight-bolt stiffened end-plate connection geometry.

17


if Equation 4.3 shows the selected bolt diameter to beinadequate.

The second method is a simplified approach which wasformulated because of the difficulty in using Equations 4.1,4.2 and 4.3, except for completely computerized designs. Themethod was developed by first generating end-plate and boltsizes using the above equations for all hot-rolled A36 steelbeam sections at various moment levels. An effective num-ber of bolts was then determined for each connection anda conservative lower bound of six bolts established. Next,it was assumed that plate thickness could be established fromtee-stub analogy bending (see Fig. 4.2), that is,

in ASD (4.4a)or

in LRFD (4.4b)

with force per bolt based on six effective boltsand an effective pitch. From the generated designsit was determined that

in ASD (4.5a)

or

in LRFD (4.5b)

The required end-plate thickness is then determined from

in ASD (4.6a)or

in LRFD (4.6b)with

in ASD (4.7a)

Fig. 4.2. Tee-stub analogy moments.

The following limitations, in addition to those given inChapter 2, apply to the simplified method:

1. The connected beam section must be hot-rolled andincluded in the "Allowable Stress Design SelectionTable" in the AISC ASD Manual.

2. The vertical pitch, from the face of the beam ten-sion flange to the first row of bolts must not exceed2½ in. The recommended minimum pitch is boltdiameter plus ½ in.

3. The vertical spacing between bolt rows, must notexceed

4. The horizontal gage, g, must be between 5½ and 7½ in.5. Bolt diameter must not be less than ¾ in. nor greater

than 1½ in.The recommendations of Hendrick and Murray (1984) can

be used to check column web yielding at either the beamtension or compression flanges (inequality 3.3) and columnweb buckling at the beam compression flange (inequality 3.4).Since Type I construction is assumed for this connection,a stiff column flange is required. Thus, unless the columnflange is considerably thicker than the end-plate, flange toweb stiffeners are required. If effective flange length effectsare neglected, the behavior of the column flange is identicalto that of the end-plate and, therefore, the column flangemust be at least as thick as the end-plate, and the columnstiffener must be as thick as the beam flange. Further, thestiffener to flange weld must be sufficient to develop thestrength of the full thickness of the stiffener plate.

If the column flange is substantially thicker than the end-plate (1.5-2 times), stiffeners may not be necessary. Basedon the work of Curtis and Murray (1989), such an unstiffenedflange can be evaluated using Equation 3.5 with

(4.8)

The referenced work included only A36 steel. Therefore, itis recommended that if the column material yield strengthis greater than 36 ksi, the column flange strength be checkedassuming A36 steel is being used.

Column web shear strength should be checked using in-equality 3.6.

4.2 DESIGN EXAMPLES

4.2.1 Allowable stress design examplesThe following three examples demonstrate the use of theabove ASD procedures. Example 4.1 uses the simplifieddesign method, Equation 4.6a. Example 4.2 uses the moreexact design method, Equations 4.1a, 4.2a and 4.3a. Exam-ple 4.3 demonstrates the ASD procedures for checking thecolumn side of the connection. For all examples, the beamtop flange is in tension and moment reversal does not occur.

18


EXAMPLE 4.1. Using the ASD procedures, design a beam-to-column end-plate connection for a moment of 700 ft-kipsand a shear force of 90 kips. The beam is a W33x118 andthe column is a W14x311. All material is A36 steel. Boltsare A325. E70XX electrodes will be used for all welding.Use Equation 4.6a to determine end-plate thickness andassume only 6 bolts are effective.

Assuming 6 bolts effective, the force per bolt is

From Table A.1, try diameter bolts (allowable capac-ity is 43.7 kips).

Assuming A325-SC bolts, the single shear capacity fromTable A.1 is 17.4 kips. The number of bolts required to resistthe applied shear is then

Bolt Selection

Use diameterA325-SC bolts fully tight-ened, 8 at beam tensionflange and 2 at compressionflange.

(Note if the four bolt unstiffened configuration shown in Fig.1.1(a) is used, the required bolt diameter is

B. End-plate design, A36 steelTry edge distance = 1½ in.

gage g = 6 in.pitchpitch between bolt rowsstiffener thickness

Note that all of the specified limitations for the simplifiedmethod are satisfied.

Minimum end-plate width is

Effective end-plate width must be less than or equal to thebeam flange width plus 1 in., e.g.12.48 in. Use 12½ in. end-plate width and

Determine effective pitch from Equation 4.5a,

and the equivalent tee-stub analogy moment from Equation4.4a

The required section modulus is then

And the required end-plate thickness from Equation 4.6a is

Check bolt bearing on end-plate (note that (1) column flangethickness is larger and (2) conservatively only the compres-sion side bolts are considered).

End-Plate Selection

C. Weld design, E70XX electrodes

i. Beam flanges to end-plate welds:By inspection, the fillet welds will be impractical. Use fullpenetration groove weld with reinforcement at beamtension flange and fillet weld (minimum for.1¼-in.plate at beam compression flange).

ii. Beam web to end-plate weld:Minimum size fillet weld is Conservatively, therequired weld to develop the bending stress in the beam webnear the tension bolts is

Use fillet weld both sides of beam web from insideface of beam flange to centerline of innermost bolt holes plustwo bolt diameters.

The applied shear is to be resisted by weld between theminimum of the mid-depth of the beam and the compres-sion flange or the inner row of tension bolts plus two boltdiameters and the compression flange. By inspection theformer governs for this example.

Use fillet weld (minimum size for 1¼-in. thick plate)both sides of beam web below tension bolt region.

19

A. Bolt design, A325-SC boltsThe beam tension flange force,


Column side limit states are checked in Example 4.3.

EXAMPLE 4.2. For the conditions of Example 4.1, deter-mine required end-plate thickness and bolt diameter usingEquations 4.1a, 4.2a and 4.3a. ASD procedures apply.

A. Trial bolt size, A325-SC boltsFrom Example 4.1, the flange force is 261.5 kips. A trial boltsize is selected assuming 6.8 bolts are effective.

From Table A.1, try diameter bolts (allowable capac-ity is 43.7 kips).

B. End-plate design, A36 steelTry: edge distance = 1½ in.

gage g = 6 in.pitchpitch between bolt rowsstiffener thickness

From Example 4.1, use 12½-in. wide end-plate with =12.48 in.

Determine the required end-plate thickness from the stiff-ness criterion, Equation 4.1a.

Determine the required end-plate thickness from the strengthcriterion, Equation 4.2a.

Check adequacy of diameter bolts using Equation4.3a.

The ultimate bolt force must be less than the tensile strengthof the bolt which is twice the allowable capacity given inTable A.1, that is

20

Thus, diameter A325-SC bolts are satisfactory. Sincethe end-plate thickness and bolt diameter are the same asin Example 4.1, the number of bolts required to resist theshear force is the same and bolt bearing is adequate. Hence,the final design using the regression based Equations 4.1a,4.2a and 4.3a is identical to that obtained using the split-teeanalogy method, Equation 4.6a. Column side limit states arechecked in Example 4.3.

EXAMPLE 4.3. Using the data, bolt design and end-platefrom Example 4.1, determine if stiffeners are required if thecolumn is a W14x311 A36 steel. Only the column side limitstates need to be checked. ASD procedures apply.


ii. Check column web buckling using inequality 3.4a, A36steel:

Therefore column web stiffeners are not required to preventcolumn web buckling.

iii. Check column flange bending, A36 steel:Since the column flange is significantly thicker than the end-plate, column flange stiffeners may not be required. Theunstiffened column flange can be investigated using Equa-tion 3.2a with appropriate modifications. From Curtis andMurray (1989), the effective column flange length, whichis equivalent to the end-plate width in Equation 3.2, is

iii. Check beam web shear yielding:

i. Check column web yielding using inequality 3.3a, A36 steel:


Fig. 4.3. Final design details for eight-bolt stiffened end-plateexamples.

iv. Check column web yielding using inequality 3.6a, A36steel:


4.2.2 Load and resistance factor design examplesThe following three examples demonstrate the use of theLRFD procedures. Example 4.4 uses the simplified designmethod, Equation 4.6b. Example 4.5 uses the more exactmethod, Equations 4.1b, 4.2b and 4.3b. Example 4.6 dem-onstrates the LRFD procedures for checking the column sideof the connection. For all examples, the beam top flange isin tension and moment reversal does not occur.

EXAMPLE 4.4. Using the LRFD procedures, design a beam-to-column end-plate connection for a factored moment of1050 ft-kips, an unfactored shear force of 90 kips and a fac-tored shear force of 135 kips. The beam is a W33x118 andthe column is a W14x311. All material is A36 steel. Boltsare A325. E70XX electrodes will be used for all welding.Use Equation 4.6b to determine end-plate thickness andassume only 6 bolts are effective.


Assuming 6 bolts effective, the force per bolt is

From Table B.1, try diameter bolts (design tension loadis 67.1 kips).

Assuming A325-SC bolts, the single shear capacity fromTable B.1 is 16.9 kips. The number of bolts required to resistthe applied shear is then 90 / 16.9 = 5.3.

Bolt Selection

Use diameterA325-SC bolts fully tightened,8 at beam tension flangeand 2 at beam compressionflange.

(Note if the four bolt unstiffened configuration shown in Fig.1.1(a) is used, the required bolt diameter is

21

The required flange thickness from Equation 3.2 is

Therefore, column web stiffeners are not required for thisexample.


B. End-plate design, A36 steelFrom Example 4.1 use:

gage g = 6 in.pitchpitch between bolt rowsstiffener thicknessend-plate width = 12½ in.effective end-plate width

Note that all of the specified limitations for the simplifiedmethod are satisfied.

Determine effective pitch from Equation 4.5b.

and the equivalent tee-stub analogy moment from Equation4.4b

The required section modulus is then

And the required end-plate thickness from Equation 4.6b is

Check bolt bearing on end-plate (note column flange thick-ness is larger and conservatively only the compression sidebolts are considered).

End-Plate Selection


i. Beam flanges to end-plate welds:By inspection, fillet welds will be impractical. Use fullpenetration groove weld with reinforcement at beamtension flange and fillet weld (minimum for 1¼-in.plate at beam compression flange).

ii. Beam web to end-plate weld:Minimum size fillet weld is Required weld to developmaximum bending stress in web near tension bolts is

Use fillet weld both sides of beam web from insideface of beam flange to centerline of innermost bolt holes plustwo bolt diameters.

The applied shear is to be resisted by weld between theminimum of the mid-depth of the beam and the compres-sion flange or the inner row of tension bolts plus two boltdiameters and the compression flange. By inspection theformer governs for this example.

Use fillet weld (minimum size for 1¼-in. thick plate)both sides of beam web below tension bolt region.

iii. Check beam web yielding:

Column side limit states are checked in Example 4.6.

EXAMPLE 4.5. For the conditions of Example 4.4, deter-mine required end-plate thicknesses and bolt diameter usingEquations 4.1b, 4.2b and 4.3b. LRFD procedures apply.

A. Trial bolt size, A325-SC boltsFrom Example 4.4, the factored flange force is 392.3 kips.A trial bolt size is selected assuming 6.8 bolts are effective.

From Table B.1, try diameter bolts (allowable capac-ity is 67.1 kips).

B. End-plate design, A36 steelFrom Example 4.1 use:

Determine the required end-plate thickness from the stiff-ness criterion, Equation 4.1b.

Determine the required end-plate thickness from the strengthcriterion, Equation 4.2b.

22


Check adequacy of diameter bolts using Equation4.3b.

The ultimate bolt force must be less than the tensile strengthof the bolt which is the design tension capacity given in TableB.1, that is

Thus, diameter A325-SC bolts are satisfactory. Sincethe end-plate thickness and bolt diameter are the same asin Example 4.4, the number of bolts required to resist theshear force is the same and bolt bearing is adequate. Hence,the final design using the regression based Equations 4.1b,4.2b and 4.3b is identical to that obtained using the split-teeanalogy method, Equation 4.6b. Column side limit states arechecked in Example 4.6.

EXAMPLE 4.6. Using the data, bolt design and end-platefrom Example 4.4, determine if stiffeners are required if thecolumn is a W14x311 A36 steel. Only the column side limitstates need to be checked. LRFD procedures apply.


ii. Check column web buckling using inequality 3.4b, A36steel:

Therefore column web stiffeners are not required to preventcolumn web buckling.

iii. Check column flange bending, , A36 steel:Since the column flange is significantly thicker than the end-plate, column flange stiffeners may not be required. Theunstiffened column flange can be investigated using Equa-tion 3.2b with appropriate modifications. From Curtis andMurray (1989), the effective column flange length, whichis equivalent to the end-plate width in Equation 3.2b, is

Therefore, column web reinforcement is not required.Final design details are the same as for the ASD Example

4.3 and are shown in Fig. 4.3.

23

i. Check column web yielding using inequality 3.3b, A36 steel:

Therefore, column web stiffeners are not required for thisexample.

iv. Check column web yielding using inequality 3.6b, A36steel:

The required flange thickness from Equation 3.2b is


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Murray, T. M. and A. Kukreti (1985), "Design of 8-BoltStiffened End-Plate Moment Connections," Papers, ThirdConference on Steel Developments, Australian Institute ofSteel Construction, Melbourne, Australia, May 20-22, 1985,pp. 145-149.

Murray, T. M. and A. Kukreti (1988), "Design of 8-bolt

Stiffened Moment End-Plates," Engineering Journal, Ameri-can Institute of Steel Construction, Volume 25, No. 2, 2ndQuarter, 1988, pp. 45-52.

Nair, R. S., P. C. Birkemoe and W. H. Munse (1969),"Behavior of Bolts in Tee Connections Subject to PryingAction," Civil Engineering Studies, University of Illinois,September 1969.

Nair, R. S., P. C. Birkemoe and W. H. Munse (1974), "HighStrength Bolts Subject to Tension and Prying," Journal ofthe Structural Division, ASCE, Vol. 100, No. 2, Feb. 1974,pp. 351-372.

Naka, T., B. Kato and M. Watabe (1966), "Research on theBehavior of Steel Beam-to-Column Connections, Laboratoryfor Steel Structures, Dept, of Architecture, University ofTokyo, Japan 1966.

Nakao, M. (1973), "Experiments on the Behavior of SteelBeam-to-Column Connections with Large Axial Force onColumn," Annual Report of the Engineering Institute, Fac-ulty of Engineering, University of Tokyo, Vol. 32, 1973.

Nethercot, D. A. (1985), "Joint Action and the Design ofSteel Frames," The Structural Engineer, Vol. 63A, No. 12,December 1985, pp. 371-379.

Newlin, D. E. and W. E. Chen (1971), "Strength and Stabil-ity of Column Web in Welded Beam-to-Column Connec-tions," Fritz Engineering Laboratory, Report No. 333.14,Lehigh University, Bethlehem, PA, May 1971.

Onderdonk, A. B., R. P. Lathrop and J. Goel (1964), "End-Plate Connections in Plastically Designed Structures," Engi-neering Journal, AISC, Vol. 1, No. 1, Jan. 1964, pp. 24-27.

Oswalt, R. E. (1978), "Two-Dimensional Finite ElementAnalysis of the Effects of Bolt Heads and Welds in Steel End-Plate Connections," Unpublished thesis, Vanderbilt Univer-sity, Nashville, TN, May 1978.

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29


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Zoetemeijer, P. (1981), "Semi-Rigid Bolted Beam-to-ColumnConnections with Stiffened Column Flanges and Flush End-Plates," Joints in Structural Steelwork, John Wiley & Sons,London-Toronto, 1981, pp. 2.99-2.118.

Zoetemeijer, P. (1981a), "Bolted Connections with FlushEnd-Plates and Haunched Beams, Tests and Computations,"Stevin Laboratory, Delft University of Technology, 1981.

Zoetemeijer, P. and M. H. Kolstein (1975), "Bolted Beam-to-Column Connections with Flush End-Plates, Tests andComputation Methods" (in Dutch), Report 6-75-20, StevinLaboratory, Delft University of Technology, 1975.

30


Appendix AASD NOMENCLATURE, DESIGN AIDS,AND QUICK REFERENCE EXAMPLES

A.1 ASD NOMENCLATURE

planar area of column connection, in.2

area of beam tension flange, in.2

gross area of plate, in.2

column stiffener area, in.2

area of beam web, in.2

beam or column flange width, in.effective end-plate width, in. (not more than

+ 1 in.)effective column flange length, in.unfactored tension force per bolt, kip sallowable tension load in bolt, kipsultimate bolt force including prying action effects,kipsvertical spacing between rows of tension bolts, in.a constant depending on the plate material yieldstress, the bolt material and the design method

depth of beam or column section, in.nominal bolt diameter, in.column web depth clear of fillets, in.required fillet weld throat size, sixteenthsedge distance, in.computed bearing stress, ksicomputed shear stress, ksiaverage yield stress for beam and end-plate mate-rials, ksiallowable bending stress for the end-plate mate-rial (0.75 times the specified yield stress), ksiallowable bending stress for column flange mate-rial (0.75 times the specified yield stress), ksiallowable tensile stress for bolt material, ksispecified minimum tensile strength for bolt mate-rial, ksicapacity of unstiffened column flange to resistapplied force, kipsunfactored beam flange force, kipsallowable bearing stress, ksispecified minimum tensile strength, ksiallowable shear stress, ksi

specified yield stress of the end-plate material, ksispecified yield stress of column material, ksihorizontal spacing between vertical bolt lines, in.distance from outer face of flange to web toe offillet, in.the column section distance, in.unfactored effective end-plate moment, in.-kipsconnected beam end moments, ft-kipsrequired number of bolts to resist beam sheareffective pitch, in.pitch, distance from center line of bolt to nearersurface of the tension flange, in. + ½ in. isgenerally enough to provide wrench clearance)factored beam flange force equal to times thebeam flange force when the flange force is dueto live and dead loads only, or by when theflange force is due to live and dead loads in con-junction with wind or earthquake forces, kipseffective pitch, in.minimum bolt tension, kipsrequired end-plate elastic section modulus, in.3

beam flange thickness, in.column flange thickness, in.required column flange thickness, in.end-plate thickness, in.required end-plate thickness from stiffnesscriterion, in.required end-plate thickness from strengthcriterion, in.end-plate to beam tension flange stiffener thick-ness (approximately equal in thickness to that ofthe beam web), in.beam web thickness, in.column web thickness, in.column web depth clear of fillets, in.single shear capacity of bolt, kipsleg size of fillet weld or reinforcement weld, in.

31


A.2 ASD DESIGN AIDS

Table A.1.Allowable Tension and Single Shear Loads for A325 and A490 Bolts

(ASD Method)a) A325 Bolts

Diameter (in.)Tension (kips)

Shear A325-SC (kips)

Shear A325-N (kips)

Shear A325-X (kips)

Minimum Bolt Tension (kips)

13.5

5.4

6.4

9.2

19

19.4

7.7

9.3

13.3

28

26.5

10.5

12.6

18.0

39

34.6

13.7

16.5

23.6

51

43.7

17.4

20.9

29.8

56

54.0

21.5

25.8

36.8

71

65.3

26.0

31.2

44.5

85

77.7

30.9

37.1

53.0

103

b) A490 Bolts

Diameter (in.)

Tension (kips)


Shear A490-N (kips)

Shear A490-X (kips)


16.6

6.7

8.6

12.3

24

23.9

9.7

12.4

17.7

35

32.5

13.2

16.8

24.1

49

42.4

17.3

22.0

31.4

64

53.7

21.9

27.8

39.8

80

66.3

27.0

34.4

49.1

102

80.2

32.7

41.6

59.4

121

95.4

38.9

49.5

70.7

148

All values from AISC ASD Manual (1980, 1989).

32

Table A.2.ASD Values of for A325 Bolts

Table A.3.ASD Values of for A490 Bolts


Table A.4.Values of

Section

W36 x 359 0.899x328 0.903x300 0.887x280 0.882x260 0.850x245 0.835x230 0.818x256 0.648x 232 0.644x210 0.588x194 0.587x182 0.579x170 0.573x160 0.554x150 0.530x135 0.463

W33x354 0.925x318 0.926x291 0.913x 263 0.909x 241 0.853x 221 0.829x 201 0.807x169 0.667x152 0.612x141 0.583x130 0.541x118 0.492

W30 x 235 0.961x211 0.905x191 0.887x173 0.861x148 0.672x132 0.606x124 0.590x116 0.558x108 0.516x 99 0.476

Section

W27x217 1.003x194 0.986x178 0.909x161 0.902x146 0.885x129 0.710x114 0.646x102 0.635x 94 0.597x 84 0.545

W24 x 176 1.021x162 0.994x146 0.959x131 0.904x117 0.877x104 0.848x103 0.711x 94 0.683x 84 0.655x 76 0.616x 68 0.560x 62 0.428x 55 0.397

W21 x 166 1.140x147 1.011x132 1.002x122 1.003x111 0.994x101 0.995x 93 0.683x 83 0.686x 73 0.683x 68 0.667x 62 0.641x 57 0.532x 50 0.465x 44 0.423

Section

W18x143 1.204x130 1.186x119 1.082x106 1.059x 97 1.076x 86 1.056x 76 1.048x 71 0.741x 65 0.751x 60 0.751x 55 0.722x 50 0.714x 46 0.604x 40 0.595x 35 0.504

W16x100 1.170x 89 1.152x 77 1.146x 67 1.149x 57 0.789x 50 0.781x 45 0.768x 40 0.772x 36 0.679x 31 0.589x 26 0.506

W14x120 1.855x109 1.899x 99 1.859x 90 1.860x 82 1.348x 74 1.394x 68 1.382x 61 1.364x 53 1.141x 48 1.115x 43 1.103x 38 0.861x 34 0.824x 30 0.734x 26 0.633x 22 0.557

Section

W12x87 1.748x79 1.732x72 1.720x65 1.706x58 1.631x53 1.527x50 1.281x45 1.266x 40 1.281x35 0.992x30 0.963x26 0.936x22 0.575x19 0.520x16 0.419x14 0.390

W10x60 1.842x 54 1.882x49 1.859x45 1.603x39 1.516x33 1.348x 30 1.045x 26 1.033x22 0.913x19 0.672x17 0.583x15 0.497x12 0.463

W 8x35 1.796x31 1.711x 28 1.495x 24 1.487x21 1.127x18 1.007x15 0.690x13 0.593x10 0.635

W 6x25 1.580x20 1.545x15 1.238x16 1.148x12 0.890x 9 0.911

W 5x19 1.867x16 1.748

W 4x13 1.442

33


A.3 ASD QUICK REFERENCE EXAMPLES

EXAMPLE A.1 (Same as Example 3.1)

4-bolt Unstiffened End-Plate

Beam W24x55 A36 steel

Column W14x159 A36 steel

A. Bolt design, A325-SC bolts

i. Tension:

ii. Shear,


i. Bending, Equation 3.1a:

ii. Check bolt bearing, end-plate controls, compression bolts:

iii. Check end-plate shear:

C. End-plate weld design, E70XX electrodes

i. Beam flanges to end-plate weld:

ii. Beam web to end-plate weld:


D. Column side, A36 steel and E70XX electrodes

i. Check column web yielding, inequality 3.3a,

34


Column web reinforcement is not required.

E. Final details:

35

ii. Check column web buckling, inequality 3.4a,

iii. Check column flange bending,

Column web stiffeners are not required.

iv. Check column web yielding, inequality 3.6a:

EXAMPLE A.2 (Same as Example 3.2)Data is same as Example A.1, except

Column W14x90 A572 Gr 50 steel

A. Column side

i. Check column web yielding, inequality 3.3a:

ii. Check column web buckling, inequality 3.4a:


Calculations to be made with

Design stiffeners and welds for

Column flange to stiffener weld, E70XX electrodes:

Column web to stiffener weld, E70XX electrodes:

Check shear stress in stiffener base metal, A36 steel:

iv. Check column web yielding inequality 3.6(a),


B. Final details:

EXAMPLE A.3 (Same as Examples 4.1 and 4.3)

36

8-Bolt Stiffened End-PlateSimplified procedure




i. Tension:

ii. Shear,


i. Bending, Equation 4.4a:



i. Beam flanges to end-plate weld:Use full penetration groove weld with reinforcement.




i. Check column web yielding, inequality 3.3a,



D. Final details:

37

ii. Check column web buckling, inequality 3.4a,



iv. Check column web yielding, inequality 3.6a:


Appendix BLRFD NOMENCLATURE, DESIGN AIDS,AND QUICK REFERENCE EXAMPLES

B.1 LRFD NOMENCLATURE

= planar area of the column connection, in.2

= area of beam tension flange, in.2

= gross area of plate, in.2

= column stiffener area, in.2

= area of beam web, in.2

= beam or column flange width, in.= effective end-plate width, in. (not more than

+ 1 in.)= effective column flange length, in.= design tension capacity of bolt, kips= factored tension force per bolt; ultimate bolt

force including prying action effects, kips= vertical spacing between rows of tension bolts, in.= a constant depending on the plate material yield

stress, the bolt material and the design method.

= depth of beam or column section, in.= nominal bolt diameter, in.= column web depth clear of fillets, in.= required fillet weld throat size, sixteenths= edge distance, in.= average yield stress for beam and end-plate mate-

rials, ksi= 0.75 of end-plate material), ksi= ASD allowable tensile stress for bolt material, ksi= specified minimum tensile strength for bolt mate-

rial, ksi= capacity of unstiffened column flange to resist

applied force, kips= factored beam flange force, kips= specified minimum tensile strength, ksi= specified yield stress of the end-plate material,

ksi= specified yield stress of column material, ksi

= horizontal spacing between vertical bolt lines, in.= distance from outer face of flange to web toe of

fillet, in.= the column section distance, in.= factored effective end-plate moment, in.-kips= factored beam moment, in.-kips= connected beam factored end moments, ft-kips= required number of bolts to resist beam shear= effective pitch, in.= pitch, distance from center line of bolt to nearer

surface of the tension flange, in. + ½ in.is generally enough to provide wrench clearance.)

= effective pitch, in.= minimum bolt tension, kips= beam flange thickness, in.= column flange thickness, in.= required column flange thickness, in.= end-plate thickness, in.= required end-plate thickness from stiffness

criterion, in.= required end-plate thickness from strength

criterion, in.= end-plate to beam tension flange stiffener thick-

ness, in.= beam web thickness, in.= column web thickness, in.= column web depth clear of fillets, in.= single shear bolt design strength, kips= factored shear force, kips= leg size of fillet weld or reinforcement weld, in.= required end-plate plastic section modulus, in.3

= resistance factor

39


B.2 LRFD DESIGN AIDS

Table B.1.Design Tension and Single Shear Strengths for A325 and A490 Bolts

(LRFD Method)

a) A325 Bolts

Diameter (in.)

Tension (kips)


Shear A325-N (kips)

Shear A325-X (kips)


20.7

5.22

10.8

14.4

19

29.8

7.51

15.5

20.7

28

40.6

10.2

21.1

28.1

39

53.0

13.4

27.6

36.8

51

67.1

16.9

34.9

46.5

56

82.8

20.9

43.1

57.4

71

100.2

25.2

52.1

69.5

85

119.3

30.0

62.0

82.7

103

b) A490 Bolts

Diameter (in.)

Tension (kips)


Shear A490-N (kips)

Shear A490-X (kips)


25.9

6.44

13.5

17.9

24

37.3

9.28

19.4

25.8

35

50.7

12.6

26.4

35.2

49

66.3

16.5

34.5

45.9

64

83.9

20.9

43.6

58.2

80

103.5

25.8

53.8

71.8

102

125.3

31.2

65.1

86.9

121

149.1

37.1

77.5

103.4

148

All values from AISC LRFD Manual (1986).

Table B.2.LRFD Values of for A325 Bolts

Table B.3.LRFD Values of for A490 Bolts

40


B.3 LRFD QUICK REFERENCE EXAMPLES

EXAMPLE B.1 (Same as Example 3.3)

4-bolt Unstiffened End-Plate

41




i. Tension:

ii. Shear,


i. Bending, Equation 3.1b:


in. Check end-plate shear:






i. Check column web yielding, inequality 3.3b,

ii. Check column web buckling, inequality 3.4b,


42

EXAMPLE B.2 (Same as Examples 4.4 and 4.6)8-Bolt Stiffened End-PlateSimplified procedure


Design stiffeners and welds for

Column flange to stiffener weld, E70XX electrodes:

Column web to stiffener weld, E70XX electrodes:

Check shear force in stiffener base metal, A36 steel:

iv. Check column web yielding, inequality 3.6(b):

E. Final details:




i. Tension:

ii. Shear,


i. Bending, Equation 4.4b:





Use full penetration groove weld with reinforcement.


43


iv. Check column web yielding, inequality 3.6b:


D. Final details:



i. Check column web yielding, inequality 3.3b,

ii. Check column web buckling, inequality 3.4b,



DESIGN GUIDE SERIESAmerican Institute of Steel Construction, Inc.One East Wacker Drive, Suite 3100Chicago, Illinois 60601-2001

Pub. No. D804 (5M194)


dg04 - extended end-plate moment connections

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