behavior o welded plate connections in precast concrete panels under simulated seismic loads

8/10/2019 Behavior o Welded Plate Connections in Precast Concrete Panels Under Simulated Seismic Loads

1/13

Christian L. Hofheins, P.E.Engineer

JM Williams and AssociatesSalt Lake City, Utah

Lawrence D. Reaveley,Ph.D., P.E.ProfessorDepartment of Civil &Environmental EngineeringUniversity of UtahSalt Lake City, Utah

Tests were performed on precast wall panelswith typical loose-plate connectors located inthe vertical joint between panels. The tests wereperformed to investigate the performance of theconnectors under simulated seismic loads. In-plane lateral cyclic loads were applied to thewall panels, which applied tension-shear andcompression-shear forces to the loose-plateconnectors. The paper describes the experimental

program and results for the welded plateconnections in ten precast concrete wall panelassemblies. Design assumptions and simplifieddesign models are also examined. The researchshows that the connection possesses little ductilecapacity and, therefore, is not suitable for use inhigh seismic regions (Zones 3 and 4). However,based on the observed failure modes, minormodifications to the connection are suggested thatwill increase the ductility of the connection.

This paper addresses the behavior of a specific loose-

plate welded connector under applied cyclic loading.

This type of connection is widely used in the United

States. Due to the limited number of tests performed, no

specific design parameters were considered in this study.

The objectives of this investigation were to:

(a) Quantify the performance of the connection in terms

of force-deflection and ductility.

(b) Check the validity of design values that are currently

used for loose-plate welded connections in hollow-core

precast concrete wall panel construction.

Behavior of Welded PlateConnections in Precast ConcretePanels Under SimulatedSeismic Loads

122 PCI JOURNAL

Chris P. Pantelides,Ph.D., P.E.

ProfessorDepartment of Civil &

Environmental EngineeringUniversity of Utah

Salt Lake City, Utah


2/13


3/13

124 PCI JOURNAL

precast walls.6The connections were

designed to be ductile, and to be the

major location of inelastic response

of the structure. Vertical joint con-

nections included different designs of

welded loose-plate and bolted ductile

connections. The connections took ad-

vantage of the interaction between the

embed and concrete by incorporating

flexural yield, tension/compression

yield, shear yield and friction sliding

concepts.

The behavior of a six-story precast

concrete office building under mod-

erate seismicity was investigated.7 It

was concluded that uneven shear dis-

tribution in a precast system causes

a high ductility demand in the panel-

to-panel joint connections. The un-

even distribution drives the connec-

tion elements into the inelastic range.

Therefore, connection details that can

be easily replaced should be used in

precast concrete structures.

As part of the PRESSS five-story

precast concrete building test, a struc-tural wall system consisting of pre-

cast concrete panels was tested under

simulated seismic loading.8 The pre-

cast concrete panels were connected to

each other and the foundation by un-

bonded vertical post-tensioning, using

threaded bars. A horizontal connection

across the vertical joint was provided

by stainless-steel energy-dissipating

U-shaped flexure plates, welded to

embed plates in both adjacent wall

panels. In addition to providing energy

dissipation, these plates provided addi-

tional resistance by shear coupling be-

tween the structural walls. The struc-

tural response of the building under

simulated seismic loads was extremely

satisfactory.

The ability of precast double tee

floor diaphragm and wall systems to

perform adequately under in-plane

seismic forces has been studied in

terms of:

(a) The behavior of connections be-

tween double tees.

(b) The analytical modeling of con-

nectors, diaphragm, and wall systems.

(c) The development of design

guidelines for double tee diaphragms

and wall systems.9

It was found that the interaction be-

tween shear and tension forces in a

flange connection between double tees

could be significant. The connectors

ductility should allow the diaphragm

to redistribute the force among indi-

vidual connectors; this ensures that all

connectors reach their full strength.9

In an experimental study of 3/8 in.

(9.52 mm) stud-welded deformed bar

anchors subject to tensile loads, it was

found that a number of specimens

fractured at the weld. Based on the test

results, quality control procedures and

revised settings were recommended

for stud welding of deformed bar an-

chors.10

The strength and ductility of sev-

eral tilt-up concrete wall panel con-

nections were investigated in a se-

ries of monotonic and cyclic tests.11

Most of the connectors tested did not

show sufficient ductility to be used in

areas of high seismicity. Even when a

connection possessed some ductility,

extensive damage to the surrounding

concrete was observed.

Presently, there is no adequate set of

seismic code requirements for the de-

sign of loose-plate connections in hol-low-core precast wall panels. Many of

the loose-plate connections currently

used in construction are proportioned

using design models that are seldom

backed up with test data.

The truss analogy, currently being

used to describe the performance of

the connection under consideration,

leads to a conservative design.

This paper addresses the behavior of

a specific loose-plate welded connec-

tor for hollow-core precast wall panelsunder cyclic loading; this type of con-

nection is widely used in high seismic

regions of the United States.

The primary objective of this re-

search was to quantify the performance

of the connections between precast

concrete panels using loose-plate con-

nectors and to assess the feasibility for

their use in regions of high seismicity.

Due to the limited number of tests

performed, no specific design parame-

ters have been considered in the study.

The assemblies had variations which

commonly occur in practice. These

included the width of the welded plate,

the length of the weld, the vertical

unevenness of the embedded angles

between adjacent panels, and the mis-

alignment of the three wall panels in

the out-of-plane direction. This paper

presents the experimental results, ana-

lytical models of the connections, and

the details of a proposed new welded

connection.

EXPERIMENTAL PROGRAM

Tests were performed by applying a

quasi-static cyclic load to three precast

hollow-core wall panels connected to-

gether with two loose-plate connectors

at each vertical joint. Ten wall panel

assemblies were tested, all using the

same loose-plate welded connection.

Description of Precast

Wall Panel Assemblies

Fig. 2. Embedded angle assembly for welded connection tested.


4/13

July-August 2002 125

Typically, hollow-core precast pan-

els are 8 ft wide, 12 to 24 ft high (2.44

x 3.66 to 7.32 m) and have six hollow

cores as shown in Fig. 1. The overall

thickness of the panels is 8 in. (203

mm). Panels 12 ft (3.66 m) high and 4

ft (1.22 m) wide were used for testing

due to space constraints in the load

frame. Panels 4 ft (1.22 m) wide were

fabricated by cutting an 8 ft (2.44 m)panel in half.

The two center hollow cores of the

8 ft (2.44 m) panels were filled with

concrete. These solid cores were re-

quired to form a pin connection at the

two outside panels at the supports of

the wall assembly. The average 28-day

compressive strength of the concrete

wall panels was found to be 7150 psi

(49 MPa) with a standard deviation of

190 psi (1.3 MPa).

Description ofWelded Connections

Two welded connections were lo-

cated between panel pairs in vertical

joints. Each welded connection com-

prises two embedded angle assem-

blies and a loose plate. Each embed-

ded angle assembly consists of a 11/2x 2 x 1/4in. (38 x 50.8 x 6.4 mm) x 6

in. (152 mm) long angle, with three 3/8

in. (9.5 mm) diameter weldable steel

deformed anchor bars. The bars are 12in. (305 mm) long, and are stud welded

to the back of the angle as shown in

Fig. 2. Fig. 3 shows the details of the

embedded angle assemblies.

Each wall panel assembly consists

of three hollow-core wall panels joined

together with four welded connections.

Two welded connections are placed 3

ft (914 mm) from the top and bottom

of the wall panels in each vertical joint

found in between the wall panels, as

shown in Fig. 4.The width of the loose plate var-

ied in some wall panel assemblies.

Eight assemblies used 3 in. (76 mm)

wide plates, and two assemblies used

2 in. (51 mm) wide plates. Test re-

sults showed that the plate width had

no effect on the maximum force or

displacement sustained by the wall as-

semblies.

The loose plate was 1/4 to3/8 in.

(6.4 to 9.5 mm) thick A36 steel, and

it was welded to the embedded angle

assembly with two 3/16 in. (4.8 mm)

fillet welds that ran along the 5 in.

(127 mm) vertical edge of the plate

as shown in Fig. 5. All welds were

performed by certified welders with an

E70 electrode and a 7018 rod.

Test Setup

A total of ten wall panel assem-

blies were tested in a load frame at

the Structures Laboratory at the Uni-

versity of Utah. A steel belt enclosed

the wall panel assembly and was con-

nected to a hydraulic actuator with a

force link. The panels were welded

together in the vertical position after

being placed in the load frame. The

entire wall assembly was pushed orpulled by a 150 kip (667 kN) hydrau-

lic actuator through the force link and

the steel belt. The steel belt transferred

the force from the hydraulic actuator

to the wall panel assembly without

restraining the panels.

The panels were supported by two

pin connections placed at the two bot-

tom corners of the wall panel assem-

bly as shown in Fig. 4. The pin used

in this connection was a 2 in. (51 mm)

diameter steel rod. The pin supports

Fig. 3. Details of embedded angle assembly.

supported the wall assembly 1.5 in.

(38 mm) above the bottom of the test

frame, making the pins the only sup-

port for the wall assembly. This al-

lowed the walls to rotate at the pins

and transfer the applied cyclical forcebetween the panels in a symmetrical

manner.

A 1.5 in. (38 mm) thick steel plate

was placed under each corner of the

center panel as shown in Fig. 4. These

plates raised the center panel up to

the same height as the outside panels.

This aligned the embedded angles to

facilitate the placement of the welded

plate. A more detailed description of

the loading system and the wall as-

sembly supports can be found in otherpublications from the University of

Utah.12,13

Test Procedure andInstrumentation

A force was applied to the top left

corner of the wall assembly with a

hydraulic actuator in a quasi-static

manner. The test was carried out in a

force-controlled mode at a rate of ap-

proximately 1 kip (4.5 kN) per second.

Loading steps began at 10 kips (44.5

Note: 1 in. = 25.4 mm.


5/13

126 PCI JOURNAL

kN) and increased by 5 kips (22.2 kN)

until the welded connections failed.

Each loading step consisted of three

cyclic load increments to simulate the

effects of an earthquake. Strain gauges

the wall panel assembly (see Fig. 4).

EXPERIMENTAL RESULTS

The tests revealed the following

characteristics for the connection stud-

ied in this research:

(a) The connection can resist rela-

tively high shear loads.

(b) The connection possesses little

ductile capacity.

(c) The connection should be de-

signed as elastic due to insufficient

ductility.

Failure Mechanism

Cracking around the connections

began near the 20 kip (89 kN) load

cycle. Cracking was initiated by the

embedded angle pushing into the sur-

face of the concrete. As soon as the

concrete crumbled away from around

the connection (see Fig. 6), the de-

formed anchor bars on the back of the

embedded angle assemblies quickly

tore away from their welds. Figs. 6(a)

Fig. 4. Setup and instrumentation of typical wall assembly. Note: 1 in. = 25.4 mm.

Fig. 5. Detailsof welded

loose-plateconnection.

Note: 1 in. =25.4 mm.

were placed on welded plates to form

a three-element rectangular rosette.

Displacement transducers were used

in all of the tests to measure the dis-

placements at various locations of


6/13


and 6(b) illustrate the typical failed

connections.

The following is a description of the

typical mode of failure for this con-

nection:

(a) The concrete around the embed-

ded connections begins to crack.

(b) The bearing capacity of the de-

formed anchor bars and embedded

angle is severely decreased.

(c) The deformed anchor bars

quickly tear free from the embeddedangles as soon as the concrete crum-

bles around the embedded angle as-

semblies.

(d) The load carrying capacity of the

connection is lost.

The welds connecting the loose-

plate to the embedded angle assem-

blies for nine of the ten wall assem-

blies were not damaged. A weld in

one wall panel assembly failed due to

poor penetration of the weld onto the

connecting plate. In general, the welddid not contribute to the failure of the

connection.

Vertical displacement transducers

DT2 and DT3 (see Fig. 4) recorded

very small relative movement between

panels, until the connections failed.

Therefore, the wall assembly moved

as a relatively rigid body until the first

connection failed.

Force-Displacement Relationship of

Wall Panel Assemblies

The hysteretic behavior of Assem-

bly 8 is typical of all wall assemblies

and is shown in Fig. 7. The shape of

the hysteresis loops demonstrates that

they were stable and did not degrade

until sudden failure. The assembly al-

lowed a displacement drift of only 0.5

percent, and did not demonstrate any

appreciable ductile behavior.

The hysteresis envelope for every

wall assembly was approximated by

a general component behavior curveas described in FEMA 273.14The gen-

eral component behavior curve for the

Fig. 6. Welded connections for Assembly 4 at failure: (a) top right connection, and (b) bottom right connection.

Fig. 7. Hysteresis curve for Assembly 8. Note: 1 kip = 4.448 kN; 1 in. = 25.4 mm.

ten assemblies tested is shown in Fig.

8. The general component behavior

curve is able to define the hysteresis

curves into important design criteria.

As defined by FEMA 273, QCE is

the expected strength of the welded

connection of the wall section, and

QCLis the lower-bound estimate of the

strength. Table 1 contains a summary

of the test data that was used to create

the general component behavior curve

of every wall assembly.The mean elastic force, QCL, equals

28.4 kips (126.3 kN), and the mean

(a) (b)


7/13

128 PCI JOURNAL

Table 1. Summary of test results for wall assemblies.

Wall Elastic force,QCL Elastic displacement Ultimate force,QCE Ultimate displacement

assembly (kips) (kN) (in.) (mm) (kips) (kN) (in.) (mm)

1 26.3 117.0 0.44 11.2 28.1 125.0 0.53 13.5

2 24.8 110.3 0.52 13.2 28.8 128.1 0.71 18.0

3 27.1 120.5 0.51 12.9 30.2 134.3 0.62 15.7

4 31.0 137.9 0.63 16.0 35.0 155.7 0.74 18.8

5 32.3 143.7 0.57 14.5 35.0 155.7 0.79 20.1

6 30.2 134.3 0.54 13.7 33.1 147.2 0.71 18.0

7 30.5 135.7 0.55 14.0 34.5 153.5 0.62 15.7

8 23.2 103.2 0.57 14.8 28.2 125.4 0.70 17.8 9 29.9 133.0 0.69 17.5 33.2 147.7 0.80 20.3

10 28.3 125.9 0.40 10.2 30.7 136.6 0.56 14.2

ultimate force, QCE (mean value of

peak on all hysteresis), equals 31.7

kips (141.0 kN). The mean elastic dis-

placement is 0.54 in. (13.7 mm) and

the mean ultimate displacement is 0.68

in. (17.3 mm). Thus, the range for the

inelastic displacement was only 0.14

in. (3.6 mm). According to FEMA

273, the wall panel assembly would

be defined as a force-controlled action

due to the small plastic range.

Strain gauges oriented in a three-

element rosette pattern were applied

to several welded plates on the wall

panel assemblies as shown in Fig. 9.

This rosette pattern was chosen so that

the principal stresses and their direc-

tions could be determined. There was

insufficient instrumentation to deter-

mine the force in each plate directly

from the strain gauges.Fig. 10 shows the principal strains

recorded by the three-element rosette

on the plate of the bottom right con-

nection of Assembly 2. Although the

plate yielded in the last loading cycle,

the connection failed immediately

thereafter. As a result, the ductility of

the connection was not significantly

increased by the yielded plate.

ANALYTICAL RESULTS

Fig. 8. General component behavior of ten wall assemblies. Note: 1 kip = 4.448 kN;1 in. = 25.4 mm.

Fig. 9. Three-element strain gauge

rosette appliedon loose-plate

connector.


8/13


A structural analysis of the wall as-

sembly was performed using the struc-

tural analysis program SAP 2000.15

The purpose of the analysis was to find

the forces across each welded connec-

tion of the wall panel assembly, and

compare them to the commonly used

design methodologies. The precast

concrete wall panels were modeled

as rigid frame elements with a dia-phragm constraint on each wall panel

(as shown in Fig. 11). The wall panel

connections were modeled as rigid

pins, which is a reasonable assumption

given their brittle mode of failure.

The nodes located at the supports of

the wall panel assembly were assigned

pin restraints. The shim supports under

the center panel (see Fig. 4) were

not considered in the model. Verti-

cal displacement transducers revealed

that the center panel rose vertically,whether the wall assembly was being

pushed or pulled. These displacements

were a result of vertical movement

occurring at the pin supports, and the

rigid body motion of the wall panel

assembly.

The holes in the panels for the pin

supports were oversized for ease of

erection in the load frame. The over-

sized holes allowed the entire assem-

bly to rise and move as a rigid body.

As a result, the bottom corners of the

middle panel never touched the shims

during loading cycles. Consequently,

the shim supports did not restrain the

panel assembly, and were not included

in the model.

The weight of each wall panel was

applied as a point load at four differ-

Fig. 10. Principal strains at bottom right plate connection of Assembly 2.

Fig. 11. Structuralanalysis modelof wall panelassembly. Note:1 kip = 4.448 kN.


9/13

130 PCI JOURNAL

ent nodes on each wall panel (see Fig.

11). The average maximum force at

failure, 31.7 kips (141.0 kN), was ap-

plied as the lateral load at the top left

corner of the wall panel assembly to

find the capacity of each welded con-

nection. The structural analysis results

are shown in Fig. 12.

For the above conditions, the shearforce at failure of the welded con-

nections was 15.0 kips (66.7 kN) on

the two left connectors, and 16.6 kips

(73.8 kN) on the two right connectors.

This is significant because the capac-

ity of this connection typically used

in design is equal to 8 kips (35.6 kN).

Structures built with these welded con-

nections were safely designed with an

approximate factor of safety of 1.9.Using this design value, the connection

will safely stay in the elastic range.

Force-Displacement Relationshipof Welded Connection

The force-displacement relation-

ship of each welded connection was

found by plotting the relative vertical

displacement of two adjoining wallpanels versus the shear force across

the welded connection. The relative

displacement of two adjoining wall

panels in the vertical direction was

found by subtracting data retrieved

from displacement transducers DT2

and DT3 (see Fig. 4).

The shear force across each connec-

tion was found as follows: the force

applied by the hydraulic actuator on

the wall assembly was multiplied by

the ratio of the average maximum

force at failure of the welded connec-

tion, or 16.6 kips (73.8 kN), to the

average maximum force at failure of

the wall panel assembly, or 31.7 kips

(141.0 kN). This assumption is reason-

able because the connections behave

in a linear elastic manner.

The hysteresis curve for the connec-

tors of eight wall panel assemblies,

was approximated by a general com-

ponent behavior curve as described in

the Guidelines for the Seismic Reha-

bilitation of Buildings, FEMA 273.14

Fig. 12. Resultsof structuralanalysis for

maximum lateralload applied

to the wallassembly.

Note: 1 kip =4.448 kN.

Fig. 13. General component behavior of the welded connectors of eight wallassemblies. Note: 1 kip = 4.448 kN; 1 in. = 25.4 mm.


10/13


Fig. 13 shows the general component

behavior curve for the connectors of

eight wall assemblies. The average

elastic force on the connectors was

14.7 kips (65.4 kN), and the average

ultimate force was 17.1 kips (76.1

kN).

The force-displacement relationship

is linear until the connection fails in a

brittle manner. This connection shouldbe designed to remain elastic due to

its brittle mode of failure and limited

ductility.

Analytical Model ofWelded Connection

The probable resisting mechanisms

of the connector under consideration

are bearing and tension actions in the

deformed anchor bars, as well as bear-

ing of the angle section. Many de-

signers currently model this welded

connection with the truss analogy as

described in the PCI Design Hand-

book.16

Fig. 14 is an illustration of the truss

analogy. The following equations are

used to describe this model:

CU= TU= Asfy (1)

VRU= (CU+ TU)cos (2)

where

CU = compression force

TU = tensile force

= capacity reduction factor =

0.9

= angle of deformed anchor

bar = 45 degrees

As = area of 3/8 in. (9.5 mm) di-

ameter deformed anchor bar

= 0.13 sq in.(71 mm2)

fy = yield strength of mild steel

reinforcement [= 60 ksi (420

MPa)]

VRU = vertical shear force resisted

by connection

The equations from the truss anal-

ogy yield a vertical shear resistance

of 8.4 kips (37.4 kN) for each connec-

tion. The analysis indicates that the

average capacity of this connection is

between 15.0 and 16.6 kips (66.7 to

73.8 kN). The truss analogy is a con-

servative design methodology when

applied to this connection.

The following is a list of some of

the differences between the truss anal-

ogy and the connection under consid-

eration:

a. The angle for this connection

equals zero, not 45 degrees (see Fig.

2).

b. The deformed anchor bars are

bent 90 degrees into the back of the

angle (see Figs. 2 and 3). The bars

will not be able to develop the full

tensile capacity as described in the

truss analogy. The deformed anchor

bars act more as 3/8in. (9.5 mm) studs

with ineffective tails rather than bars

in tension.

c. The truss analogy does not ac-

count for the bearing of the angle as-

sembly into the concrete. Angle bear-

ing is one of the main force resisting

mechanisms of the connection.

Fig. 15 illustrates that the deformed

Fig. 14. Truss analogy model for welded connection.

Fig. 15. Statics ofdeformed anchorbar at failure forcurrent connection.Note: 1kip = 4.448kN, 1 in. = 25.4mm; 1 k-in. = 133

N-m.


11/13

132 PCI JOURNAL

anchor bar cannot fully develop in ten-

sion due to the eccentric load from the

bend in the bar. Assuming the force

taken by each vertical deformed an-

chor bar is 8.3 kips (36.9 kN) (half of

the total vertical shear force taken by

the connection), the maximum shear

and moment taken by each vertical de-

formed anchor bar is 8.3 kips and 10.4

kip-in. (36.9 kN and 1.17 kN-m), re-spectively. The eccentric load causes

the deformed anchor bars to quickly

tear free from their welds as soon as

the concrete crushes around the con-

nection.

PROPOSED NEW WELDEDCONNECTION

The most effective way to improve

this connection is to provide a larger

surface area for concrete bearing andto minimize eccentric loads from the

deformed anchor bars. Fig. 16 is a

drawing of a proposed new embedded

angle assembly. The angle is replaced

by a 6 in. (152 mm) long ST2x3.85

to create a greater bearing area in the

concrete.

One continuous deformed anchor

bar replaces the two vertical deformed

anchor bars of the previous connec-

tion. The vertical deformed anchor

bar is attached to the back of the em-

bedded angle assembly with a 4 in.

(102 mm) long, 3/16in. (4.8 mm) fillet

weld. The vertical deformed anchor

bar is bent at 5 degrees to minimize

eccentric loads and to ensure adequate

concrete cover.

The strength of this fillet weld

can be described by Eq. (3), and the

strength of the base metal can be de-

scribed by Eq. (4), as:17

Rn= 0.75te(0.6Fexx) (3)

Rn= 0.75t(0.6Fu) (4)

where

Rn = strength of fillet weld or base

material

Fexx = strength of electrode = 70 ksi

(483 MPa)

Fu = tensile strength of base mate-

rial = 60 ksi (420 MPa)

te = 0.707a

a = weld size = 3/16in. (4.8 mm)

t = thickness of base material =5/16in. (7.9 mm)

Eq. (3) yields the strength of the fil-

let weld as 4.2 kips per in. (0.74 kN/

mm), and Eq. (4) yields the strength

of the base material as 8.4 kips per in.

(1.47 kN/mm). A 4 in. (102 mm) long

weld gives a strength of 16.8 kips (74.7

kN), which is significantly higher than

the allowable shear resistance of the

welds in the tested connection. In addi-

tion, the concrete will not easily break

away from the connection due to the

increased bearing area with the web

of the structural tee embedded into the

wall.

DISCUSSION OF

TEST RESULTS

Engineers prefer the panel connec-tions, not the panels themselves, to

be the weak link in the system. This

investigation has shown that the con-

nections are in fact the weakest link.

Although the loose-plate connection

used in this research effectively trans-

ferred the applied shear forces, the

connection failed in a brittle manner.

The small displacement ductility

exhibited by the welded connections

is lost as soon as the deformed an-

chor bars on the back of the embeddedangle fracture from their welds. Fail-

ure occurs before shear yielding can

take place in the welded plate.

These tests reveal that hollow-core

precast concrete panels can be used in

seismic regions provided that the con-

nections can be improved. To this end,

a new welded connection is proposed;

ductility may be restored to the sys-

tem by increasing the surface area for

concrete bearing and by reducing the

eccentric load in the deformed anchor

bars.

If the connection is a location of

ductile inelastic deformation, the pre-

cast concrete panels will remain elastic

under seismic response. Damage to

the overall structure will be reduced

and repair of the structure will be less

costly. Ductility in shear will allow the

force to redistribute among individualconnectors. Ductility will enable all

connectors to reach their full strength,

thereby increasing the overall force re-

sisting capability of the structure.

For existing connections of the type

tested in this investigation, a seismic

retrofit option has been studied using

a carbon fiber composite connection,

which will be published shortly.

CONCLUSIONS

Simulated seismic load tests of

Fig. 16. Details of proposed new embedded angle assembly. Note: 1 in. = 25.4 mm.


12/13


loose-plate vertical connections be-

tween precast concrete wall panels

were performed. Based on the results

of this investigation, the following

conclusions can be drawn:

1. The loose-plate connection com-

monly used in precast construction can

resist relatively high shear forces.

2. The connection fails in a brittle

manner when the deformed anchorbars tear free from the embedded an-

gles, which occurs as soon as the con-

crete crumbles around the embedded

angle assemblies; as a consequence,

the connection possesses little ductile

capacity.

3. The connection should be de-

signed to remain elastic; in its current

form, the connection is not suitable for

use in areas of high seismic regions

(Zones 3 and 4).

4.The design methodologies com-

monly used for this connection are

conservative.

5. The connection can be modified

to increase its ductile behavior by pro-

viding more surface area for concretebearing, and by minimizing eccentric

loads in the deformed anchor bars.

ACKNOWLEDGMENT

The authors would like to acknowl-

edge the funding provided by XXsys

Technologies, Inc., and the Center

for Composites in Construction at the

University of Utah.

The authors wish to express their

gratitude to Eagle Precast Company

(Monroc, Inc.), for providing the pre-

cast wall specimens.

The authors would like to thank

Vladimir Volnyy and Professor Janos

Gergely for their assistance with thetests. In addition, the authors are

grateful to Philip Richardson and Carl

Wright of Eagle Precast Company for

their suggestions.

Lastly, the authors want to express

their appreciation to the PCI JOUR-

NAL reviewers for their thoughtful

and constructive comments.1. Rostasy, F. S., Connections in Precast Concrete Structures

Continuity in Double-T Floor Construction, PCI JOURNAL,

V. 7, No. 4, 1962, pp. 18-48.

2. Scoggin, H. L., and Pfeiffer, D. W., Cast-in-Place Concrete

Residences with Insulated Walls-Influence of Shear Connec-

tors on Flexural Resistance,Journal of the PCA Research and

Development Laboratories, V. 9, No. 2, 1967, pp. 2-7.

3. Abdul-Wahab, H. M. S., Ultimate Shear Strength of Vertical

Joints in Panel Structures,ACI Structural Journal, V. 88, No.

2, March-April 1991, pp. 204-213.

4. Spencer, R. A., and Neille, D. S., Cyclic Tests of Welded

Headed Stud Connections, PCI JOURNAL, V. 21, No. 3,

May-June 1976, pp. 70-81.

5. Stanton, J. F., Hawkins, N. M., and Hicks, T. R., PRESSS

Project 1.3: Connection Classification and Evaluation, PCIJOURNAL, V. 36, No. 5, September-October 1991, pp. 62-71.

6. Schultz, A., Tadros, M. K., Juo, X. M., and Magana, R. A.,

Seismic Resistance of Vertical Joints in Precast Shear Walls,

Proceedings, XII FIP Congress, Washington, DC., May 29 -

June 2, 1994.

7. Low, S.-G., Behavior of a Six-Story Office Building Under

Moderate Seismicity, University of Nebraska, Lincoln, NE,

May 1995.

8. Priestley, M. J. N., Sritharan, S., Conley, J. R., and Pampanin,

S., Preliminary Results and Conclusions from the PRESSS

Five-Story Precast Concrete Test Building, PCI JOURNAL,

V. 44, No. 6, November-December 1999, pp. 42-67.

9. Pincheira, J. A., Oliva, M. G., and Kusumo-Rahardjo, F. I.,Tests on Double-Tee Flange Connectors Subjected to Mono-

tonic and Cyclic Loading, Research Report, University of

Wisconsin, Madison, WI, 1998.

10. Strigel, R. M., Pincheira, J. A., and Oliva, M. G., Reliability

of 3/8 in. Stud-Welded Deformed Bar Anchors Subject to Ten-

sile Loads, PCI JOURNAL, V. 45, No. 6, November-Decem-

ber 2000, pp. 72-82.

11. Lemieux, K., Sexsmith, R., and Weiler, G., Behavior of Em-

bedded Steel Connectors in Concrete Tilt-Up Panels, ACI

Structural Journal, V. 95, No. 4, July-August 1998, pp. 400-

413.

12. Pantelides, C. P., Reaveley, L. D., Gergely, I., Hofheins, C.,

and Volnyy, V., Testing of Precast Wall Connections, Uni-

versity of Utah, Department of Civil and Environmental En-

gineering, Report UUCVEEN 97-02, 97-03, 98-01, Salt Lake

City, UT, 1997-98.

13. Hofheins, C., Welded Loose-Plate Connections for Hollow-Core Precast Wall Panels, M.Sc. Thesis, Department of Civil

& Environmental Engineering, University of Utah, Salt Lake

City, UT, May 1999.

14. Building Seismic Safety Council, NEHRP Guidelines for the

Seismic Rehabilitation of Buildings, FEMA Publication 273,

Federal Emergency Management Agency, Washington, DC,

October 1997.

15. SAP2000 Analysis Reference, Computers and Structures, Inc.,

V. I, Berkeley, CA, 1997.

16. PCI Committee on Industry Handbook, PCI Design Hand-

book: Precast and Prestressed Concrete, Fifth Edition, Pre-

cast/Prestressed Concrete Institute, Chicago, IL, 1999.

17. Salmon, C. G., and Johnson, J. E., Steel Structures Design andBehavior, Fourth Edition, Harper Collins College Publishers

REFERENCES

APPENDIX A NOTATIONInc., New York, NY, 1996.

Ab = area of reinforcing bar

As = area of deformed anchor bar

CU = compression force

Fexx = strength of electrode

fs = steel stress

Fu = tensile strength of base material

fy

= yield stress of reinforcement

n = number of reinforcing bars

QCE= expected strength

QCL= lower-bound strength

Rn = strength of fillet weld or base material

t = thickness of base material

te = effective area of weld

TU = tensile force

VRU = vertical shear force resisted by connection

Vs = shear strength of connection = angle of deformed anchor bar


13/13

= capacity reduction factor

behavior o welded plate connections in precast concrete panels under simulated seismic loads

Documents