the changing strength of clay and its application to ...€¦ · the changing strength of clay and...

THE CHANGING STRENGTH OF CLAY AND ITS

APPLICATION TO OFFSHORE PIPELINE DESIGN

By

Fauzan SAHDI

B.Eng. (Hons)

This thesis is presented for the degree of

Doctor of Philosophy

at

School of Civil and Resource Engineering

Centre for Offshore Foundation Systems

March 2013

DECLARATION

This thesis contains published work, which has been co-authored. The bibliographical

details of the work and where it appears in the thesis are outlined below:

1. Sahdi, F., Boylan, N., White, D.J., and Gaudin, C. 2010. The influence of coloured

dyes on the undrained shear strength of kaolin. In 7th International Conference on

Physical Modelling in Geotechnics, Zurich, Switzerland. (Chapter 2)

The estimated percentage contribution of the candidate is 80%.

2. Sahdi, F., Gaudin, C., and White, D.J. 2013. The strength properties of ultra-soft

kaolin. Canadian Geotechnical Journal (Submitted: March 2013). (Chapter 5)


3. Sahdi, F., Gaudin, C., White, D.J., Boylan, N., and Randolph, M.F. 2012. Centrifuge

modelling of active slide-pipeline loading in soft clay. Géotechnique (Submitted:

December 2012). (Chapter 6)


That man can have nothing but what he strives for

– Quran 53:39

Acquire knowledge and impart it to the people

– Muhammad (peace be upon him)

i

ABSTRACT

Offshore pipelines form an integral part of any offshore hydrocarbon production field

and ensuring the security of these vital ‘arteries’ from either environmental or

operational hazards are of paramount importance.

This thesis is concerned with the changing strength of soft seabed sediments through the

processes of consolidation and disturbance, as these sediments form and interact with

seabed infrastructure. These processes involve failure and remoulding, as well as

healing through reconsolidation. Therefore, it is necessary to consider alternating

episodes of cyclic loading and recovery in cases involving storms and the intervening

calm periods, pipeline startup and shutdown episodes, and the triggering, runout and

reconsolidation of mass movement events – including their impact and flow past

pipelines. An important theme is the whole life response of a sediment during genesis

and when acted on by seabed infrastructure. Rate effects are particularly important

considering the wide range of strain rates involved in these processes – from slow

changes during consolidation, to the rapid strain rates within submarine slides.

The effect of clay microfabric arrangement during monotonic and subsequent cyclic

strength degradation is explored using centrifuge cyclic T-bar tests. The different

arrangements of clay microfabric were induced by adding synthetic dyes in kaolin

samples. SEM (scanning electron microscope) and XRD (X-ray diffraction) analyses

were used to correlate the increase in initial strengths to the arrangement of kaolin

platelets, as reflected by the XRD Fabric Index. However, the remoulded shear

strengths for the samples with different initial fabric arrangements were found to be

identical.

The development of an innovative soil characterisation tool, namely the vertically

orientated penetrometer (VOP) is described in detail. The VOP was developed with the

aim of providing accurate surficial strength data via a sensing element that is analogous

to a rigid pile under horizontal displacement. Extensive centrifuge tests were performed

to study the VOP behaviour during monotonic and cyclic horizontal translation in

normally consolidated (NC) and highly overconsolidated (HOC) clays. Analyses of the

test results demonstrated that the concepts normally used to estimate the ultimate

resistance of a laterally loaded pile can also be used in reverse to infer the shear strength

ii

of the soil. It is shown that the VOP could potentially measure the true intact soil

strength directly, as opposed to the strain-softened strength normally obtained in

conventional full-flow penetrometer tests. Cyclic VOP test results revealed that it has

the same capability to provide soil sensitivity measurements as other full-flow

penetrometers. A parametric study on VOP viscous rate effects confirmed that the rate

dependency is similar in intact and remoulded samples and independent of the water

content of soil. The VOP concept is potentially applicable in the field, perhaps to

provide continuous strength data over long horizontal regions – e.g. for pipeline route

surveys.

A simplified macroelement effective stress framework is developed to predict the

remoulding and reconsolidation behaviour of clay. The framework is compared to

centrifuge VOP cyclic experimental data. The reported experiments demonstrated that

even within a cyclic shearing episode, reconsolidation, which leads to strength recovery,

can occur. The framework captures the measured initial degradation and subsequent rise

in VOP resistance accurately, demonstrating the framework applicability to quantify the

changes in the surrounding soil operative strength throughout the operational lifetime of

a pipeline.

Having used a VOP penetrometer to identify the effects of rate, remoulding and

reconsolidation on the strength of normally consolidated and overconsolidated kaolin

clay, a further study was performed on underconsolidated kaolin clay to examine

whether these relationships apply through to the fluid regime, when the clay has a far

lower strength than is usually considered in geotechnical engineering.

Drum centrifuge samples under self-weight consolidation were characterised by

numerical large strain consolidation analysis, and were used to assess the changing

strength and strength degradation properties of kaolin from strengths of less than 0.1

kPa and stronger. A simple framework of correlations showed how the strength

degradation properties measured from the cyclic T-bar tests evolved during

consolidation, and with liquidity index. Encouraging agreement was found with other

published data derived at higher strengths (and lower liquidity indices). These

correlations provide a simple basis for incorporating strength degradation behaviour into

the analyses of various soil-pipeline interactions at high moisture contents, across the

solid-fluid boundary.

iii

The behaviour of pipelines under submarine slide impact was explored in an extensive

drum centrifuge test programme. These centrifuge pipe tests varied the relative

influence of ‘fluid’ drag and ‘solid’ bearing resistance by covering 4 orders of

magnitude of pipe velocity (from 0.004 – 4.2 m/s) and two orders of magnitude of soil

strength (from 0.08 to 1.7 kPa). From the extensive experimental data, a hybrid viscous-

inertial type framework is proposed to estimate the horizontal impact forces of a

submarine slide on a stationary pipeline. The framework is validated by reinterpretation

of other physical tests data reported in the literature, using the hybrid approach.

Total pore pressure data gathered from the pipe test programme suggests the presence of

vortex shedding at high non-Newtonian Reynolds number. The presence of this

phenomenon highlights the importance of quantifying the resulting vertical lift forces.

Reinterpretation of vortex shedding induced lift force data reported in the literature

showed that the hybrid approach for quantifying the horizontal slide loading is equally

applicable in estimating this vertical force component.

In summary, this study has examined the strength of clay across a wider range of strain

rates and moisture contents than has been considered previously, providing a validated

framework for considering the effects of remoulding, reconsolidation and viscous-

inertial drag on the resulting soil-pipeline interaction forces.

v

ACKNOWLEDGEMENTS

Praise and thanks be to Allah, Most Gracious, Most Merciful, for giving me the strength

and perseverance to complete this thesis. May Allah accept this humble effort of mine.

I would sincerely like to thank Professor David White for his generous and invaluable

support and countless discussions throughout my PhD years. His approach in always

using positive and encouraging words is truly inspirational. Special mention goes to Dr.

Noel Boylan for his friendly and encouraging attitude particularly in helping me with

centrifuge tests. I would also like to extend my deepest gratitude to Professor

Christophe Gaudin who effectively became my co-supervisor for the last 2 years. His

readiness to give excellent research supervision anytime and anywhere is much

appreciated.

In addition, I would like to thank Professor Mark Randolph for giving me the

opportunity to conduct research at COFS and for the occasional discussions. Many

thanks to Professor Martin Fahey for his help with MinTaCo.

I appreciate the financial support from the Ministry of Higher Education Malaysia

(MOHE), UNIMAS and COFS for me to pursue a PhD study at UWA.

I am indebted to the people from my former workplace Jurutera Jasa, Dr. Dominic Ong,

Mr. Tai Lee Yon, Dr. Ting Wen Hui and Mr. Anthony Law for paving my way to

COFS.

It is truly a privilege to work alongside technicians Bart (for the good times we had

cleaning the centrifuge and mixing clay), Shane (a.k.a ‘PID/LabView guru’) – thanks

for all your help with data acquisition, Phil (indeed, you are the ‘Majorstrain’), Tuarn

(for the hardware installations), Dave (for the beautiful ‘metal sculptures’), Khin,

Claire, Binaya and Don. It wouldn’t be possible to generate a single graph presented in

this thesis if it wasn’t for their expert help in centrifuge and lab tests and equipment

developments. Many thanks to Dr. Jeremy, Peter, Lyn and John for helping me with the

SEM and also Dr. Saowanuch, Dr. Naoko and Michael for their help with the XRD. The

vi

help of the COFS’s admin and IT staffs Eileen, Monica, Lisa, Ivan, Keith, Kan Yu, and

Wenge is much appreciated.

To my friends Bassem and Azrul, I must say, I much enjoyed the relaxing ‘Asar prayer

walk’ we had every afternoon and I thank you for that. To the ‘badminton gang’ Jit

Kheng, Indranil, Yusuke, Xi Hong, Stefanus and Xiaojun, thank you for helping me

find an excuse to leave office early for the much needed weekly stress relief badminton

session. Special thanks to Han Eng, KK Lee, Dr. James Schneider and Dr. Norhisham

Bakhary for all the help when I first arrived to COFS. Friends and colleagues David

Yong, Ariel, Vickie, Kar Lu, Amin, Youhu, Divya, Santiram, Zack, Neyamat, Dr. Al

Sidqi, Suradi, Dr. Tutun, thank you for the helpful discussions and friendship all these

years.

For always providing me with support, love and encouragement, I would like to thank

my wife Hafizah. Truly, you are my better half, and your great patience in putting up

with my swinging mood and taking care of little cheeky Ayesha (who in her own way

provided me with bundles of joy) is simply incredible. My deepest gratitude goes to my

parents and sister for their unconditional love and selfless support. I can only hope to

return the favour by being a good son and brother and I hope I have made you proud.

vii

CONTENTS

ABSTRACT ....................................................................................................................... i

ACKNOWLEDGEMENTS ............................................................................................. v

CONTENTS .................................................................................................................... vii

DECLARATION OF CANDIDATE CONTRIBUTIONS…………………………xi

CHAPTER 1. INTRODUCTION ................................................................................ 1.1

1.1 Overview .......................................................................................................... 1.1

1.2 Research objectives .......................................................................................... 1.2

1.2.1 Rate effects ...................................................................................... 1.2

1.2.2 Effects of cyclic loading.................................................................. 1.4

1.3 Thesis outline ................................................................................................... 1.6

CHAPTER 2. THE INFLUENCE OF COLOURED DYES ON THE UNDRAINED

SHEAR STRENGTH OF KAOLIN ............................................................................ 2.1

2.1 Abstract ............................................................................................................ 2.2

2.2 Introduction ...................................................................................................... 2.2

2.3 Sample preparation .......................................................................................... 2.3

2.3.1 Samples for strength characterisation ............................................. 2.3

2.3.2 Samples for SEM, XRD, index and pH tests .................................. 2.4

2.4 Results .............................................................................................................. 2.5

2.4.1 Index properties and pH .................................................................. 2.5

2.4.2 T-bar penetrometer testing .............................................................. 2.5

2.4.3 SEM and XRD analyses .................................................................. 2.7

2.5 Discussion ........................................................................................................ 2.9

2.6 Conclusions .................................................................................................... 2.11

2.7 Acknowledgements ........................................................................................ 2.12

CHAPTER 3. DEVELOPMENT OF THE VERTICALLY ORIENTED

PENETROMETER FOR SHALLOW SEABED CHARACTERISATION ........... 3.1

3.1 Abstract ............................................................................................................ 3.2

3.2 Introduction ...................................................................................................... 3.2

3.3 Background literature ....................................................................................... 3.4

3.3.1 Laterally loaded pile ........................................................................ 3.4

3.3.2 Strain rate effects............................................................................. 3.7

3.4 Experimental programme ................................................................................. 3.8

3.4.1 Experimental apparatus ................................................................... 3.8

3.4.2 Sample preparation.......................................................................... 3.9

viii

3.4.3 Centrifuge testing procedure ........................................................... 3.9

3.5 T-bar tests: shear strength and sensitivity ...................................................... 3.11

3.6 Results for VOP experiments ......................................................................... 3.12

3.6.1 Bearing capacity factors – NC sample .......................................... 3.13

3.6.2 Bearing capacity factors – HOC sample ....................................... 3.15

3.6.3 Soil sensitivity ............................................................................... 3.16

3.6.4 Strain rate effects .......................................................................... 3.18

3.6.5 Peak load in NC soil accounting for softening and rate effects .... 3.19

3.7 Possible refinements of the VOP ................................................................... 3.20

3.7.1 Back-analysis and bearing factors ................................................. 3.20

3.7.2 Instrumentation ............................................................................. 3.22

3.8 Conclusions .................................................................................................... 3.22


CHAPTER 4. EFFECTS OF CYCLIC DISTURBANCE AND

RECONSOLIDATION ON THE SHEAR STRENGTH OF CLAY ....................... 4.1

4.1 Abstract ............................................................................................................ 4.2

4.2 Introduction ...................................................................................................... 4.3

4.3 Test methodology ............................................................................................. 4.5

4.3.1 Experimental equipment ................................................................. 4.5

4.3.2 Clay properties and sample preparation .......................................... 4.6

4.3.3 Test procedure ................................................................................. 4.6

4.4 Undrained shear strength and sensitivity of sample ........................................ 4.7

4.5 Reconsolidation during cyclic VOP test .......................................................... 4.9

4.6 Framework for back analysis ......................................................................... 4.11

4.6.1 Strength analysis considering drainage condition ......................... 4.11

4.6.2 Strength evolution due to repeated disturbance and

reconsolidation ............................................................................................... 4.13

4.6.3 Dissipation of excess pore water pressure .................................... 4.16

4.7 Remoulding-reconsolidation parameters ....................................................... 4.18

4.7.1 Failure state parameters ................................................................ 4.18

4.7.2 Soil sensitivity and degradation rate parameter ............................ 4.19

4.8 Comparison of model with test data .............................................................. 4.19

4.9 Steady state cyclic resistance ......................................................................... 4.22

4.10 Conclusions .................................................................................................... 4.23


CHAPTER 5. THE STRENGTH PROPERTIES OF ULTRA-SOFT KAOLIN

CLAY ............................................................................................................................. 5.1

5.1 Abstract ............................................................................................................ 5.2

5.2 Introduction ...................................................................................................... 5.3

ix

5.2.1 Importance of quantifying strength degradation ............................. 5.3

5.2.2 Consolidation at very low stresses and through large-strains ......... 5.5

5.2.3 Objectives ........................................................................................ 5.6

5.3 Properties of kaolin affecting cyclic T-bar resistance ...................................... 5.6

5.3.1 Overview ......................................................................................... 5.6

5.3.2 Index properties ............................................................................... 5.7

5.3.3 Compressibility and permeability ................................................... 5.7

5.3.4 Undrained strength ratio .................................................................. 5.8

5.4 Test methodology ............................................................................................. 5.8

5.4.1 Overview of test setup ..................................................................... 5.8

5.4.2 Sample preparation.......................................................................... 5.9

5.4.3 Minimising stress error in consolidating soil .................................. 5.9

5.4.4 T-bar tests ...................................................................................... 5.12

5.5 Centrifuge test results..................................................................................... 5.13

5.5.1 Average degree of consolidation, U% .......................................... 5.13

5.5.2 Correction of T-bar resistance using cyclic data ........................... 5.13

5.6 Validation of cyclic correction ....................................................................... 5.15

5.6.1 General numerical validation procedure ....................................... 5.15

5.6.2 Calibration of numerical analysis.................................................. 5.16

5.6.3 Validating calibrated numerical parameters .................................. 5.17

5.6.4 Base excess pore pressure ............................................................. 5.18

5.6.5 Evolution of undrained shear strength and voids ratio ................. 5.18

5.7 Effect of water content on sensitivity, ductility and remoulded strength ...... 5.20

5.7.1 Sensitivity and ductility ................................................................ 5.20

5.7.2 Remoulded shear strength ............................................................. 5.21

5.8 Correlations for strength degradation ............................................................ 5.23

5.9 Performance of correlations ........................................................................... 5.25

5.10 Conclusions .................................................................................................... 5.27


CHAPTER 6. CENTRIFUGE MODELLING OF ACTIVE SLIDE-PIPELINE

LOADING IN SOFT CLAY ........................................................................................ 6.1

6.1 Abstract ............................................................................................................ 6.2

6.2 Introduction ...................................................................................................... 6.2

6.3 Experimental apparatus .................................................................................... 6.6

6.3.1 UWA drum centrifuge..................................................................... 6.6

6.3.2 T-bar penetrometer .......................................................................... 6.7

6.3.3 Model pipeline ................................................................................ 6.7

6.4 Sample preparation .......................................................................................... 6.8

6.5 Test procedure .................................................................................................. 6.9

6.5.1 Investigation of kaolin consolidation behaviour - Sample 1 ......... 6.11

x

6.5.2 Pipe tests in fully consolidated sample - Sample 1 ....................... 6.11

6.5.3 End effects tests (EET) - Sample EET .......................................... 6.12

6.5.4 Tests in samples at U < 100% (Samples 2 –12) ............................ 6.13

6.6 Results: Consolidation behaviour of kaolin ................................................... 6.13

6.6.1 Sample settlement ......................................................................... 6.13

6.6.2 Evolution of shear strength and density with consolidation time . 6.14

6.7 Results: End effects tests (EET) .................................................................... 6.16

6.8 Results: Horizontal force ............................................................................... 6.16

6.8.1 Overview ....................................................................................... 6.16

6.8.2 Viscous effects .............................................................................. 6.17

6.8.3 Inertial drag coefficient (fluid mechanics approach) .................... 6.18

6.8.4 Hybrid approach ............................................................................ 6.18

6.9 Results: Pore water pressure responses .......................................................... 6.21

6.10 Estimation of mean vertical lift force ............................................................ 6.22

6.11 Conclusions .................................................................................................... 6.23


CHAPTER 7. CONCLUDING REMARKS ............................................................... 7.1

7.1 Introduction ...................................................................................................... 7.1

7.2 Principal outcomes: rate effects ....................................................................... 7.1

7.2.1 Viscous effects ................................................................................ 7.2

7.2.2 Combined viscous-inertial effects ................................................... 7.2

7.3 Principal outcomes: effects of cyclic loading and strength recovery .............. 7.3

7.3.1 Effect of soil microstructure on strength degradation ..................... 7.4

7.3.2 Decay of strength from the undisturbed to remoulded states.......... 7.5

7.3.3 Degradation and recovery of undrained shear strength .................. 7.6

7.3.4 Strength degradation-index properties relationship ........................ 7.7

7.4 Overall summary .............................................................................................. 7.8

7.5 Recommendations for future research ............................................................. 7.8

CHAPTER 8. REFERENCES ..................................................................................... 8.1

xi

DECLARATION OF CANDIDATE CONTRIBUTIONS

In accordance to the regulations of the University of Western Australia, this thesis is

submitted as a series of papers. Chapter 2 has been published, while Chapters 5 and 6

have been submitted for journal publications. Chapters 3 and 4 have been prepared as

draft journal papers to be submitted in the near future. The contributions of the

candidate and the co-authors for the papers comprising Chapters 2 to 6 are outlined as

follows:

Paper 1

This paper is presented as Chapter 2 and the estimated contribution of the candidate for

this paper is 80%. This paper was published in the 7th

International Conference on

Physical Modelling in Geotechnics proceedings as:

Sahdi, F., Boylan, N., White, D.J., and Gaudin, C. 2010. The influence of coloured dyes

on the undrained shear strength of kaolin. In 7th International Conference on Physical

Modelling in Geotechnics, Zurich, Switzerland.

The work outlined in this paper includes the usage of the geotechnical centrifuge,

scanning electron microscope (SEM) and X-ray diffraction (XRD) devices as well as

simple index properties tests on the soil samples. All these experimental works were

conducted fully by the candidate with the necessary guidance and training as follows:

Geotechnical centrifuge – The candidate devised the test plan in consultation

with Dr. Noel Boylan, Professor David White and Professor Christophe Gaudin.

The physical tests were performed by the candidate with operational guidance

from Mr. Bart Thompson (drum centrifuge technician).

xii

SEM – The candidate performed all the SEM observations at the Centre for

Microscopy, Characterisation and Analysis, UWA upon receiving proper

training from Mr. Peter Duncan and Dr. Jeremy Shaw.

XRD – Dr. Naoko Zwingman, Dr. Saowanuch Tawornpruek and Mr. Michael

Smirk trained the candidate to operate the XRD. All XRD tests were performed

by the candidate.

Index tests – The candidate conducted all the index tests with some guidance

from Mrs. Claire Bearman and Mr. Binaya Bhattarai.

The candidate interpreted the experimental data and wrote the initial version of the

paper which was proof read by Dr. Noel Boylan, Professor David White and Professor

Christophe Gaudin.

Chapter 3

This chapter is the first draft journal paper prepared mostly by the candidate. This

chapter will be sent to the American Society of Civil Engineers’s Journal of

Geotechnical and Geoenvironmental Engineering, and is co-authored by Professor

David White, Dr. Noel Boylan, Professor Christophe Gaudin and Dr. James Schneider.

The candidate contributed about 70% of the work reported in this chapter.

Dr. James Schneider initially designed the first vertically oriented penetrometer (VOP)

version – i.e. a device used heavily in the experiments reported in this chapter. The VOP

was originally designed to quantify the strength of a submarine slide as it flows around

the VOP. The candidate and Dr. James conducted extensive 1g trial tests (not reported

in the chapter) on this device before finalising the design. Realising the potential for

using the VOP as an in-situ soil characterisation tool (apart from the original intent

stated earlier), the candidate planned (in consultation with Dr. Noel Boylan, Professor

David White and Professor Christophe Gaudin) and performed all the beam centrifuge

xiii

tests, with assistance from the beam centrifuge technician, Mr. Don Herley. These tests

utilised a second VOP version designed by Dr. James and the candidate.

The candidate performed the necessary experimental data interpretations and wrote the

majority of the chapter under the supervision of Professor David White, Dr. Noel

Boylan, Professor Christophe Gaudin and Dr. James Schneider.

Chapter 4

This chapter is the first draft journal paper prepared mostly by the candidate. The

candidate contributed 80% of the work reported in this chapter. This chapter will be sent

to the Canadian Geotechnical Journal, and is co-authored by Professor David White and

Professor Christophe Gaudin.

The candidate performed all beam centrifuge tests using the VOP with assistance from

Mr. Don Herley (beam centrifuge technician).

Comparing with the test data, the candidate developed an analytical framework in

consultation with Professor David White and wrote the first draft chapter. Based on the

suggestions from Professor David White and Professor Christophe Gaudin, the

candidate finalised the chapter.

Paper 2

The second paper is presented as Chapter 5. The paper version of Chapter 5 was slightly

modified to reduce its length and recently submitted as:

Sahdi, F., Gaudin, C., and White, D.J. 2013. The strength properties of ultra-soft kaolin.

Canadian Geotechnical Journal (Submitted: March 2013).

The candidate contributed to about 75% of the work reported in this paper.

xiv

All T-bar centrifuge tests were planned and performed by the candidate in consultation

with Professor Christophe Gaudin. Mr. Bart Thompson (drum centrifuge technician)

provided operational support and guidance during tests.

The candidate conducted numerical back-analyses using MinTaCo with some guidance

from Professor Martin Fahey. The candidate formulated an empirical framework using

the data from the cyclic T-bar tests and those available in the existing literature.

The candidate wrote the paper under the supervision of Professor Christophe Gaudin

and Professor David White.

Paper 3

The third paper is presented as Chapter 6 and the estimated contribution of the candidate

for this paper is 75%. This paper was submitted as (after some content reduction):

Sahdi, F., Gaudin, C., White, D.J., Boylan, N., and Randolph, M.F. 2012. Centrifuge

modelling of active slide-pipeline loading in soft clay. Géotechnique (Submitted:

December 2012).

In order to perform the experiments, some hardware and software upgrades were needed

on the drum centrifuge tool-table. Electronic technicians, Mr. Shane De Catania and Mr.

Tuarn Brown installed a new motion controller card (PCI-7344) in order to achieve the

high velocities needed in the experiments. For ease of control, Mr. Shane De Catania

developed a new LabView interface to allow the candidate to pre-programme specific

test details. Using the candidate’s design, Mr. Dave Jones and Mr. Phillip Hortin

fabricated and instrumented (respectively) the model pipe required in the experiments.

Assisted by Mr. Shane De Catania and Mr. Tuarn Brown, the candidate then conducted

extensive trials and tuning of the drum centrifuge tool-table in order to achieve the

desired high angular velocities with minimal vibrations and kinematic-related errors.

xv

All tests were planned and performed by the candidate in consultation with Professor

Christophe Gaudin. Sample preparation and physical tests were assisted by Mr. Bart

Thompson.

Subsequent data interpretation and paper/chapter write up were performed by the

candidate under the guidance of Professor Christophe Gaudin, Professor David White,

Dr. Noel Boylan and Professor Mark Randolph.

I hereby declare that, except where specific reference is made to the work of others, the

content of this dissertation are original and have not been submitted in whole or in part

for consideration for any other degree of qualification at this, or any other, university.

Fauzan SAHDI, March 2013

1.1

CHAPTER 1. INTRODUCTION

OVERVIEW 1.1

For the past few decades the most important offshore energy resources have been oil

and gas. The construction of the first offshore structure in the Gulf of Mexico (in 1947)

signified the birth of the offshore oil and gas industry. The structure, made up of a

wooden frame, stood at only a mere 6 m from the seabed (Geer 1982). By the late

1980s, the oil and gas industry expanded to the continental margins of the Southern

Californian Coast, the North Sea, the Middle East, the Indonesian Archipelago, the

South China Sea, the southeastern and northwestern shelf areas of Australia, and the

Artic shore of Alaska (Poulos 1988).

To date, the offshore oil and gas industry has expanded near the continental slope with

water depths of 1000 – 2000 m (Puech et al. 2005; Equid 2008; Randolph et al. 2011;

Randolph and Gourvenec 2011). In deep water environments such as those found

offshore Brazil, West Africa, South East Asia, Australia and New Zealand, hydrocarbon

developments are heavily reliant on floating facilities rather than platforms founded on

the seabed. A typical example of an in-field floating production facility is shown in

Figure 1.1. In any hydrocarbon production field, pipelines are of major importance,

linking the various production wells to the buoyant processing facilities, often via steel

catenary risers (SCR). Export pipelines provide a direct route for which hydrocarbon

products (after undergoing the required processing offshore) can be relayed to shore

efficiently.

It is clear that the viability of offshore hydrocarbon productions relies heavily on the

safe and economical design of offshore pipelines, thus, avoiding detrimental

environmental and economic impact should pipeline failure occur. Besides accounting

for the mechanical, structural and hydrodynamic loading aspects in a pipeline design,

various soil-pipeline interaction considerations are equally important. Figure 1.2 depicts

some of the environmental and operational hazards posed on offshore pipelines, which

warrant careful geotechnical consideration. These include mass movement events, or

submarine slides and the various modes of intermittent episodes of cyclic soil-pipeline

1.2

interactions followed by calm periods. This thesis will focus on some of the

geotechnical aspects that arise when considering: (i) submarine slide-pipeline

interaction; (ii) cyclic interaction of pipelines and risers with the seabed.

The soil undrained shear strength is perhaps the most important parameter controlling

the mechanisms of submarine slide-pipeline and cyclic soil-pipeline interactions. The

former mechanism (see Figure 1.2) involves the coupled effects of soil weakening (due

to slope failure and subsequent slide runout) and rate-enhanced soil strength and inertial

effects (owing to the increased velocity of the slide downslope). The latter mechanism

similarly causes soil strength changes due to intermittent episodes of cyclic loading

followed by a calm period. A common theme that will be addressed in this thesis is

therefore the changes in the operative soil strength, which influences the calculations of

forces exerted on a pipeline due to interactions with these mechanisms.

RESEARCH OBJECTIVES 1.2

This thesis aims to provide a better understanding of the evolving soil undrained

strength due to pipeline-slide and cyclic pipeline-soil interactions via extensive

laboratory experiments, which include the usage of the geotechnical centrifuge, X-ray

Diffraction (XRD) and Scanning Electron Microscopy (SEM). This thesis also includes

complementary analytical and numerical analyses to either provide a better

understanding of the governing mechanisms or back-calculate soil parameters important

for subsequent analyses of the experiment data. The development of a new penetrometer

tool to support assessments of near-seabed soil strength is also described.

The main objectives of this thesis are described briefly in the following subsections.

Rate effects 1.2.1

This particular research objective relates directly to slide-pipeline interactions. Pipelines

built in the proximity of continental slopes may cross the path of potential submarine

slides. Submarine slides often have great mobility and involve large volumes, with run-

out distances sometimes exceeding 100 km (Locat and Lee 2002). A good example

1.3

depicting the magnitude and mobility of submarine slides is the case of the historic

Storegga slide (off the coast of Norway) where the Ormen Lange gas field is located. As

shown in Figure 1.3, the slide runout distance exceeded 400 km. Clearly, submarine

slides pose serious consequences on the viability of offshore pipelines. The effect of

submarine slides can be seen in the case of the Mississippi Delta slide when Hurricane

Camille struck (Bea et al. 1983). The slide caused complete destruction of two offshore

structures and another displaced 1 m downslope. Earthquakes, wave loading, tides,

sedimentation, glaciation, formation of diapirs, erosion and gas hydrate disassociation

are some of the factors which can trigger submarine slides (Locat and Lee 2002), even

in areas with slopes of only about l° - 5° (Edgers and Karlsrud 1982; De Blasio et al.

2004a). Despite the risk, offshore structures are still located in areas prone to submarine

slides, which cannot be avoided if this is where the hydrocarbon reserves lie. For

example, the Mad Dog and Atlantis fields in the Gulf of Mexico, with pipelines for the

Mardi Gras export running through the Sigsbee Escarpment, exhibit evidences of past

debris flow cases (Figure 1.4).

Rerouting pipelines around slide hazards is expensive, whilst underestimation of a

hazard could lead to failure and a loss of containment in service. Therefore, engineers

need to assess the potential for submarine slides along the pipeline route and the likely

consequences on the pipeline. It is estimated that submarine slides can travel to

velocities of 7 – 30 m/s (Bjerrum 1971; Imran et al. 2001; Canals et al. 2004; De Blasio

et al. 2004b). Since the magnitude of the slide impact on a pipeline is dependent on both

the velocity and strength of the slide material, a proper method of quantifying this effect

is important.

During the initial stages of a submarine slide (see Figure 1.2), the failed mass possesses

geotechnical properties similar to that of the intact parent slope. As such, slide impact

on a pipeline can be estimated based on the conventional bearing capacity solution (e.g.

Bea et al. 1983; Randolph and Houlsby 1984; Georgiadis 1991; Martin and Randolph

2006) whilst augmenting the slide shear strength to account for the high strain rates,

v/Dpipe (where v is the slide velocity and Dpipe is the pipe diameter) (e.g. Biscontin and

Pestana 2001; Einav and Randolph 2005; Boukpeti et al. 2012).

1.4

Further downslope, the coupled effects of intense shearing within the slide mass and

interaction with the ambient water cause severe soil remoulding. At this stage, the slide

material is known as a debris flow or a turbidity current (see Figure 1.2). Although the

strength of the debris flow or turbidity current is small, the combined effects of weight

and velocity (inertia) are still significant enough to cause damage to a pipeline. To

account for the inertial effects, a fluid mechanics approach is often utilised to quantify

the drag forces (Zakeri et al. 2008; Zakeri 2009; Zakeri et al. 2009). This approach does

not consider the shear strength of the slide as an independent contribution to the impact

force.

The geotechnical and fluid mechanics approaches outlined above can only fully capture

the impact load of a submarine slide across two extreme conditions. The former neglects

the inertial effects, which dominate when the remoulded slide strength approaches zero,

while the latter links the impact load directly to the density and velocity of the slide

even if the impact load is more appropriately a function of the slide strength during the

early stages of a slide (Zhu and Randolph 2011). To address the disassociation between

the geotechnical and fluid mechanics approaches, the first objective of this thesis is:

Objective 1: to study the rate effects of a clay material across a wide range of strain

rates, shear strengths and densities in order to assess the potential of unifying the

geotechnical and fluid mechanics approaches that are used to estimate the slide impact

on a pipeline.

Effects of cyclic loading 1.2.2

Determination of soil strength at the shallowest depths below the seabed (typically

about 1 – 2 m) is of paramount importance for the design of offshore pipelines. Careful

interpretation of in-situ penetrometer tests (White et al. 2010) and novel techniques to

measure strength in box core samples (Low et al. 2008b) may be utilised to improve the

accuracy of determining the near seabed strength. Despite these novel techniques, only

discrete strength measurements can be made along a pipeline route, which may span

over 100 km to shore.

1.5

Accurate measurement of the near seabed strength does not, however, alleviate the

difficulty in estimating the change in the operative soil strength near the vicinity of the

pipeline throughout its lifetime. An offshore pipeline may interact with the seabed in a

cyclic manner, thus remoulding the seabed soil repeatedly. Examples of cyclic loading

events on pipelines are depicted in Figure 1.2. Hydrodynamic loads on a floating

platform may cause cyclic movement of a steel catenary riser (SCR) near the seabed.

This will cause the surrounding soil strength to degrade exponentially, thus reducing the

riser-soil stiffness (Hodder et al. 2009; Hu et al. 2011). During some instances, the

hydrodynamic load on the floating platform may be enough to cause the riser to lift off

from the seabed. This induces further weakening of the seabed due to water

entrainment. Another example of cyclic softening of the seabed soil is the controlled

lateral buckling of on bottom pipelines due to temperature fluctuations during startup

and shutdown events (Figure 1.2). If the cyclic loading event is fast enough, positive

excess pore pressure will be generated (for clays and silts with stress histories on the

‘wet side of the critical state line’) – i.e. undrained cyclic loading. However, discounting

thixotropic effects, the seabed strength may recover, as positive excess pore pressure

dissipates during a calm period.

The effects of pipeline-soil remoulding and reconsolidation can be simulated using a T-

bar penetrometer, which itself is essentially a scaled down version of a pipeline section.

An example of these two effects (White and Hodder 2010) is illustrated in Figure 1.5.

These episodic T-bar tests were conducted in the centrifuge on a sample of lightly

overconsolidated kaolin. The operative strength was extracted from a depth of 2.25 m

(prototype scale) below the seabed. It can be seen that the 2 calm periods (allowing

reconsolidations) caused the remoulded strength in the last cyclic test to increase by a

factor of about 2 compared to that during the first cyclic episode.

White and Hodder (2010) showed that the evolution of the operative strength depicted

in Figure 1.5 can be simulated using an exponential strength degradation model (Einav

and Randolph 2005) coupled with concepts of critical state soil mechanics (Schofield

and Wroth 1968; Atkinson and Bransby 1978). Their proposed framework showed good

agreement with the T-bar data (see Figure 1.5). However, the parameters used in the

strength degradation model of Einav and Randolph (2005) can only be obtained from

1.6

back-calculation of a full T-bar cyclic test, and no attempts to link these parameters with

soil properties have been made. Moreover, the remoulding and reconsolidation

framework proposed by White and Hodder (2010) assumed that no pore pressure

dissipation occurs within a cyclic episode, which is compatible to their observed

experimental data. In contrast, Rismanchian et al. (2011) showed that the operative soil

strength may increase even within a cyclic episode.

The selection of an appropriate design soil strength for pipe-seabed and SCR-seabed

interaction is also hampered by the very weak sediments involved. Conventional

geotechnical engineering operates at significantly higher stress levels, whereas close to

the mudline and within debris flows the soil consistency is close to the liquid limit or

wetter. The mechanisms, which control the cyclic strength degradation at low stresses,

have not been properly addressed.

Attempts to address these aspects will form the second objective of this thesis:

Objective 2: to study the changing strength of clay due to remoulding and

reconsolidation using penetrometer tests, and explore improved methods for defining

the strength and strength degradation parameters of very soft sediments. A critical state-

based remoulding and reconsolidation framework, which allows pore pressure

dissipation within a cyclic episode, will also be explored.

THESIS OUTLINE 1.3

Apart from Chapters 1 and 7, the remaining portion of this thesis is presented as a series

of technical papers comprising of a published peer-reviewed conference paper (Chapter

2), 2 submitted journal papers (Chapters 5 and 6) and 2 journal papers drafted for

publications (Chapters 3 and 4). A chart showing the organisation and interrelationships

(where applicable) of the chapters under the two main research objectives is shown in

Figure 1.6. The organisation of this thesis is as follows:

Chapter 2 outlines an experimental study to investigate the strength behaviour

of clay with different microfabrics under intense cyclic remoulding.

Experimental methods using centrifuge T-bar tests coupled with X-ray

1.7

Diffraction and Scanning Electron Microscopy studies are presented. Both

monotonic and cyclic shear strengths derived from T-bar tests are linked with

the orientation of clay platelets, and the influence of microstructure on the initial

strength and strength degradation is studied.

Chapter 3 presents a centrifuge testing programme evaluating the potential of a

novel penetrometer (named the vertically oriented penetrometer, VOP) to

improve the accuracy of site investigation practice for offshore pipeline design.

The experimental data are compared with existing published work of a laterally

loaded pile to derive suitable bearing factors. This study also presents the

gradual softening of clay from the ‘intact’ to the stain-softened operative

strength and further cyclic strength degradation characteristics particular to the

VOP, which are equally applicable for pipeline cyclic loading. This chapter also

presents a parametric study of viscous rate effects in intact and remoulded clay

relevant to the initial stages of a submarine slide.

Chapter 4 continues the investigation of the effects of cyclic loading on clay

observed using the VOP, with allowance of soil strength regain through

reconsolidation within a cyclic loading episode. An analytical framework that

combines an exponential strength decay model with critical state soil mechanics

and a 1-dimensional consolidation solution is outlined. VOP bearing factors

derived from Chapter 3 are utilised to estimate the resulting evolution of VOP

resistance. This framework is compared to centrifuge test results where the VOP

is cycled at strain rates that permit reconsolidation in between passes of the VOP

at a particular soil location.

Chapter 5 presents an attempt to link the exponential strength degradation trend

during a cyclic loading event of a T-bar (also applicable for offshore pipelines)

with the basic properties of clay. An empirical improvement to the exponential

strength degradation model is proposed based on extensive centrifuge T-bar

cyclic tests on kaolin clay at different degrees of consolidation (reflecting a wide

range of water contents) coupled with other field and centrifuge data reported in

the literature. Large strain numerical back analyses of the consolidation process

are performed using MinTaCo (Mine Tailings Consolidation), and applied in

subsequent analyses of the centrifuge test results outlined in Chapter 6.

1.8

Chapter 6 continues the rate effects theme of Chapter 3. Instead of the VOP,

centrifuge model pipe test programme is conducted to quantify the velocity

effects of a submarine slide in the viscous to inertial domains. This experiment

involves dragging the model pipe at velocities ranging from 0.004 to 4.2 m/s in

very soft consolidating kaolin samples. Centrifuge test results and data obtained

from the literature are combined to calibrate a hybrid geotechnical-fluid

mechanics approach, which quantifies both horizontal and vertical drag forces

on an offshore pipeline under impact of a submarine slide.

Chapter 7 provides the conclusions and highlights the important findings of this

study. Recommendations for future work are also presented.

Chapter 8 provides the list of references used in this thesis.

1.9

Figure 1.1. In-field buoyant processing facility and the associated pipeline and riser networks

(Jayson et al. 2008)

1.10

Figure 1.2. Examples of environmental and operational hazards on offshore pipelines

Slo

pe

fail

ure

Init

ial

slid

e

blo

ck

Deb

ris

flo

w

or

turb

idit

y

curr

ent

Cy

clic

mo

tio

n o

f a

stee

l ca

ten

ary

ris

er

at t

he

tou

chd

ow

n

zon

e

Tem

per

atu

re-i

nd

uce

d

pip

elin

e b

uck

lin

g

Wav

e-in

du

ced

mo

vem

ent

on

flo

atin

g p

latf

orm

Ex

po

rt

pip

elin

e

An

cho

red

to

sea

bed

1.11

Figure 1.3. Runout extent of the Storegga slide (> 400 km) located between the VØring Plateau and

the North Sea Fan (De Blasio et al. 2004b)

Figure 1.4. Mad Dog and Atlantis fields located near potential submarine slide hazard in the

Southern Green Canyon area of the Gulf of Mexico (Jeanjean et al. 2005)

1.12

Figure 1.5. Change in the operative strength due to episodic events of T-bar cyclic remoulding and

reconsolidation (White and Hodder 2010)

1.13

Figure 1.6. Thesis organisation and interrelationships between chapters

Th

e ch

an

gin

g s

tren

gth

of

cla

y a

nd

its

ap

pli

cati

on

to

off

sho

re p

ipel

ine

des

ign

Th

e ch

an

gin

g s

tren

gth

of

cla

y a

nd

its

ap

pli

cati

on

to

off

sho

re p

ipel

ine

des

ign

Ob

ject

ive

1:

Rat

e ef

fect

s

Ob

ject

ive

1:

Rat

e ef

fect

s

Ob

ject

ive

2:

Beh

avio

ur

un

der

cy

clic

load

ing

Ob

ject

ive

2:

Beh

avio

ur

un

der

cy

clic

load

ing

Ch

ap

ter

1:

Intr

od

uct

ion

of

thes

is c

on

ten

t

Ch

ap

ter

1:

Intr

od

uct

ion

of

thes

is c

on

ten

t

Ch

ap

ter

3:

Stu

dy

th

e v

isco

us

effe

cts

of

inta

ct a

nd

rem

ou

lded

clay

usi

ng

th

e v

erti

call

y

ori

ente

d p

enet

rom

eter

(VO

P)

Ch

ap

ter

3:

Stu

dy

th

e v

isco

us

effe

cts

of

inta

ct a

nd

rem

ou

lded

clay

usi

ng

th

e v

erti

call

y

ori

ente

d p

enet

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eter

(VO

P)

Ch

ap

ter

6:

Qu

anti

fyin

g t

he

vis

cou

s

and

in

erti

al e

ffec

ts o

f a

sub

mar

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slid

e im

pac

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n

off

sho

re p

ipel

ines

Ch

ap

ter

6:

Qu

anti

fyin

g t

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cou

s

and

in

erti

al e

ffec

ts o

f a

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slid

e im

pac

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sho

re p

ipel

ines

Ch

ap

ter

5:

Em

pir

ical

fra

mew

ork

to

inte

rpre

t th

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ipel

ine-

seab

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Ch

ap

ter

5:

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pir

ical

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to

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Ch

ap

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3:

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so

ften

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eff

ects

du

rin

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on

oto

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d

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isp

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men

ts o

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the

VO

P

Ch

ap

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3:

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so

ften

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eff

ects

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on

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d

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the

VO

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Ch

ap

ter

2:

Eff

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of

clay

mic

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stre

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th d

egra

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beh

avio

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Ch

ap

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2:

Eff

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of

clay

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Ch

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cycl

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Ch

ap

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4:

Co

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Ch

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7:

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Ch

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7:

Co

ncl

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on

s an

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utu

re

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rk

Bea

rin

g

fact

ors

Bea

rin

g

fact

ors

2.1

CHAPTER 2. THE INFLUENCE OF COLOURED DYES

ON THE UNDRAINED SHEAR STRENGTH OF KAOLIN

FOREWORD

This chapter falls within Objective 2 (see Section 1.2.2). The link between soil

microstructure and shear strength is investigated in this chapter. In particular, the

degradation of clay undrained shear strength due to cyclic loading measured with the

T-bar penetrometer (which is essentially a model pipe) is studied. The outcomes of this

chapter can be used to understand the role of clay microstructure in practical offshore

pipeline-soil interactions that involve cyclic remoulding of the founding soil.

2.2

ABSTRACT 2.1

Colouring of clay with dyes can be a useful method for monitoring processes in physical

modelling where different soil masses interact. In the present study, centrifuge

modelling of the interaction of a submarine slide and a model seabed necessitated the

addition of coloured dyes to aid differentiation of the slide and the seabed after a slide

event. Assessments of pH, index properties and undrained shear strength, supported by

Scanning Electron Microscopy (SEM) and X-Ray Diffraction (XRD) studies, have been

used to understand the influence of different dyes on kaolin clay. The addition of

different dyes was found to influence the initial undrained shear strength (su-in) of the

dyed samples while the remoulded shear strength (su-rem) was similar for all samples.

XRD and SEM tests revealed micro-structural changes in the dyed samples, which may

have caused the increase in su-in.

INTRODUCTION 2.2

Colouring of soil can be a useful method of differentiating different soil masses in

physical modelling. However, the influence of the dye on the geotechnical properties of

the soil itself is not always considered. This study was carried out as part of a project at

the Centre for Offshore Foundation Systems (COFS) aimed at simulating submarine

slides in the geotechnical drum centrifuge at the University of Western Australia

(UWA). This project aims to study the breakdown of the slide material from its intact

condition along a model seabed. One of the objectives of this project is to quantify the

degree of erosion of the model seabed during a slide event. To achieve this, the seabed

(which consists of kaolin clay) is dyed to aid differentiation from the slide runout

material, which is left undyed. Prior to conducting the slide experiments, this study was

carried out to investigate the influence of dyes on the geotechnical properties of kaolin

clay. The outcomes of this study are reported in this paper.

Many physical experiments including centrifuge modelling, triaxial tests, and also

modelling at unit gravity (1g) use different types of dyes to colour the sample area of

interest relative to the surrounding soil. Examples of published experimental works

which used dyes to better identify the sample area of interest are reported by Mohrig et

2.3

al. (1999) and Black et al. (2007). Mohrig et al. (1999) investigated whether a debris

flow can increase its mass by mobilizing an antecedent failure deposit using large flume

tests at unit gravity (1g). For these experiments, they used dyed slurry for the debris

flow and found that the dying technique changed the properties of the slurry. Black et

al. (2007) modelled the behaviour of clay reinforced with sand columns in a triaxial cell.

The sand columns were dyed to differentiate them from the surrounding clay. However,

the possibility of the dye influencing the behaviour of sand columns was not considered.

Dyes contain chemicals that may change the microstructure or influence the

interparticle attraction forces and can ultimately cause changes in the geotechnical

properties of a clay. To quantify the effects of different dyes on the geotechnical

properties of kaolin clay, pH tests, basic index testing as well as assessment of the shear

strength from T-bar penetrometer tests were carried out. Scanning Electron Microscopy

(SEM) and X-ray Diffraction (XRD) studies were also utilised to investigate the

influence of dyes on the micro-fabric of the samples.

SAMPLE PREPARATION 2.3

Samples for strength characterisation 2.3.1

A total of 6 samples were prepared for the characterisation of the strength of the dyed

and un-dyed clays by T-bar tests. Table 2.1 lists the different samples, the dye type and

the concentration of dye used. Samples B1-3A, B1-4A and B1-6A were coloured with

the minimum dye concentrations that gave satisfactory colour contrast compared to

standard kaolin, which is usually white. The concentration of the Art Spectrum®

Red

dye was doubled for sample B1-5A to study the effect of increasing the dye

concentration. Samples B1-1A and B1-2A were control samples which were left un-

dyed. To prepare each coloured sample, kaolin slurry at a water content of

approximately twice the liquid limit (120%) was first thoroughly mixed in a mechanical

mixer for 10 minutes. The appropriate amount of dye was then added into the slurry and

mixed for a further 20 minutes until the sample was homogeneous. The control samples

(B1-1A and B1-2A) were prepared by mixing the kaolin slurry in the mechanical mixer

for 30 minutes.

2.4

T-bar penetrometer tests were conducted in the UWA drum centrifuge. Details of the

testing regime are provided in Section 2.4.2. The samples were contained in Polyvinyl

chloride (PVC) tubes which were held in a rectangular steel box in the drum centrifuge

channel. Each PVC tube had a diameter of 63 mm and a height of 140 mm. A 10 mm

thick sand layer was placed at the base of each PVC tube to facilitate two-way drainage

during consolidation. The samples were poured into the PVC tubes while the centrifuge

was spinning at 20 g before being finally increased to 100 g for four days to ensure full

self-weight consolidation.

Samples for SEM, XRD, index and pH tests 2.3.2

To provide sufficient strength for sample trimming to small sizes (typically in mm’s),

samples for the Scanning Electron Microscopy (SEM) and X-ray Diffraction (XRD)

tests were prepared by consolidating the slurry-dye mixtures in steel cylinders. Each

cylinder had a height of 250 mm and a diameter of 72 mm. A top platen was used to

apply an axial stress of 100 kPa to each sample. Apart from the consolidation method,

the dye concentrations and mixing procedures used to prepare the samples were similar

to the preparation methods used for the centrifuge samples. These samples are denoted

by the letter “B” as shown in Table 2.1. Unlike for the centrifuge tests, only one control

sample (B1-1B) was prepared.

After consolidation, two rectangular sections from each sample, perpendicular to the

consolidation loading axis were trimmed to sizes of 7 × 9 × 1 mm3

and 15 × 20 × 10

mm3

for the SEM analyses. The smaller section was prepared for freeze-drying while

the larger section was trimmed for oven-drying. Two additional rectangular sections

normal (horizontal section) and parallel (vertical section) to the direction of the loading

platen (during consolidation in the cylinder) measuring 27 × 21 × 4.4 mm3 were

trimmed for XRD tests. The remaining portion in each sample was used for index and

pH tests.

2.5

RESULTS 2.4

Index properties and pH 2.4.1

Liquid limit and plastic limit tests were conducted in accordance with AS 1289.3.9

(1991) and AS 1289.3.2.1 (1995) respectively. Sample B1-1B (a control sample) was

found to have a liquid limit of 56% and a plastic limit of 29% which gives a plasticity

index of 27%. Index properties of the other dyed samples fall within the range of 56-

57% for the liquid limit and 29-30% for the plastic limit. There is, therefore, no

apparent change in the index properties due to the inclusion of these dyes.

The pH of the dyes range from 7.6-9.2 with the exception of the Copic®

Various Ink

Apple Green dye where the pH is 2.81. However, the pH of the solid consolidated

samples (both dyed and un-dyed) were very similar (pH 7.3-7.7).

T-bar penetrometer testing 2.4.2

T-bar penetrometer tests were conducted at 100 g after full consolidation of the samples.

The results for the centrifuge tests will be described hereafter in prototype scale units,

unless stated otherwise.

T-bar penetrometer tests (Stewart and Randolph 1991) using a T-bar with a diameter of

5 mm and 20 mm in length (model scale) were carried out to determine the initial shear

strength (su-in) profile and the strength degradation from penetrometer cycling of each

sample. Each test consisted of an initial penetration to a depth of about 8 times the bar

diameter (8DT-bar) followed by a cyclic penetration and extraction regime over a sample

depth of 4DT-bar to 8DT-bar. An additional penetration to a sample depth of 10DT-bar was

also performed before full extraction from the sample. All T-bar tests were conducted at

a rate of 1 mm/s to ensure undrained conditions. The T-bar resistance, qT-bar, was

converted to the undrained shear strength, su, using a bearing factor, NT-bar, as shown in

Equation 2.1:

2.6

barT

barT

N

q

us

2.1

A constant value of NT-bar = 10.5, representing the theoretical value for an intermediate

roughness condition between fully smooth and fully rough (Martin and Randolph 2006)

was used to calculate su. Figure 2.1 shows the undrained shear strength profiles derived

from the first T-bar penetration (su-in) for samples B1-1A to B1-6A. It is evident that

sample B1-6A, coloured with 0.025 ml/g Copic®

Various Ink Green (lowest dye

concentration), resulted in the smallest change in the su-in profile compared to control

samples B1-1A and B1-2A, with only minor differences (<10%) at depths greater than

4DT-bar. The addition of dyes in the 3 other samples resulted in an increase of su-in with

B1-5A having the highest initial strength followed by B1-3A and B1-4A.

From cyclic tests, the influence of dye on the strength degradation properties can be

investigated by plotting a graph of degradation factor against the cycle number. An

example of this graph is shown in Figure 2.2 for sample B1-1A. The degradation factor

is defined as the strength during a particular phase, su-n, normalised by the initial su-in

derived from the initial T-bar penetration. These su-n values were taken at the mid depth

of a particular penetration or extraction phase. Following Randolph et al. (2007), the

first cycle is taken as 0.25 and subsequent cycles are cumulatively taken as increments

of 0.5 (0.75, 1.25, etc.). An exponential degradation curve was fitted to the data points

in Figure 2.2, in the manner that is commonly used to capture the strength degradation

pattern (Einav and Randolph 2005):

95

3(n 0.25)

N

rem remΔ n Δ 1 Δ e

2.2

where ∆(n) is the degradation factor during the nth

cycle, ∆rem is the ratio of the fully

remoulded strength (su-rem) to the initial strength (su-in) and N95 is the number of cycles

required to achieve 95% degradation from the initial strength. The magnitudes of the

strength gradient with depth (ksu), sensitivity (St) (which is the inverse of ∆rem), and N95

are shown in Figure 2.3. Figure 2.4 shows the strength degradation profiles for all the

samples where the su-n is normalised by the in-situ vertical effective stress (σ'vo) at the

2.7

midpoint of the cyclic phase for each sample to account for the slight differences in

sample heights between samples. The normalised undrained shear strengths for all

samples were found to converge within a narrow band after approximately 9.25 cycles.

Figure 2.1 and Figure 2.3 indicate clearly that the Copic® Various Ink Apple Green dye

(B1-6A) does not affect either the strength of the clay nor the sensitivity. In contrast, the

other dyes (B1-3A to B1-5A) result in significant increases of the initial strength

gradients (ksu). Although there is no clear trend of the rate of strength degradations for

both dyed and un-dyed samples (represented by the N95 parameter), the fully remoulded

strength (su-rem) is similar for all samples as demonstrated in Figure 2.4 (the sensitivity

varies only due to the change in the initial strength).

SEM and XRD analyses 2.4.3

Investigation of soil micro-fabric using SEM requires complete removal of the pore

fluid from within a sample (Mitchell and Soga 2005). The freeze-drying and oven-

drying methods were employed to investigate the best sample preparation method for

the SEM and XRD. In the freeze-drying method, each trimmed sample (7 × 9 × 1 mm3

in size) was immersed in liquid nitrogen at -196°C for instant freezing to avoid the

formation of ice crystals. The sample was then fractured in half using a sharp razorblade

while still immersed in liquid nitrogen. Care was taken to ensure that at least one of the

sample edges was parallel to the consolidation loading axis (vertical section) for SEM

observations. After freezing, the samples were transferred into an Emitech K775X

freeze drier for sublimation. In the oven drying method, the samples (15 × 20 × 10

mm3) were dried in an oven heated to a temperature of 50°C. After drying, each oven-

dried sample was fractured in half to generate an undisturbed surface for SEM

observation which was parallel to the loading (during consolidation) axis. Prior to SEM

analyses, the freeze-dried and oven-dried samples were coated with a 2 nm thick layer

of platinum to avoid surface charging during SEM observations.

SEM observations were performed using the Zeiss 1555 Field Emission Scanning

Electron Microscope (FESEM). The SEM pictures revealed similar micro-structure for

all samples prepared by freeze-drying and oven-drying. Figure 2.5 and Figure 2.6 show

the SEM micrographs of samples B1-1B and B1-5B, including their loading directions

2.8

respectively. These samples were prepared by oven-drying. The SEM pictures were

taken at a magnification of ×25000. Both samples appear to have similar flocculated

fabric with random particle orientations.

XRD tests were conducted to provide a more quantitative analyses of the soil micro-

fabric. XRD is capable of identifying minerals as well as revealing soil particle

orientations. Pore fluid within the trimmed small rectangular sections (see Section 2.3.2)

was removed by oven-drying at 50°C. The surface of each dried section was then

cleaned using adhesive tape to minimize disturbance from trimming prior to XRD tests.

The Philips PW1830 X-ray Diffractometer was used to analyse the samples. XRD

patterns were obtained using Cu-Kα radiation with a wavelength (λL) of 1.5418 Å. All

XRD tests were performed at a scanning rate of 1.5 °/min between angles (2θ) of 5° to

85°. Identification of minerals can be achieved using Bragg’s law (Mitchell and Soga

2005) which is shown in Equation 2.3:

1 L sn 2d sin 2.3

where n1 is the reflection order, and ds is the spacing between atomic planes. Because

no two minerals have the same ds-spacings, the angle of diffraction, θ can be used to

indentify the minerals present within a soil mass.

Figure 2.7 and Figure 2.8 depict the XRD patterns of the horizontal and vertical sections

respectively for all of the samples. The Miller Index associated to the basal peaks (001

and 002) and prism peaks (130 and 022 ) at their respective 2θ angles which are

normally used to identify kaolinite minerals are also shown. For clarity, only XRD

patterns between 2θ of 5° to 65° are shown in Figure 2.7 and Figure 2.8. Irradiated x-ray

waves from the edges and faces of clay particles will result in prism and basal peaks

respectively. High intensities of both basal and prism peaks indicates a random micro-

fabric (Sachan and Penumadu 2007a). Therefore, the clay platelets for all dyed and un-

dyed samples appear to have random particle orientations. The only differences

observed in the XRD patterns are the intensities of the basal and prism peaks. For

example, this can be seen in the XRD patterns of the (002) basal peaks for both

horizontal and vertical sections in Figure 2.7 and Figure 2.8 respectively. Note that on

2.9

the upper right sides of both figures, the (002) basal peaks are enlarged for clarity.

These differences in intensities can be used to quantify the kaolin particle orientations

using the Fabric Index, FI, (Odom 1967; Gillott 1970; Yoshinaka and Kazama 1973) by

using the (002) basal peaks according to Equation 2.4:

A

A A

VFI

H V

2.4

where VA is the area of the basal diffraction peak for the vertical section (parallel to the

axis of loading during consolidation) and HA is the basal peak area for the horizontal

section (perpendicular to loading platen axis). Theoretically, the Fabric Index can range

from 0 to 1.0 where 0 indicates that all the faces of clay platelets are perfectly

perpendicular to the loading axis (horizontal section), 0.5 indicates a perfect random

orientation (edge-to-face) and 1.0 signifies that the faces of clay particles are all aligned

parallel to the loading axis (vertical section).

Figure 2.9 shows the FI obtained using the (002) basal peaks plotted against the strength

gradient, ksu, for all the samples. It can be seen that the stronger the sample, the higher

the magnitude of the FI, thus providing a micro-structural parameter that is consistent

with the macroscopic initial undrained strength behaviour.

DISCUSSION 2.5

As shown in Figure 2.1 and Figure 2.3, the strength gradient, ksu, of sample B1-6A

(dyed) is only about 8.5% higher than the strength gradients of the control samples, B1-

1A and B1-2A. The dyes in samples B1-3A to B1-5A appear to increase the ksu of

standard kaolin by approximately 47 - 81.5%. The effect of doubling the concentration

of the Art Spectrum® Red dye

(B1-5A) resulted in an increase of 24% in the ksu

compared to sample B1-4A. The increase in the strength measured from initial T-bar

penetrations, su-in, may be attributed to the way the dye chemicals influence the particle

orientation of kaolin. SEM observations (Figure 2.5 and Figure 2.6) showed that the

samples have a flocculated micro-fabric with random particle orientations. Results of

XRD tests indicated a trend of increasing Fabric Index (FI) with increasing ksu in the

2.10

samples. Samples containing the Art Spectrum®

Red and Artelier®

Carbon Black dyes

appear to have higher magnitudes of FI, which indicates that the kaolin particle

orientations are more random compared to samples B1-1A/B and B1-6A/B with a FI of

0.41 each, indicating that more clay particles are aligned perpendicular to the

consolidation loading direction. The FI for samples B1-3A/B and B1-5A/B exceeded

0.5. This means that the kaolin particle faces are slightly more aligned parallel to the

consolidation loading axis. Soils with flocculated, random particle orientations are often

regarded as having higher magnitude of strength (Mitchell and Soga 2005) compared to

dispersed micro-fabric (Falamaki et al. 2008). A possible explanation for this is that the

more random arrangement of fabric provides a compact, truss-like ‘structural’ support

to the soil matrix compared to a fabric having more face-to-face contacts, which is less

compact. This truss-like structure is easily collapsed upon shearing and this is reflected

by the rapid loss of initial strength from cyclic T-bar tests (Figure 2.4).

From Figure 2.3, the higher magnitudes of ksu for samples B1-3A, B1-4A and B1-5A

are mirrored by the increase in sensitivity, St, for these samples, indicating that the

remoulded strength is essentially similar for all samples. This effect is shown by the

normalised strength degradation with cycle number as depicted in Figure 2.4. It can be

hypothesised that during cyclic T-bar remoulding, reorientation of the particles into

similar patterns as the un-dyed sample occurs, leading to similar remoulded shear

strength (su-rem). The addition of different dyes appears to only influence the initial

particle orientations and attractions without affecting its mineralogy, as confirmed by

the similar XRD patterns in Figure 2.7 and Figure 2.8. This may explain the similar

liquid and plastic limits for all the samples. Although two of the dyes were alkaline (pH

9.2) and acidic (pH 2.81), the pH of all solid consolidated samples were similar. This

may be attributed simply to dilution, due to the small concentrations of dyes used to

prepare the samples.

The mechanism that leads to the increasing degree of randomness of the particle

orientations in samples B1-3A/B, B1-4A/B and B1-5A/B (as depicted in Figure 2.9) is

unclear and requires chemical analyses of the dyes used in this study. However, this

phenomenon may be induced by the probable presence of reactive ions in the dyes that

2.11

will influence the surface charges on the kaolin platelets leading to an increase in the

degree of edge-to-face contacts.

All of the samples have good colour contrast relative to the normal un-dyed samples

(which is usually white). Visual observation also revealed that there was no dye

leaching into the surrounding water for each coloured sample. It should be noted that

the Art Spectrum®

Red at a concentration of 0.031 ml/g was chosen to colour the model

seabed. Although the addition of the Copic®

Various Ink Green dye resulted in the least

amount of alteration in the behaviour of kaolin, colouring the seabed using the Art

Spectrum®

Red at a concentration of 0.031 ml/g can achieve better colour contrast.

Furthermore, the overall cyclic strength degradation profile and su-rem are similar to un-

dyed kaolin.

CONCLUSIONS 2.6

This paper examines the influence of dyes on the geotechnical properties of kaolin clay.

The motivation of this study is the requirement to provide a model seabed with distinct

colour relative to the slide runout material for centrifuge submarine slide modelling

where quantifying the level of seabed erosion is of interest.

T-bar penetrometer tests revealed an increase in the initial undrained shear strength (su-

in) of kaolin clay for samples coloured with the Art Spectrum®

Red and Artelier®

Carbon

Black dyes. This increase in su-in is believed to be associated with an increase in the

degree of edge-to-face contacts of the kaolin platelets. This is suggested by the Fabric

Index that was derived from XRD tests, which correlated well with the strength

gradients (ksu) of the samples. After many cycles of T-bar penetration and extraction,

the remoulded shear strengths (su-rem) for all samples were similar, indicating a similar

micro-structure after the samples are fully remoulded. The mineralogy and index

properties of the kaolin clay were unaffected by the addition of the different dyes.

2.12

ACKNOWLEDGEMENTS 2.7

The research presented in this paper forms part of a Joint Industry Project administered

and supported by the Minerals and Energy Research Institute of Western Australia, and

by BP, BHP Billiton, Chevron, Petrobras, Shell and Woodside. The financial support of

all the participants is gratefully acknowledged. This research is being undertaken within

the CSIRO Wealth from Oceans Flagship Cluster on Subsea Pipelines. The authors

acknowledge the facilities, scientific and technical assistance of the Australian

Microscopy and Microanalysis Research Facility at the Centre for Microscopy,

Characterisation and Analysis, The University of Western Australia, a facility funded by

The University, State and Commonwealth Governments. The authors are also grateful

to Bart Thompson, Claire Bearman, Dr. James Schneider, Dr. Saowanuch Tawornpruek,

Peter Duncan and Dr. Jeremy Shaw for their technical assistance. The first author

acknowledges the financial support received from the Ministry of Higher Education

Malaysia and Universiti Malaysia Sarawak (UNIMAS).

2.13

Table 2.1 Details of samples with dye mixtures

Sample Name of dye Colour Dye concentration (ml/g)*

B1-1A/B - - -

B1-2A - - -

B1-3A/B Artelier®

Carbon Black Black 0.031

B1-4A/B Art Spectrum®

Red Red 0.031

B1-5A/B Art Spectrum®

Red Red 0.063

B1-6A/B Copic®

Various Ink Apple Green Green 0.025

* Expressed in terms of volume of dye (ml) per unit mass of dry kaolin (g).

A Sample for T-bar tests.

B Sample for SEM, XRD, index and pH tests

2.14

Figure 2.1. Undrained shear strength from the first T-bar penetration

Figure 2.2. Cyclic strength degradation for sample B1-1A

0

1

2

3

4

0 1 2 3 4 5 6

Dep

th,

z (

mm

)Undrained shear strength of initial penetration, su-in

B1-1A B1-2A

B1-3A B1-4A

B1-5A B1-6A

0

0.2

0.4

0.6

0.8

1

0 2 4 6 8 10 12Deg

rad

ati

on

fa

cto

r, s

u-n

/su

-in

Cycle number

From T-bar

Eq. 2.2

Dep

th, z

(m)

2.15

Figure 2.3. Comparison of ksu, St and N95 for all samples

Figure 2.4. Cyclic variation in the normalised undrained shear strength

0

1

2

3

4

5

B1-1A B1-2A B1-3A B1-4A B1-5A B1-6A

Ma

gn

itu

de

Sample

Strength gradient, ksu (kPa/m)

Sensitivity, St

N95


Sensitivity, St

N95

0

0.1

0.2

0.3

0 2 4 6 8 10 12Un

dra

ined

str

ength

rati

o,

s u-n

/σ' v

o

Cycle number

B1-1A B1-2A

B1-3A B1-4A

B1-5A B1-6A

2.16

Figure 2.5. SEM micrograph of sample B1-1B (vertical section) – undyed sample prepared by oven-

drying

Figure 2.6. SEM micrograph of sample B1-5B (vertical section) – coloured with Art Spectrum® Red

dye (0.063ml/g) prepared by oven-drying

Direction of consolidation loading

Direction of consolidation loading

2.17

Figure 2.7. XRD patterns for horizontal sections

Figure 2.8. XRD patterns for vertical sections

0

1000

2000

3000

4000

5 15 25 35 45 55 65

B1-1

B1-3B1-4

B1-5B1-6

(002)

(001)

Basal

peaks

)022((130)

Prism peaks

0

2000

4000

24 25 262θ

Inte

nsi

ty

2θ

Inte

nsi

ty

B1-1B

B1-3B

B1-4B

B1-5B

B1-6B

0

1000

2000

3000

4000

5 15 25 35 45 55 65

B1-1

B1-3

B1-4

B1-5

B1-6

(001) (002)

Basal

peaks

(130))022(

Prism peaks

2θ

Inte

nsi

ty

0

2000

4000

24 25 262θ

Inte

nsi

ty

B1-1B

B1-3B

B1-4B

B1-5B

B1-6B

2.18

Figure 2.9. Fabric Index of samples

0.35

0.4

0.45

0.5

0.55

0.6

0.5 0.7 0.9 1.1 1.3 1.5

Fa

bri

c In

dex

, F

I


B1-1A/B B1-6A/B B1-4A/B

B1-3A/B B1-5A/B

3.1

CHAPTER 3. DEVELOPMENT OF THE VERTICALLY

ORIENTED PENETROMETER FOR SHALLOW SEABED

CHARACTERISATION

FOREWORD

This chapter reinforces the cyclic undrained strength degradation theme of Chapter 2

by discussing the use of the T-bar and a new device – the vertically oriented

penetrometer (VOP) – to quantify the undrained strength degradation characteristics of

clay (Objective 2 – see Section 1.2.2). It documents the capability of the VOP to not

only measure the initial (operative) and remoulded strengths typical for a T-bar, but to

also measure the intact undisturbed strength. This is vital when considering the peak

resistance of pipeline-soil interactions, where the soil strength has not decreased to the

operative strength typically measured during a full-flow failure of a T-bar. Suitable

bearing factors used to convert the VOP measured resistance to the undrained strength

are established, which will be used later in Chapter 4. This chapter also addresses, for

the first time in this thesis, the effects of high strain rates (Objective 1 – see Section

1.2.1) in the viscous regime – i.e. when the undrained strength of a failed slide material

is sufficiently high and the velocity of the slide is comparatively low (typical during the

initial slide phase). This rate effects theme is extended further in Chapter 6.

As a whole, this chapter serves a three-fold purpose, because it addresses both

Objectives 1 and 2 and it paves the way for improved surficial strength measurements

of the seabed, which is of paramount importance for offshore pipeline design.

3.2

ABSTRACT 3.1

Current site investigation practice for offshore pipeline design relies heavily on soil

parameters gathered from boreholes or in-situ testing sites often spaced many

kilometres apart. Estimation of the geotechnical parameters between site investigation

areas are then achieved using geophysical data, with decreasing accuracy as the spacing

between boreholes or in-situ soundings increases. This paper describes a centrifuge test

programme designed to evaluate the performance of a novel penetrometer, namely the

vertically oriented penetrometer (VOP). The VOP is a potentially valuable addition to

the range of tools used to characterise the strength behaviour of soft soils, both in small

scale centrifuge models and, following practical development, in the field. This can be

achieved by dragging the VOP along a planned site investigation route, and deriving the

soil strength from the measured resistance using an approach common for laterally

loaded piles. Centrifuge VOP horizontal drag tests in normally consolidated (NC) and

highly overconsolidated (HOC) kaolin samples confirmed that the shear strength profile

of the soil can be derived using a framework idealising the VOP as a laterally loaded

pile. The strain rate dependency of the VOP resistance was also studied by conducting

tests at velocities ranging from 1 to 30 mm/s, and the change in resistance per log cycle

increase in shearing rate was found to be consistent with other conventional

penetrometers. The VOP, in its present configuration, was also found to provide good

estimates of the average soil sensitivity. The VOP eliminates some limitations of

conventional penetrometers when assessing very shallow seabed behaviour. The entry

and exit of a conventional penetrometer from the seabed leads to a transitional

behaviour from shallow to deep embedment, and contrasting behaviour when extracting

and penetrating due to buoyancy and the lifted soil above the penetrometer. These

effects are not present when using the VOP.

INTRODUCTION 3.2

Offshore pipelines are one of the most important elements for deep water hydrocarbon

developments. Pipelines are either used to transport hydrocarbon products between

production wells and processing facilities, or directly to shore. Successful engineering

3.3

design of an offshore pipeline network depends on careful considerations of soil-

pipeline interactions. In turn, design parameters for this particular aspect greatly hinge

upon quality site investigation along the pipeline route. Further from shore, fine-grained

sediments dominate (Randolph 2004) and therefore one of the most important

parameters for pipeline design is the shear strength, su, and its variation due to both

remoulding and subsequent reconsolidation (Randolph and White 2008).

Hampered by the difficulty in obtaining quality soil samples for laboratory element

testing, there is an increasing reliance on using in-situ tests such as the cone, T-bar and

ball penetrometers and vane shear. Although the vane shear can be used to measure both

intact and remoulded strengths, these measurements can only be done at discrete depths

and are more time consuming than other penetrometers. The cone, T-bar and ball

penetrometers are often preferred over the vane shear because of the ability to provide

continuous strength measurement with depth, with the latter two penetrometers being

capable of providing reliable intact and remoulded strengths. However, these

conventional penetrometers are affected by unequal overburden and pore pressures,

which are more critical for the cone penetrometer compared to full-flow penetrometers

(T-bar and ball) (Chung 2005; Low et al. 2010; DeJong et al. 2011). White et al. (2010)

showed that buoyancy can also affect the T-bar measurements especially in soils with

high unit weights. Furthermore, neglecting rate effects, the operative strength inferred

from full-flow penetrometers may be lower than the true intact soil strength, owing to

strain softening (Zhou and Randolph 2007; Zhou and Randolph 2009b).

Typical pipeline routes may span over more than 100 km on seabed with changing

geology. According to ISSMGE (2005) guidelines, in-situ tests and soil samplings

should be conducted up to vertical depths of 1 to 2 m for un-trenched sections and 1 m

below the depth of any pipeline trench, at a spatial frequency of every 0.3 to 5 km

laterally along the proposed pipeline route (depending on the site variability). Between

lateral spacings of boreholes or in-situ soundings, geotechnical parameters are often

predicted by means of interpolation in tandem with geophysical data. Clearly, the

accuracy of this method decreases as the distance between geotechnical test sites

increases.

3.4

This paper describes the development of a novel device for measuring the strength of

the soil, named the vertically oriented penetrometer (VOP). This device was originally

designed for use in small scale centrifuge modelling activities, as a mean for

characterising the strength behaviour of a shallow layer of soil, as it moves past the

penetrometer. As the development of the VOP advanced, it became apparent that this

new penetrometer concept could potentially also improve the accuracy of site

investigation practice geared for submarine pipeline design by means of continuously

dragging the VOP horizontally, perhaps using a machine much like the existing pipeline

ploughs (or the smaller cable ploughs). Alternatively, the device could be mounted on a

Remotely Operated Vehicle (ROV) or Autonomous Underwater Vehicle (AUV), and be

deployed for drags over horizontal regions of varying length scale.

The VOP is used to assess the strength profile of the soil via the same concepts that are

applied to assess the capacity of a laterally loaded pile. Centrifuge validation tests were

conducted in normally consolidated (NC) and heavily overconsolidated (HOC) samples

of kaolin to provide a better understanding of the behaviour of the VOP under pure

horizontal translation. The peak and steady load-displacement results are compared with

existing methods of estimating the ultimate horizontal pressure on a laterally loaded

single pile. In addition, the effect of strain rates and the suitability of the VOP to derive

the soil sensitivity from cyclic tests are also discussed.

Because the VOP resembles a laterally loaded pile, a brief review of the pertinent

literature of this aspect is provided in the next section. The background review also

includes the rate dependent nature of soil behaviour.

BACKGROUND LITERATURE 3.3

Laterally loaded pile 3.3.1

The failure mechanism of a pile under pure horizontal translation (with no rotation) is

shown schematically in Figure 3.1. It can be seen in Figure 3.1(a) that a 2-way wedge

mechanism can exist near the soil surface, involving both active (pile rear) and passive

(pile front) failure wedges which may be idealised as having similar heights (Murff and

3.5

Hamilton 1993). This failure mechanism is usually transient, and upon larger

displacement of the pile, suction is lost at the pile rear causing a gap to form and

consequently, only the passive wedge remains in front of the pile (Figure 3.1(b)). At

depth, an easier mode of soil failure compared to the conical wedge failure (either 2-

way or 1-way) occurs in the form of a plane strain full-flow mechanism. In the case of

the 2-way wedge failure, the plane strain mechanism is mobilised at a shallower depth

compared to when a gap forms behind the pile.

For the case of a gap forming behind the pile, the most common approach of estimating

the ultimate horizontal pressure on the pile, qh is the Matlock (1970) method:

vh u uoˆ'q 3s Js z for f

ˆ ˆz z 3.1

and

h uq 9s for fˆ ˆz z 3.2

where σ'vo is the in-situ effective vertical stress, J is an empirical constant (which

depends on Matlock’s field data) ranging from 0.25 to 0.5 and z is the depth below the

soil surface (z) normalised by the pile diameter (Dp), z/Dp. Equation 3.1 is applicable

down to a normalised critical depth, fz , below which Equation 3.2 applies. For the case

of a uniform soil, fz can be calculated as:

p

fp

u

6Dz

γ' DJ

s

3.3

However, if the soil stratigraphy has a strength gradient, the point of intersection

between the two functions given by Equation 3.1 and Equation 3.2 is taken as fz .

3.6

Neglecting the third term in Equation 3.1, an alternative approach to express the

ultimate horizontal pressure on a pile is (Randolph and Houlsby 1984; Murff and

Hamilton 1993; Jeanjean 2009):

h h uq N s (2-way wedge mechanism/full-flow failure) 3.4

or

h h u voq N s ' (gap at pile rear) 3.5

where Nh is the horizontal bearing capacity factor. Rigorous lower and upper bound

solutions for the case of a full-flow mechanism below the wedge mechanism presented

by Randolph and Houlsby (1984) showed that the bearing factor varies between 9.14

(fully smooth) to 11.94 (fully rough). Above the full-flow depth, lower magnitudes of

Nh apply. To analyse the complete failure mechanism of a laterally loaded pile, the

plasticity solution of Murff and Hamilton (1993) which combines the effects of wedge

failure, shearing at both wedge-soil and pile-soil interfaces, and deep full-flow failure

(similar to Randolph and Houlsby (1984)) may be utilised. As many as four parameters

representing the kinematically admissible failure mechanisms are optimised to yield the

least upper bound qh. Based on several parametric studies on clay with different linear

strength profiles, Murff and Hamilton (1993) proposed an empirical expression for Nh

as a simple fit to the plasticity solutions, which takes the form of:

ξz

h h d h d h

ˆ

sN N N N e

3.6

where Nh-d is the bearing capacity factor at depth (Randolph and Houlsby 1984) and Nh-s

is the bearing capacity factor at the soil surface. ξ depends on the ratio of the strength at

the soil surface, su-m, to the strength gradient, ksu times the pile diameter Dp and was

calibrated using least-square method as:

3.7

ξ 0.25 0.05 if η < 6 3.7

and

ξ 0.55 if η ≥ 6 3.8

where η = su-m/ksuDp.

In the case where a transient suction develops behind the pile, Nh calculated using

Equation 3.6 should be multiplied by two, reflecting the symmetrical active and passive

wedges (Murff and Hamilton 1993). However, the resulting qh should not exceed that

corresponding to the deep failure mechanism:

h h u h d uq 2N s N s 3.9

To account for suction, Jeanjean (2009) used Equation 3.4 and Equation 3.6, with Nh-s

and Nh-d of 8 and 12 respectively.

On the other hand, if a gap forms behind the pile, Murff and Hamilton (1993) suggested

using Equation 3.5 to Equation 3.8 along the full pile length.

Strain rate effects 3.3.2

During undrained shearing, it is well demonstrated that clay exhibits a rate-dependent

strength. This is due to viscous effects where at high strain rates, the strength increase is

proportional to the increase in strain rate, γ (usually characterised for an object moving

through soil as v/D, where v = velocity and D = characteristic dimension of the object

moving). Even for the relatively low velocities at which penetrometers are operated

(typically 1-20 mm/s), the strain rate within the deforming soil is far above that usually

adopted within laboratory testing of soil elements (Randolph 2004). The increase in

resistance (qh) of an object moving through soil at high strain rates can be captured

using the following power law model:

3.8

m

h ref

ref

q q

3.10

where qref is the reference resistance at a reference strain rate, refγ , and m is the rate

dependency parameter. Both the strain rate and reference strain rate can be treated as

characteristics values, given by v/D and (v/D)ref, whilst recognising that around a

moving object there is a spatial variation in strain rate. The viscous effects of common

site investigation tools such as the vane shear (Biscontin and Pestana 2001), T-bar and

ball penetrometers (Low et al. 2008a; Lehane et al. 2009; Boukpeti et al. 2012) are well

quantified. In general, m ranges between 0.05 – 0.17.

EXPERIMENTAL PROGRAMME 3.4

The centrifuge testing programme described in this paper was conducted in the UWA

beam centrifuge (Randolph et al. 1991). These centrifuge validation tests were

conducted in normally consolidated (NC) and heavily overconsolidated (HOC) samples

of kaolin, so that the behaviour of the VOP can be studied across two extreme

conditions (NC and HOC clays).

Experimental apparatus 3.4.1

The vertically oriented penetrometer (VOP) as shown in Figure 3.2 is cylindrical in

shape. This particular prototype version has a total length of 130 mm and is made from

6061 T6-grade Aluminium. A threaded section having a length of 25 mm and a larger

diameter of 7 mm provides a fixed connection to the actuator, thus rendering the VOP

as a ‘cantilevered beam’ when displaced laterally in the soil sample. The VOP is

equipped with 4 levels of strain gauges located above the part of the VOP that is

penetrated into the soil. At each strain gauge level there are 2 strain gauges on opposite

sides. These provide a profile of bending moment above the soil, which can be

processed to determine the total lateral resistance on the embedded portion of the

device. The four pairs of strain gauges, labelled A, B, C and D in Figure 3.2, are located

3.9

at 55 mm, 70 mm, 85 mm and 100 mm respectively from the VOP tip. The strain

gauges are attached on the VOP using Vishay®

AE-10 adhesive kit. They are coated

with a layer of epoxy resin that protects the strain gauges from water ingress. The

unthreaded section is hollow with an outer diameter (OD), DVOP of 4.72 mm and an

inner diameter (ID) of 2.9 mm. Because of this opening, the VOP resembles an open-

ended pile. During testing, the lower ungauged part of the VOP is penetrated to a

particular depth within the soil and then translated horizontally.

Sample preparation 3.4.2

The centrifuge tests were carried out using the UWA kaolin clay. Its mechanical

properties are well studied and described in detail by Stewart (1991) and Lehane et al.

(2009). To prepare the samples, dry kaolin clay powder was first mixed with water at

twice the liquid limit (120%) under vacuum in a barrel mixer. The well-mixed slurry

was then transferred to a strongbox with inner measurements of 650 mm long, 390 mm

wide and 325 mm in height. A 30 mm sand layer was previously placed at the bottom of

the strongbox to establish two-way drainage during sample consolidation. The

centrifuge was then spun for 5 days at 100 g until primary consolidation was achieved

and the final clay sample height was 205mm. The first series of tests was conducted in a

normally consolidated (NC) sample. After completion of the first series of test, a second

series of test were conducted in a heavily overconsolidated (HOC) sample. The HOC

sample was prepared by first bringing the centrifuge to a halt, after which 70 mm of

sample height was trimmed off. Before further testing, the sample was then spun in the

centrifuge at 50 g for 1 day to allow complete excess pore pressure dissipation. The

resulting overconsolidation ratio (OCR) profile of the HOC sample is shown in Figure

3.3. This OCR profile was calculated based on the moisture content measurements taken

at various depths in both NC and HOC samples, which were then used to convert to the

corresponding unit weights and therefore stress history of the HOC sample.

Centrifuge testing procedure 3.4.3

In the centrifuge tests, the VOP threaded section was attached rigidly to a ‘fast’ actuator

capable of a maximum horizontal displacement velocity of 100 mm/s as depicted in

3.10

Figure 3.4. Table 3.1 (NC sample) and Table 3.2 (HOC sample) provide a summary of

the testing programme, depicted according to the chronological order of the tests. The

centrifuge tests were conducted at 100 g and 50 g for the NC and HOC samples

respectively.

T-bar tests for shear strength profiling were conducted before commencement and after

the completion of the VOP tests in both NC and HOC samples. These tests were

conducted at a penetration rate of 1 mm/s using a T-bar with a diameter, DT-bar, of 5 mm

and length, LT-bar, of 20 mm. Each T-bar test consisted of a cyclic penetration and

extraction regime to infer the sensitivity, St, of the soil.

Figure 3.4 shows an overview of the VOP test sequence. All VOP tests were conducted

under displacement control – i.e. moving the VOP horizontally between specified

displacement limits (rather than load limits). Before commencing horizontal translation,

the VOP was installed vertically to the desired embedment at a rate of 0.5 mm/s. All

VOP tests involved many cycles of forward and backward sweeps in order to compare

the resulting variation in resistance – and thus sensitivity, St – to that derived from the

T-bar. The test velocities (v), horizontal sweep distances (u), number of cycles (n), and

total vertical embedment (zT) are shown in Table 3.1 and Table 3.2 for tests in NC and

HOC samples, respectively. The VOP tests in the HOC sample were conducted at

shallower depths to avoid plastic yielding of the VOP. It is noteworthy that, even at the

slowest VOP horizontal test velocity, v, of 1 mm/s, the soil shearing around the VOP is

expected to be fully undrained. Assuming that the coefficient of consolidation of the

UWA kaolin, cv is 2.6 m2/year (Stewart 1991), the resulting dimensionless velocity, V

(= vDVOP/cv) is 57, well above the undrained limit of approximately 10 based on

variable rate T-bar penetration studies reported by House et al. (2001) and Lehane et al.

(2009).

To avoid potential boundary effects, both T-bar and VOP tests were conducted at a

minimum distance of 10DVOP from the edges of the strongbox. An ample spacing of at

least 4DVOP was also allowed between test sites to avoid overlapping of the T-bar and

VOP failure mechanism ‘influence zones’.

3.11

T-BAR TESTS: SHEAR STRENGTH AND SENSITIVITY 3.5

Figure 3.5 depicts the strength derived from initial T-bar penetrations, su-in, in both the

NC and HOC samples. The strength profile can be derived directly from the T-bar

penetration resistance, qT-bar, using the usual total stress approach:

T baru

T bar

qs

N

3.11

It is common practice to adopt a constant bearing factor, NT-bar of 10.5 (reflecting a T-

bar of intermediate roughness) similar to that of a laterally loaded pile at depth

(Randolph and Houlsby 1984). However, much like the case of a strip footing, lower

magnitudes of NT-bar are required to account for the different failure mechanism near the

surface. The transition of surface heave failure to the full-flow failure mechanism

depends on the strength ratio, su/γ′DT-bar (where γ′ = soil effective unit weight, DT-bar =

T-bar diameter). The resulting variation of NT-bar can be estimated based on the iterative

procedure outlined by White et al. (2010). The su-in profiles shown in Figure 3.5 have

been adjusted for surface and buoyancy effects, and the average depths of the full-flow

mechanism for the NC and HOC samples are 1.5DT-bar and 4.6DT-bar respectively. The

su-in profiles for the NC and HOC samples can be represented using a simple linear

relationship:

u in u m sus s k z (kPa) 3.12

The magnitudes of su-m and ksu for both NC and HOC samples are summarised in

Table 3.3.

The results of the full T-bar cyclic penetration and extraction profiles for NC and HOC

samples are plotted in Figure 3.6(a) and Figure 3.7(a) respectively. It should be noted

that the NT-bar shallow correction can only be used for the penetration phase as an

extraction process involves a different failure mechanism, and is influenced by any

cavity remaining from the penetration stage. To avoid unnecessary complexity, a

constant NT-bar factor of 10.5 is adopted for the cyclic tests. Furthermore, the cyclic test

3.12

depths in NC and HOC samples are located in the deep-failure mechanism zones, where

NT-bar = 10.5 applies. The continuous degradation in strength from continuous cycles of

penetration and extraction can be quantified using the degradation factor, defined as the

shear strength at a particular cycle, su-n, normalised by the initial T-bar penetration

strength, su-in. These strength values are extracted from the mid-depth of each cyclic T-

bar stage.

The resulting decrease in degradation factor with cycle number derived from all the T-

bar tests are shown in Figure 3.6(b) (NC sample) and Figure 3.7(b) (HOC sample). The

cycle number, n, is calculated as recommended by Randolph et al. (2007), where the

first penetration is designated as n = 0.25 and the n for subsequent extractions and

penetrations, cumulatively calculated in increments of 0.5 (n = 0.75, 1.25, etc.). In this

way, a cycle number of 1 corresponds to the level of strain and remoulding experienced

by the soil after the disturbance caused by two passages of the T-bar. As the T-bar is

passing, the local soil has experienced, on average, only half of the deformation that

will ultimately be caused by the passing of the bar.

From the average degradation curves in Figure 3.6(b) (NC sample) and Figure 3.7(b)

(HOC sample), the sensitivity, St, of each sample can be estimated from the inverse of

the average degradation factors once the degradation curve has stabilised (at about n >

8.25 – 10.75). The resulting values of St are shown in Table 3.3. This approach neglects

any potential changes in NT-bar during the cyclic phase common for conventional full-

flow penetrometers (Yafrate et al. 2009; Zhou and Randolph 2009a; Low et al. 2010).

Generally, as depicted in Figure 3.5 to Figure 3.7 and Table 3.3, no systematic

differences in strength profiles and sensitivities between T-bar tests are observed in both

NC and HOC samples. This indicates the sample uniformity laterally across the sample,

and over the testing period.

RESULTS FOR VOP EXPERIMENTS 3.6

Typical total horizontal load (Fh) – normalised displacement (u/DVOP, with u the VOP

displacement and DVOP the VOP diameter) responses for tests in NC and HOC samples

3.13

are shown in Figure 3.8. These values of Fh are derived by comparing the bending

moment responses of any 2 pairs of strain gauges (Figure 3.2). The full cyclic test

regime generally consists of a peak load, which decreases to a steady value upon large

displacement. Similar to a T-bar cyclic test response, Fh decreases further to a fully

remoulded value with increasing cycles of forward and backward sweeps. In order to

reliably estimate the strength, su using the VOP, it is important to understand the

variation of the VOP bearing factor, Nh with depth during the first forward sweep of the

VOP. The back analyses of Nh corresponding to the steady load will be studied first,

followed by the Nh for peak load, using only the VOP tests at a horizontal velocity of

1 mm/s. These tests are chosen because the operative strain rate is closest to that for the

T-bar tests, thus allowing the T-bar strength profile to be used directly. The peak and

steady loads are calculated using the various methods outlined in Section 3.3.1. The

contribution of the shearing resistance at the base of the VOP (Figure 3.1) is neglected,

since the base area of the VOP is small relative to the shaft. This section also explores

the suitability of the VOP to infer the soil sensitivity and addresses the effects of strain

rates on the measured VOP loads.

Bearing capacity factors – NC sample 3.6.1

Figure 3.9 shows the surficial soil condition surrounding the VOP when displaced

horizontally in NC sample. It is evident that a small gap is visible near the mudline at

the rear of the VOP. This suggests that a 1-way wedge mechanism (Figure 3.1(b)) may

be confined to a very shallow depth. Moreover, the peak to steady load transition

observable in Figure 3.8 for the test in NC sample (VOP_NC4) is often associated with

the loss of suction (due to the formation of a rear gap) at the pile/VOP rear (e.g. Murff

and Hamilton 1993; Zhang et al. 2011). For this reason, the estimated total steady load

using the Matlock (1970) (Equation 3.1 and Equation 3.2) and Murff and Hamilton

(1993) (Equation 3.5 and Equation 3.6) methods are compared to that measured from

the VOP tests (v = 1 mm/s) in the NC sample in Figure 3.10, assuming that a gap forms

behind the VOP. For the estimate using the Murff and Hamilton method, the near

surface VOP bearing factor, Nh-s is taken as 2 (following Zhang et al. 2011) and the full-

flow bearing factor, Nh-d is taken as 10.5 (corresponding to a T-bar full-flow

mechanism). The resulting horizontal pressure (qh) distributions with VOP depths are

3.14

plotted in Figure 3.11. The steady load calculated using the Matlock method is limited

to the ultimate load for deep failure mechanism at depths below 2.22DVOP (Figure 3.11).

As depicted in Figure 3.10, the Matlock method appears to predict the steady load better

than the Murff and Hamilton method, where the latter overpredicts the steady load by

more than 40%. As shown in Figure 3.11, the Murff and Hamilton method (Equation

3.5) exceeds the pressure distribution calculated, using a constant Nh-d = 10.5 at z ≥ 1,

causing the overestimation of the steady load. This was also noted in the back

calculation of the horizontal pressure on a laterally displaced pile reported by Zhang et

al. (2011). Beyond fz ,

the deep flow mechanism is an easier mode of failure compared

to the wedge mechanism, and Equation 3.5 should be modified as:

h h u vo h d uq N s ' N s 3.13

where Murff and Hamilton’s method for calculating Nh is retained (Equation 3.6 to

Equation 3.8). This modified method (Modified M&H) results in significantly better

estimates of the steady load (Figure 3.10). From Figure 3.11, it is clear that the deep

failure mechanism dominates when the VOP is pushed laterally in NC soil. This is

supported by the presence of a shallow gap at the VOP rear (Figure 3.9).

Figure 3.12 shows the estimated peak load using the methods of Matlock (1970),

Jeanjean (2009) and a constant Nh-d = 10.5. The 1-sided wedge mechanism of

Matlock (1970) underestimates the peak loads by more than 30%. Numerous other

studies (Murff and Hamilton 1993; Jeanjean 2009; Zhang et al. 2011) have also shown

that the Matlock method tends to underestimate both peak and steady loads – although

this is perhaps an acceptable practical situation, from the perspective of pile design.

Surprisingly however, both the Jeanjean and a constant Nh-d = 10.5 methods also

underestimate the peak loads, even when the former was calibrated for peak load

conditions. In tandem with the visual evidence shown in Figure 3.9 (shallow gap), it is

hypothesised that the peak to steady state transformation in NC sample is not

predominantly governed by the loss of suction at the VOP rear.

An alternative explanation for the high peak lateral resistance is consolidation of the soil

around the VOP following penetration. However, Figure 3.13 shows no relationship

3.15

between the increase in peak loads with consolidation time factor, T = cht/Deq2 for all

the tests in the NC sample where t is the elapsed time from the end of the vertical

installation of the VOP to the start of the horizontal translation. As recommended by

Randolph (2003), the VOP equivalent diameter (to account for the open end condition

of the pile), is taken as Deq ≈ 2VOP wD t , where tw is the VOP wall thickness. The

horizontal coefficient of consolidation, ch is taken as 2cv (Randolph and Hope 2004).

The absence of a unique trend in Figure 3.13 suggests that the VOP peak load in the NC

sample is not associated with the increase in su following a ‘setup’ period common to

those found in driven piles (Randolph et al. 1979) and anchors (Richardson et al. 2009).

The potential factor that causes this high peak load compared to the steady load is

revisited later in this paper (Section 3.6.5) once the rate effects of the VOP are

quantified.

Bearing capacity factors – HOC sample 3.6.2

In HOC sample, it can be seen that a gap forms at the VOP rear Figure 3.14. Figure 3.15

shows the comparison between the estimated and measured total steady loads for the

VOP tests in the HOC sample conducted at v = 1 mm/s using the Matlock (Equation 3.1

and Equation 3.2) and Murff and Hamilton (Equation 3.5 and Equation 3.6) methods,

assuming that a gap forms behind the VOP (consistent to that shown in Figure 3.14).

The pressure distributions calculated using these methods are shown in Figure 3.16.

Similar to that in the NC sample, the Matlock method also underestimates the steady

loads for all the 3 tests in the HOC sample (Figure 3.15). The Matlock method estimates

that the deep failure mechanism is mobilised at a much larger depth in the HOC case as

shown in Figure 3.16, and may be the cause of the lower estimated steady loads. As

observed in Figure 3.16, the pressure distribution calculated using the ordinary Murff

and Hamilton method intersects that calculated using a constant Nh-d = 10.5 at fz = 4,

implying that above this normalised depth, the ordinary Murff and Hamilton approach

applies and below this depth, the ultimate horizontal pressure should be calculated using

Nh-d = 10.5. In Figure 3.15, the modified Murff and Hamilton approach resulted in no

improvement in the estimated steady load for test VOP_HOC1 (zT/DVOP = 2.12, where

zT is the total embedment), and only modest improvements for tests VOP_HOC2

3.16

(zT/DVOP = 4.23) and VOP_HOC3 (zT/DVOP = 5.3). This is because the ordinary Murff

and Hamilton method is not applicable at only 5.4% and 24.5% of the total embedment

lengths (below 4DVOP) for tests VOP_HOC2 and VOP_HOC3 respectively, which are

associated to the easier flow-round failure mode.

For the case where a peak load is mobilised in the HOC sample, symmetrical active and

passive wedges are assumed to form near the soil surface as the VOP is displaced

horizontally. A comparison of the measured total and estimated peak loads using the

Jeanjean, Murff and Hamilton and Matlock methods is shown in Figure 3.17 for tests at

v = 1 mm/s in HOC sample. For the Murff and Hamilton method, the variation of Nh is

calculated according to Equation 3.6 to Equation 3.8, and the resulting qh distribution

with VOP depth calculated according to the limiting condition shown in Equation 3.9.

The resulting estimates using the Murff and Hamilton method are closer to the

measured peak loads compared to the Jeanjean method where the former has an average

of less than 1% error whereas the latter has about 3% average error. As expected, the

Matlock method (suited for only a 1-wedge mechanism) greatly underpredicts the peak

loads.

Considering the visual observation shown in Figure 3.14 and the results for the

estimated steady loads (Figure 3.15 and Figure 3.16) and peak loads (Figure 3.17), it is

highly likely that the peak to steady state transformation of the VOP load-displacement

response in the HOC sample is governed mostly by the formation of a gap at the VOP

rear.

Soil sensitivity 3.6.3

The degradation factor vs. cycle number responses for the VOP cyclic tests in the NC

and HOC samples at different velocities are each plotted in Figure 3.18 (solid symbols –

zT = 30 mm, open symbols – zT = 45 mm) and Figure 3.19 (solid symbols – zT =

10 mm, open symbols – zT = 20 mm, dotted lines – zT = 25 mm). In these figures, the

degradation factor is defined as the horizontal load at a particular cycle number, Fh-n

(subscript n denoting cycle number), normalised by the horizontal load from the first

VOP forward sweep (n = 0.25), Fh-0.25. The degradation factor can be treated as

3.17

analogous to the T-bar degradation factor, assuming (as for the T-bar) that the Nh (or

NT-bar) profile is unchanged through the cycles. These loads are extracted from the mid-

point of each cycle, when the load had reached a steady state, and the cycle numbering

system follows that of the T-bar (Section 3.5). The average T-bar degradation curves for

the NC and HOC samples are also plotted in the corresponding figures for comparison.

As shown in Figure 3.18, the degradation responses of the VOP cyclic tests and the T-

bar in NC sample are very similar, except for the VOP cyclic tests at v = 1 mm/s

(VOP_NC1 and VOP_NC2). At v = 1 mm/s, the time between the VOP passing the

mid-point of the cyclic distance is 40 s (prototype = 4.63 days). This delay causes the

reconsolidation effect to dominate over the soil remoulding and is believed to be the

reason behind the slightly increasing degradation factors at n > 7.75. This

reconsolidation effect is discussed further in Sahdi (2012) using a framework couched

in critical soil mechanics terms. On the other hand, compared to the T-bar, the

degradation curves of the VOP cyclic tests in HOC sample (Figure 3.19) show

significantly lower degradation factors and highly brittle responses, where the load is

only about 20 - 40% of the initial load during the second pass of the VOP (n = 0.75).

A comparison of the sensitivity, St, derived from the VOP and the T-bar cyclic tests in

NC and HOC samples is depicted in Figure 3.20. The VOP St values are calculated from

the inverse of the average degradation factors once the degradation curves have

stabilised. Generally, in the NC sample, the VOP St is close to the T-bar St (Figure 3.20

(a)), except for the tests affected by reconsolidation (v < 3 mm/s). However, in the HOC

sample, the VOP cyclic tests yielded St values that are much higher than those derived

from the T-bar tests. In the NC sample, the full-flow mechanism dominates along the

VOP length, which means that the VOP acts essentially as a T-bar during the cyclic

tests, with flow-round occurring at most depths, producing similar levels of remoulding

in the surrounding soil. In the HOC sample however, the wedge mechanism extends up

to 4DVOP, and a gap forms instantly after the first VOP forward sweep (Figure 3.14),

resulting in a higher apparent St with decreasing VOP total embedment (zT)

(Figure 3.20 (b)).

3.18

It should be recognised that in practise, any VOP cyclic test is perhaps only practical

within a limited horizontal distance, given the difficulty in dragging the VOP using a

suitable test station/platform. This limited distance is also important to avoid the effects

of strength hardening of the soil attributed to reconsolidation (White and Hodder 2010;

Rismanchian et al. 2011).

Strain rate effects 3.6.4

The viscous effects observed during the VOP tests in both the NC and HOC samples are

quantified in Figure 3.21 where the data for the tests in the NC sample are plotted as

open symbols, while the test data for the HOC sample are represented by solid symbols.

The peak (Fhp), steady (Fhs) and remoulded (Fhr) loads are normalised by the respective

reference loads (represented as subscript ref) at a reference velocity of 1 mm/s. Due to

reconsolidation effects for the cyclic tests at v = 1mm/s in NC sample (Section 3.6.3),

the Fhr-ref requires correction. The magnitudes of Fhr-ref are adjusted by dividing the

respective Fhs with the average St derived from VOP cyclic tests at v ≥ 3 mm/s. The

entire normalised peak, steady and remoulded loads are fitted with the power law model

(Equation 3.10) and the respective best fit m (rate dependency parameter) values are

shown in Figure 3.21.

Although the data in Figure 3.21 are somewhat scattered, especially at higher velocities,

the changes in loads with increasing velocities generally follow the same trends in both

NC and HOC samples (reflected by the similar best-fit m in each sample). This implies

that m is essentially independent of the water content of the soil as outlined by Boukpeti

et al. (2012). It can also be seen from Figure 3.21(b) and Figure 3.21(c) that for both

intact (NC and HOC samples) and remoulded soils (NC sample), m is essentially

identical (at 0.08). This is consistent with the findings of Low et al. (2008a) for intact

and remoulded Burswood clay. Surprisingly, the rate dependency of the peak loads for

both NC and HOC samples (m = 0.14) is higher than that for the steady and remoulded

loads. This may be linked to the transition in failure mode from a 2-way (associated

with the peak resistance) to a 1-way wedge mechanisms (associated with the steady

resistance) after peak resistance has been reached. In particular, the strain rate

dependency parameter, m associated with the rear wedge, which experiences mainly

3.19

dilation (with associated reduction in pore pressure – i.e. suction) may be different to

that associated with the frontal wedge, which experiences compression (with associated

increase in pore pressure).

Peak load in NC soil accounting for softening and rate effects 3.6.5

Closer inspection of Figure 3.8 reveals that the loss of peak resistance in HOC sample is

much more sudden (stabilising at about 0.5DVOP) compared to that in the NC sample

(stabilises at about 2DVOP). This more gentle transition from the peak to the steady load

in the NC sample is very similar to the response observed numerically when a T-bar is

wished-in-place in softening soil, then displaced (Zhou and Randolph 2007; Zhou and

Randolph 2009b). In these studies, the soil around the T-bar softened according to the

accumulated plastic strain. The initial resistance when the T-bar was first moved was

controlled by the intact soil strength, but the steady resistance on the T-bar was lower,

being equal to an operative soil strength in which the soil around the bar has already

undergone varying levels of strain and softening.

The VOP tests can be considered to be analogous to these numerical simulations. When

the VOP first begins moving laterally, the steady flow mechanism – in which the

surrounding soil is partly softened – is not established (although some lesser amount of

strain will have been induced during the penetration process). Consequently, the initial

VOP resistance represents the ‘true’ intact strength of the soil, which is not detectable

during steady T-bar penetration. To confirm this inference, a correction can be applied

to the estimated steady resistance during VOP translation and compared with the

measured peak loads using a suitable bearing factor and a corrected ‘true’ strength, su-c.

To obtain this new corrected strength (su-c), the steady strength – given by su-in (strength

from the first T-bar penetration) – must be modified by a multiplicative factor, denoted

as Cs. To account for rate effects, the su-c (only for VOP tests at v > 1 mm/s) should be

augmented for viscous effects. The overall corrective procedure can be written as:

m

u c s u in

ref

s C s

3.14

3.20

The strain softening correction factor (CS) is equal to the ratio of the rate independent

‘true’ intact strength, su-int to su-in. For a T-bar test in the UWA kaolin, Hodder et al.

(2010) showed that Cs is approximately 1.6. The rate dependency parameter, m is taken

as 0.14 (for peak loads).

Figure 3.22 compares the calculated (by correcting the strength via Equation 3.14 and

assuming a constant Nh-d = 10.5) and measured peak loads for all the VOP tests in the

NC sample. The new corrected estimates are in excellent agreement with the measured

peak loads. This suggests that in NC soil conditions, where flow-round occurs

throughout most of the VOP depth, the initial VOP resistance indicates the ‘true’ intact

soil strength, as opposed to the ‘steady’ flow T-bar strength. In this way, the VOP

provides additional information on the softening and remoulding behaviour of the soil,

compared to the conventional T-bar.

POSSIBLE REFINEMENTS OF THE VOP 3.7

Back-analysis and bearing factors 3.7.1

From the back analyses to determine suitable VOP bearing factors for both peak and

steady states in NC and HOC samples, it is evident that the estimated loads using the

empirical equation proposed by Murff and Hamilton (1993) (albeit modified for the case

where a gap forms behind the VOP) yielded the closest estimates to the measured loads.

However, this approach requires prior knowledge of the strength profile of the seabed

soil, and may hamper interpretation of the VOP shear strength when applied in field

situations due to the uncertainty in selecting suitable bearing factors. This limitation

also applies to the back-analysis of near-surface T-bar penetrometer resistance (White et

al. 2010). This is inevitable because the strength profile invariably affects the bearing

factor profile close to the soil surface.

Apart from very small displacements, the predominant mode of near-surface failure

when the VOP of the scale tested here is displaced along a horizontal route, is the 1-

sided wedge mechanism (Figure 3.1 (b)) although for normally consolidated to lightly

overconsolidated clays, the wedge mechanism may be restricted to above 1 - 2DVOP

3.21

only. Since the VOP is forced to overcome both the soil self-weight and interface shear

strength in the 1-sided wedge mechanism, the depth at which the full-flow mechanism is

mobilised, fz , should depend on the normalised strength to weight ratio of the soil,

su/γ'DVOP.

This effect has been recognised previously for a shallow T-bar penetration (White et al.

2010) – which is a plane-strain failure mechanism – and spudcan penetration (Hossain

et al. 2005; Hossain et al. 2006) into clay – which is an axisymmetric mechanism.

Inspection of the equation proposed by Matlock (1970) (Equation 3.3), shows that fz

can be expressed directly as a function of su/γ'DVOP. The fz back calculated from tests

in NC and HOC samples shown in Figure 3.11 and Figure 3.16 respectively are plotted

in Figure 3.23 as a function of su/γ'DVOP. The full-flow depth – strength ratio

relationships for laterally loaded piles (Matlock 1970), T-bar (White et al. 2010) and

spudcan (Hossain et al. 2005; Hossain et al. 2006) are also included in the same plot for

comparison. Although the centrifuge data is currently very limited, the two test data

points appear to follow the same trend as the other plotted lines.

Also, the parameter ξ in Equation 3.6 is dependent on su-m/ksuDVOP. For NC (with zero

surface strength intercept) and uniform clays the ξ is constant at 0.25, implying that the

rate of increase in the bearing factor Nh (Equation 3.6) is the same regardless of

su/γ'DVOP. Therefore, it is also likely that ξ may also be a function of su/γ'DVOP instead. If

proper relationships of fz - su/γ'DVOP and ξ - su/γ'DVOP can be established, a framework

for interpreting the soil strength using suitable bearing factors may be possible.

However, in offshore hydrocarbon project areas like the Gulf of Mexico and West

Africa, the seabed sediments tend to be normally consolidated or lightly

overconsolidated (Gemenhardt and Focht 1970; Endley et al. 1981; Puech et al. 2005),

implying that the wedge mechanism will be limited to a very shallow depth, and a

constant factor of 10.5 may be used. In these soils, interpretation of a VOP test would

be relatively simple.

3.22

Instrumentation 3.7.2

Unlike a T-bar penetrometer – which provides a continuous profile of resistance with

depth – the VOP in its current form only provides bending moment data that indicates

the net force and effective depth resulting from the profile of lateral resistance. This

information can be combined from VOP tests at different penetration depths, adding

data that can be used to help the back-calculation of the strength profile. Alternatively,

the current VOP instrumentation may be improved by fitting multiple bending strain

gauges throughout the depth of the VOP including the portion embedded in the soil. The

soil resistance can then be determined by double differentiation of the bending moment

profile. However, the use of instrumentation above the mudline allows for varying wall

thickness and diameters – to maximise the strain gauge sensitivity. Additional gauging

could be added to measure the axial load during penetration, providing further

information to support assessments of the intact soil strength.

Also, an important parameter for pipeline design is the determination of the soil

coefficient of consolidation cv. This can possibly be achieved by installing pore pressure

transducers on the VOP, and the cv may be back calculated by comparing the dissipation

of the measured excess pore pressure with the consolidation solution of a driven pile in

clay given by Randolph and Wroth (1979). On the other hand, a twitch test often used

for the cone and full-flow penetrometers (e.g. Randolph and Hope 2004; Chung et al.

2006) may be utilised to infer cv. However, the ability of the VOP to infer cv should be

experimentally validated.

CONCLUSIONS 3.8

A series of centrifuge tests in normally consolidated (NC) and highly overconsolidated

(HOC) clay samples aimed at studying the performance of a novel penetrometer is

described in this paper. This new penetrometer, named the vertically oriented

penetrometer (VOP), provides an alternative tool for assessing the near-surface strength

of the seabed. An ambitious goal would be to gather continuous data on the lateral

variation in seabed shear strength along a proposed pipeline route by means of dragging

the VOP horizontally. This would lead to greater site investigation accuracy compared

3.23

to the current practice of performing geotechnical investigation at largely spaced

discrete areas. A more modest aim is to provide better near-surface strength data at

discrete locations in either centrifuge samples or in the field, compared to the T-bar

penetrometer. The centrifuge test programme involved evaluating suitable bearing

factors for back-analysis of the VOP, a parametric study of rate effects, and also a study

of the suitability of the VOP to derive the soil sensitivity via multiple cycles of forward

and backward sweeps.

For the first forward sweep of the VOP for tests in both NC and HOC samples, a peak

load first appears followed by a steady state load at larger displacements. The steady

loads for the tests in NC and HOC samples were predicted well using established

concepts for a laterally loaded pile. Back calculation of the steady loads in NC and HOC

samples revealed that the full-flow mechanism is achieved at 1 and 4 VOP diameters

respectively, and the depth of full-flow is found to be a function of the soil strength su,

the effective soil unit weight, γ' and the VOP diameter, DVOP, i.e. su/γ'DVOP. The loss of

peak load for tests in HOC sample is linked to the loss of suction at the VOP rear. Due

to the dominating effect of the flow-round mechanism for the VOP tests in NC sample,

it was found that the peak to steady load transition is caused by strain softening. This

means that the measured peak load may provide a new manner in which the in-situ

strength can be deduced instead of the strain soften strength usually obtained from the

ball or T-bar penetrometers.

A parametric study into viscous effects on the VOP resistance in both the NC and HOC

samples showed that the rate dependency of the intact soil is similar to that in a

remoulded soil, but higher for the peak load. The back-calculated rate dependency

parameters were consistent with other published results for kaolin clay.

For the NC sample, the sensitivity of the soil derived from the VOP cyclic tests showed

good agreement with the T-bar sensitivity. However, a gap which forms at the VOP rear

for in the HOC sample prevents proper insight into the soil sensitivity, because the

failure mechanism changes between the cycles, as well as the soil strength.

At this stage, the VOP used in this centrifuge study represents only a prototype of a

potential field instrument. If a field scale version is to be developed, the dimensions and

3.24

instrumentation could be optimised and a suitable deployment platform would need to

be developed. However, the results presented here show that the instrument has a

comparable capability to the T-bar penetrometer to gather information on near-surface

soil strength. An advantage of the VOP over the T-bar is that it avoids buoyancy effects

related to vertical entry and exit from the seabed. A disadvantage is that for the current

level of instrumentation, the VOP back-analysis is more complex, since only the

resultant force and its effective location are measured directly.


This research was supported by the Minerals and Energy Research Institute of Western

Australia, and by BP, BHP Billiton, Chevron, Petrobras, Shell and Woodside. The

financial support of all the participants is gratefully acknowledged. The authors are

grateful to James Schneider, Don Herley, Phil Hortin and Shane De Catania for their

technical assistance. The first author is also grateful for the financial support received

from the Ministry of Higher Education Malaysia (MOHE) and Universiti Malaysia

Sarawak (UNIMAS).

3.25

Table 3.1. T-bar and VOP centrifuge test details in NC sample (g-level: 100 g)

Test

reference

VOP total

embedment,

zT (mm)

VOP

horizontal

displacement,

u (mm)

T-bar total

vertical

displacement,

(mm)

VOP

horizontal

cyclic

distance

(mm)

T-bar

vertical

cyclic

depth

(mm)

Test

velocity,

v

(mm/s)

Number

of cycles,

n

T-bar_NC1 - - 100 - 18-48 1 10.75

VOP_NC1 30 40 - 0-40 - 1 9.75

VOP_NC2 45 40 - 0-40 - 1 9.75

VOP_NC3 30 40 - 0-40 - 3 9.75

VOP_NC4 45 40 - 0-40 - 3 9.75

VOP_NC5 30 100 - 0-100 - 10 9.75

VOP_NC6 45 100 - 0-100 - 10 9.75

VOP_NC7 30 100 - 0-100 - 30 9.75

VOP_NC8 45 100 - 0-100 - 30 9.75

T-bar_NC2 - - 60 - 29-49 1 10.75

Table 3.2. T-bar and VOP centrifuge test details in HOC sample (g-level: 50 g)

Test

reference

VOP

embedment,

zT (mm)

VOP

horizontal

displacement,

u (mm)

T-bar total

vertical

displacement,

(mm)

VOP

horizontal

cyclic

distance

(mm)

T-bar

vertical

cyclic

depth

(mm)

Test

velocity,

v

(mm/s)

Number

of cycles,

n

T-bar_HOC1 - - 60 - 31-51 1 11.75

VOP_HOC1 10 40 - 0-40 - 1 9.75

VOP_HOC2 20 40 - 0-40 - 1 9.75

VOP_HOC3 25 40 - 0-40 - 1 9.75

VOP_HOC4 10 40 - 0-40 - 3 9.75

VOP_HOC5 20 40 - 0-40 - 3 9.75

VOP_HOC6 25 40 - 0-40 - 3 9.75

VOP_HOC7 10 100 - 0-100 - 30 9.75

VOP_HOC8 20 100 - 0-100 - 30 9.75

VOP_HOC9 25 100 - 0-100 - 30 9.75

T-bar_HOC2 - - 60 - 32-52 1 11.75

T-bar_HOC3 - - 60 - 31-51 1 11.75

3.26

Table 3.3. Parameters derived from T-bar tests in NC and HOC samples

Sample

Average shear

strength intercept at

soil surface, su-m (kPa)

Average shear strength

gradient, ksu (kPa/mm)

Average shear strength

gradient, ksu (kPa/m) -

prototype scale

Average

sensitivity,

St

NC 0 0.09 0.9 2.3

HOC 3.815 0.12 2.4 2.2

3.27

(a)

(b)

Figure 3.1. Failure mechanism of a pile in pure horizontal translation at: (a) small displacement;

(b) large displacement

Displacement

Frictional resistance

Passive wedgeActive wedge

Full-flow mechanism

Full-flow mechanism

Displacement

Passive wedgeGap

Frictional resistance

3.28

Figure 3.2. (a) Photo of VOP; (b) Schematics of VOP

Figure 3.3. Overconsolidation ratio profile of HOC sample

0

40

80

120

160

1 10 100

Sa

mp

le d

ep

th (

mm

)

Overconsolidation ratio, OCR

N = 50 g

3.29

Figure 3.4. VOP beam centrifuge test setup

‘Fast’ actuator

VOP

(2) VOP displacement

Strongbox

(1) VOP embedment

NOTE: VOP operation -

(1) followed by (2)

3.30

(a)

(b)

Figure 3.5. Undrained shear strength from T-bar tests in: (a) NC sample; (b) HOC sample

0

10

20

30

40

50

60

70

0 2 4 6

Dep

th b

elo

w m

ud

lin

e, z (

mm

)

Undrained shear strength, su-in (kPa)

T-bar_NC1

T-bar_NC2

su-in = 0.09zT-bar kPa

N = 100 g

0

10

20

30

40

50

60

70

0 5 10 15

Dep

th b

elo

w m

ud

lin

e, z (

mm

)


T-bar_HOC1

T-bar_HOC2

T-bar_HOC3

su-in = 3.815 + 0.12zT-bar kPa

N = 50 g

3.31

(a)

(b)

Figure 3.6. Cyclic T-bar test results in NC sample: (a) Cyclic tests profiles; (b) T-bar strength

degradation factor extracted from mid-point of cyclic test range

0

20

40

60

80

100

-8 -6 -4 -2 0 2 4 6 8 10

Dep

th b

elo

w m

ud

lin

e, z (

mm

)

Undrained shear strength, su (kPa)

T-bar_NC1

T-bar_NC2

N = 100 g

0.2

0.4

0.6

0.8

1

0 2 4 6 8 10 12

Deg

ra

da

tion

fa

cto

r, s

u-n

/su

-in

Cycle number

T-bar_NC1

T-bar_NC2

Average

St = 2.3

3.32

(a)

(b)

Figure 3.7. T-bar cyclic tests results in HOC sample: (a) Cyclic tests profiles; (b) T-bar strength

degradation factor extracted from mid-point of cyclic test distance

0

10

20

30

40

50

60

70

-8 -4 0 4 8 12

Dep

th b

elow

mu

dli

ne,

z (

mm

)


T-bar_HOC1

T-bar_HOC2

T-bar_HOC3

N = 50 g

0.2

0.4

0.6

0.8

1

0 2 4 6 8 10 12

Deg

ra

da

tio

n f

acto

r, s

u-n

/su

-in

Cycle number

T-bar_HOC1

T-bar_HOC2

T-bar_HOC3

Average

St = 2.2

3.33

Figure 3.8. Typical load – displacement profile for test in NC (test VOP_NC4) and HOC (test

VOP_HOC6) samples

Figure 3.9. Soil condition during VOP horizontal translation in NC sample – shallow gap visible at

the VOP rear

-4

0

4

8

12

0 1 2 3 4 5 6 7 8 9

Ho

riz

on

tal lo

ad

, F

h(N

)

Horizontal displacement, u/DVOP

VOP_NC4

VOP_HOC6Peak

Steady

Remoulded

3.34

Figure 3.10. Comparison of calculated and estimated steady load for VOP tests (v = 1mm/s) in NC

sample

Figure 3.11. Calculated horizontal pressure distribution along VOP length in NC sample (steady

load, v = 1mm/s)

0.7

0.8

0.9

1

1.1

1.2

1.3

1.4

1.5

VOP_NC1 VOP_NC2

Est

ima

ted

lo

ad

/Measu

red

lo

ad

Test name

Matlock (1970)

Murff and Hamilton (1993)

Modified M&H

0

0.5

1

1.5

2

2.5

0 1 2 3 4 5 6 7 8 9

No

rma

lise

d d

epth

, z/

DV

OP

Horizontal pressure, qh (kPa)

Matlock (1970) - shallow

Matlock (1970) - deep


Nh-d = 10.5Nh-d = 10.5

2.22

1

3.35

Figure 3.12. Comparison of calculated and estimated peak load for VOP tests (v = 1mm/s) in NC

sample

Figure 3.13. Effect of consolidation time on the peak load in NC sample (all tests)

0

0.2

0.4

0.6

0.8

1

1.2

VOP_NC1 VOP_NC2

Est

ima

ted

lo

ad

/Measu

red

lo

ad

Test name

Nh-d = 10.5

Jeanjean (2009)

Matlock (1970)

Nh-d = 10.5

1.4

1.6

1.8

2

2.2

1 2 3 4 5 6

Pea

k lo

ad

/Ste

ad

y lo

ad

Time factor, T = cht/Deq2

30 mm embedment

45 mm embedment

3.36

Figure 3.14. Soil condition during VOP horizontal translation in HOC sample – gap visible

Figure 3.15. Comparison of calculated and estimated steady load for VOP tests (v = 1mm/s) in

HOC sample

0.4

0.5

0.6

0.7

0.8

0.9

1

1.1

VOP_HOC1 VOP_HOC2 VOP_HOC3

Est

ima

ted

lo

ad

/Measu

red

lo

ad

Test name

Matlock (1970)


Modified M&H

3.37

Figure 3.16. Calculated horizontal pressure distribution along VOP length in HOC sample (steady

load, v = 1mm/s)

Figure 3.17. Comparison of calculated and estimated peak load for VOP tests (v = 1mm/s) in HOC

sample

0

2

4

6

8

10

0 20 40 60 80 100 120

Norm

ali

sed

dep

th, z/

DV

OP

Horizontal pressure, qh (kPa)

Matlock (1970) - shallow

Matlock (1970) - deep


10.5Nh-d = 10.5

4

9

0

0.2

0.4

0.6

0.8

1

1.2

VOP_HOC1 VOP_HOC2 VOP_HOC3

Est

ima

ted

lo

ad

/Measu

red

lo

ad

Test name

Jeanjean (2009)


Matlock (1970)

3.38

(a)

(b)

(c)

(d)

Figure 3.18. Load degradation response for test in NC sample at: (a) v = 1 mm/s; (b) v = 3 mm/s;

(c) v = 10 mm/s; (d) v = 30 mm/s

0.2

0.4

0.6

0.8

1

0 2 4 6 8 10 12

Fh

-n/F

h-0

.25

Cycle number

VOP_NC1

VOP_NC2

T-bar

Affected by reconsolidation0.2

0.4

0.6

0.8

1

0 2 4 6 8 10 12

Fh

-n/F

h-0

.25

Cycle number

VOP_NC3

VOP_NC4

T-bar

0.2

0.4

0.6

0.8

1

0 2 4 6 8 10 12

Fh

-n/F

h-0

.25

Cycle number

VOP_NC5

VOP_NC6

T-bar

0.2

0.4

0.6

0.8

1

0 2 4 6 8 10 12

Fh

-n/F

h-0

.25

Cycle number

VOP_NC7

VOP_NC8

T-bar

3.39

(a)

(b)

(c)

Figure 3.19. Load degradation response for test in HOC sample at: (a) v = 1 mm/s; (b) v = 3 mm/s;

(c) v = 30 mm/s

0

0.2

0.4

0.6

0.8

1

0 2 4 6 8 10 12

Fh

-n/F

h-0

.25

Cycle number

T-bar

VOP_HOC1

VOP_HOC2

VOP_HOC3

0

0.2

0.4

0.6

0.8

1

0 2 4 6 8 10 12

Fh

-n/F

h-0

.25

Cycle number

T-bar

VOP_HOC4

VOP_HOC5

VOP_HOC6

0

0.2

0.4

0.6

0.8

1

0 2 4 6 8 10 12

Fh

-n/F

h-0

.25

Cycle number

T-bar

VOP_HOC7

VOP_HOC8

VOP_HOC9

3.40

(a)

(b)

Figure 3.20. Comparison of soil sensitivity derived from the VOP and T-bar cyclic tests in: (a) NC

sample; (b) HOC sample

0.6

0.7

0.8

0.9

1

1.1

VO

P s

en

siti

vit

y/T

-ba

r se

nsi

tivit

y

Test name

Affected by reconsolidation

0

1

2

3

4

5

6

VO

P s

en

siti

vit

y/T

-ba

r se

nsi

tivit

y

Test name

3.41

(a)

(b)

(c)

Figure 3.21. Effect of rate on VOP resistance for: (a) peak load; (b) steady load; (c) remoulded load

0

0.5

1

1.5

2

2.5

1 10 100

Fh

p/F

hp

-ref

v (mm/s)

HOC sample

NC sample

NC best fit, m = 0.1405

HOC best fit, m = 0.1398

Combined best fit, m = 0.140.4

0.6

0.8

1

1.2

1.4

1.6

1 10 100

Fh

s/F

hs-

ref

v (mm/s)

NC sample

HOC sample

HOC best fit, m = 0.0772

NC best fit, m =0.0804

Combined best fit, m = 0.08

0.4

0.6

0.8

1

1.2

1.4

1.6

1 10 100

Fh

r/F

hr-

ref

v (mm/s)

NC sample

Best fit, m = 0.08

3.42

Figure 3.22. Estimation of peak loads for VOP tests in NC sample (all) considering strain softening

and rate effects

Figure 3.23. Relationship of the full-flow depth with the VOP normalised strength ratio

0

0.2

0.4

0.6

0.8

1

1.2

1.4

Est

ima

ted

pea

k lo

ad

/Mea

sured

peak

lo

ad

Test name

No adjustment

Adjusted for rate effects only

Adjusted for strain softening and rate effects

0.1

1

10

100

0 2 4 6 8 10 12

Fu

ll f

low

dep

th

su/γ'DVOP or su/γ'Dp @ T-bar @ spudcan

Matlock (1970), J = 0.25

Matlock (1970), J = 0.5

Hossain et al. (2005)

Hossain et al. (2006)

White et al. (2010)

NC sample

HOC sample

4.1

CHAPTER 4. EFFECTS OF CYCLIC DISTURBANCE

AND RECONSOLIDATION ON THE SHEAR STRENGTH

OF CLAY

FOREWORD

The investigation of the undrained strength degradation behaviour of clay due to cyclic

disturbance (Objective 2 – see Section 1.2.2) is considered further in this chapter by

including the effects of reconsolidation. These two competing effects are quantified

using the VOP (a device introduced earlier in Chapter 3). A framework to capture the

measured remoulding and reconsolidation behaviour of clay is outlined in this chapter.

This framework includes the use of the bearing factors during VOP lateral sweeps

discussed in the previous chapter. In summary, this chapter identifies and quantifies the

surprisingly significant strength recovery that follows remoulding events. This effect

can significantly raise the resistance forces between pipelines and the seabed.

4.2

ABSTRACT 4.1

The process of remoulding and subsequent recovery of soil shear strength is a common

geotechnical problem, especially in the offshore environment, where cyclic

environmental loading is by nature, intermittent. It is however often ignored in the

design of a geotechnical structure. A novel site investigation tool, namely the vertically

oriented penetrometer (VOP) – which can be used to obtain the soil shear strength via

analytical concepts familiar to laterally loaded piles – was used in a centrifuge testing

programme to provide additional insights into the changing soil strength arising from

successive disturbance and reconsolidation events. This test series involved cyclically

dragging the 4.72 mm diameter VOP (whilst embedded at depths of 30 or 45 mm below

the mudline) horizontally at velocities ranging from 0.3 to 30 mm/s in normally

consolidated kaolin, at a centrifuge acceleration corresponding to a scale factor of 100.

For tests at or slower than 1 mm/s, the VOP resistance initially reduces, but a trend of

increasing resistance after many cycles is evident, especially for tests at 0.3 mm/s.

Ultimately, the recovery in strength from reconsolidation can exceed the weakening

from remoulding. A simple framework using the critical state soil mechanics concept

that links the current soil strength to the state in specific volume – vertical effective

stress space is shown to provide reasonable predictions of the changing measured VOP

resistance, quantifying both the remoulding and reconsolidation processes. The degree

of strength recovery is predicted well using a simple vertical consolidation solution for

hydraulic fill, to describe the dissipation of excess pore pressure created by the VOP

shearing. Both the test data and the model framework reveal that recovery of the soil

strength is possible even within a cyclic shearing episode, provided that the time for

reconsolidation is sufficient. This behaviour is directly analogous to the soil response

during horizontal cyclic movements of offshore pipelines and piled foundations. In

these cases, shutdown-startup events and storms cause episodes of remoulding,

interspersed with long periods of reconsolidation. This study illustrates the soil strength

response to these events, and a basis by which it can be quantified.

4.3

INTRODUCTION 4.2

Many geotechnical problems involve episodes of rapid soil disturbance. In contractile

fine-grained soils, this causes a decrease in effective stress due to the accumulation of

positive excess pore pressure at constant volume (undrained) conditions, and an

associated reduction of undrained shear strength. When repeated over a large number of

cycles, the undrained shear strength will decrease to a minimal value, with the soil

reaching a fully remoulded state. During a calm period between two successive

disturbance events, the excess pore pressure will dissipate, resulting in densification of

the soil skeleton (i.e. a reduction of voids ratio or moisture content) and recovery of the

effective stress. The study of the strength recovery of a fine grained soil following a

disturbance event has been the subject of interest since the late 1970s for earthquake

engineering applications (France and Sangrey 1977; Ohara and Matsuda 1988;

Yasuhara and Andersen 1991; Hyde et al. 2007). These studies were based on

laboratory soil element tests using either the direct simple shear or triaxial apparatus.

However, these tests which studied reconsolidations of either silts or clays each

involved prior cyclic shearing which terminated before the effective stress reached the

critical state – i.e. tests halted at the pre-failure state.

In actual fact, many other repetitive loading events involve shearing the soil element

well past the critical state. This is especially prevalent in the offshore environment.

Simulation of reconsolidation following complete failure of the soil element is

impossible in the conventional simple shear or triaxial apparatus. However, it is possible

to simulate this important aspect of soil behaviour in the geotechnical centrifuge.

Examples of offshore events involving severe cyclic disturbance followed by

reconsolidation are shown in Figure 4.1.

Some of these scenarios have been investigated in various geotechnical centrifuge

model tests. Zhang et al. (2011) (Figure 4.1(a)) investigated the effects of repeated

lateral displacements and reconsolidation on pile-soil lateral (p-y) stiffness. In their

study, an initial episode of cyclic shearing led to a significant drop in lateral stiffness.

However, after 5 episodes of cyclic shearing and reconsolidation, the stiffness recovered

to 85% of the initial magnitude. Spudcan reinstallation into a previous footprint (Figure

4.4

4.1(b)) also involves reconsolidation following failure of the foundation soil. This effect

was investigated by Gan et al. (2012) who found that in normally consolidated clay, the

load required to reinstall the spudcan on a previous installation site increased with

elapsed reconsolidation time.

Figure 4.1(c) and Figure 4.1(d) depict examples of cyclic shearing events involving

offshore pipelines, where the former is a repetitive vertical movement of a steel catenary

riser at the riser-soil touchdown zone, and the latter illustrates an element of pipe within

an engineered lateral buckle arising from temperature-induced expansions and

contractions. Centrifuge tests simulating riser-soil interaction (Figure 4.1(c)) reported

by Hodder et al. (2009) showed that the ‘remoulded’ soil stiffness increased

continuously through 3 cyclic shearing and reconsolidation episodes. Rismanchian et al.

(2011) conducted centrifuge tests simulating the event depicted in Figure 4.1(d). Back-

calculation suggested that the strength of the berms which formed at the extremities of

the lateral buckling extent increased within a cyclic episode.

Figure 4.1(e) shows a submarine slide event, which perhaps causes the most severe

form of strength degradation compared to that depicted in Figure 4.1(a) to Figure 4.1(d).

However, after deposition of the slide material further downslope, the moisture content

may decrease below that of the intact parent slope. This is supported by controlled

laboratory test findings documented by Boylan et al. (2012).

It is evident that a framework that incorporates both effects of soil disturbance and

subsequent reconsolidation is highly desirable. White and Hodder (2010) formulated a

pioneering framework to capture the effects of complete soil remoulding and

reconsolidation by incorporating both strain softening and critical state soil mechanics

(CSSM) concepts. Their framework was compared to a centrifuge T-bar penetrometer

test involving 3 episodes of remoulding interspersed by 2 events of reconsolidation, and

encouraging agreement was achieved. Using the same test data for comparison, Hodder

et al. (2012) incorporated a more complex form of the strain softening law whilst

maintaining the same CSSM framework, coupled with a lateral dissipation solution of

Bolton (1979) to estimate the gain in soil strength during a calm period. However, these

studies assumed that no reconsolidation occurs within a cyclic shearing event, which

4.5

may happen, as suggested by Rismanchian et al. (2011) in the back analysis of their test

results.

This paper presents a methodology that captures the effects of reconsolidation occurring

within a cyclic shearing event using the simple framework of White and Hodder (2010).

A centrifuge test programme using a novel penetrometer involving cyclic lateral

displacements at different velocities was conducted. This penetrometer is named the

vertically oriented penetrometer (VOP), and is developed for improving the

geotechnical characterisation of centrifuge model samples, and potentially also future

site investigation practice along pipeline routes (Sahdi 2012). The results of the

centrifuge tests will be presented first, followed by the underlying concepts of the

framework used to back analyse the test data. Finally, the framework will be compared

to the centrifuge test results. In combination, the VOP and the model framework

provide a basis to quantify the changing strength of near-seabed soils during episodes of

disturbance and reconsolidation.

TEST METHODOLOGY 4.3

Experimental equipment 4.3.1

All tests were performed in the UWA beam centrifuge at an enhanced acceleration of

100 g. The Accutronic Model 661 beam centrifuge has a maximum spinning radius of

1.8 m and a maximum capacity of 40 g-tonnes, which is equivalent to a maximum load

capacity of 200 kg at 200 g. A more detailed description of the UWA beam centrifuge is

provided by Randolph et al. (1991).

The vertically oriented penetrometer (VOP) as shown in Figure 4.2 is cylindrical in

shape. This new penetrometer (Sahdi 2012) was developed as a new technology to

support the assessment of near-surface soil strength in small scale testing – for example

in the geotechnical centrifuge. However, the horizontal profiling capability of this tool

means that, it has the potential to be applied in situations that require the

characterisation of a shallow zone with a long lateral extent, such as a pipeline route. In

essence, it provides a continuous measurement of the shallow shear strength along the

4.6

route that it is dragged. This can be a more efficient investigation approach than discrete

vertical penetrations spaced at lateral intervals.

The shear strength of the soil can be estimated from the measured lateral resistance

using the same concepts applied to a laterally loaded pile (e.g. Murff and Hamilton

1993). This prototype VOP has a total length of 130 mm and is made from Aluminium

6061 T6. A threaded section having a length of 25 mm and a larger diameter of 7 mm

provides a fixed connection to the actuator, thus rendering the VOP as a cantilevered

beam when loaded laterally by the soil sample, whilst displaced horizontally. The VOP

is instrumented with 4 levels of strain gauges with each consisting of 2 strain gauges on

opposite sides to detect the net lateral pressure on both the front and back sides of the

device. These strain gauges (labelled as A, B, C and D in Figure 4.2) are located at

55 mm, 70 mm, 85 mm and 100 mm from the VOP tip and are protected with a layer of

epoxy resin. The unthreaded section is hollow with an outer diameter (OD), DVOP of

4.72 mm and an inner diameter (ID) of 2.9 mm.

Clay properties and sample preparation 4.3.2

The centrifuge tests were carried out using commercial kaolin clay. The mechanical

properties of the kaolin used in The University of Western Australia (UWA) are well

studied and provided in Table 4.1. To prepare a model seabed for centrifuge testing, dry

kaolin clay was initially mixed with water at twice the liquid limit (120%) under

vacuum in a barrel mixer. The well-mixed slurry was then transferred to a strongbox

with inner measurements of 650 mm long, 390 mm wide and 325 mm in height. Prior to

this, a 30 mm sand layer was placed at the bottom of the strongbox to facilitate drainage

during sample consolidation. The centrifuge was then spun for 5 days at 100 g until

primary consolidation was completed. This resulted in a model seabed with a normally

consolidated strength profile having a total height (model scale) of 205 mm.

Test procedure 4.3.3

In the centrifuge tests, the VOP threaded section was attached rigidly to an actuator

capable of a maximum horizontal velocity of 100 mm/s. Table 4.2 provides a summary

4.7

of the testing details, depicted according to the tests order. All tests were performed at a

centrifuge acceleration level corresponding to 100 g.

T-bar tests for shear strength profiling were conducted before commencement and after

completion of the VOP tests. These tests were conducted at a penetration rate of 1 mm/s

using a T-bar with a diameter of 5 mm and a length of 20 mm. Each T-bar test consisted

of a cyclic penetration and extraction regime to deduce the sensitivity, St, of the soil.

The T-bar test details are shown in Table 4.2.

All VOP and T-bar tests were conducted under displacement control with repetitive

cycles of horizontal or vertical displacements respectively. Before commencing

horizontal translation, the VOP was installed vertically to the desired embedment at a

rate of 0.5 mm/s. The VOP test velocities (v), horizontal travel distances (u), number of

cycles (n), and total vertical embedment (zT) are shown in Table 4.2.

To avoid potential boundary effects, all T-bar and VOP tests were conducted at a

minimum distance of 10DVOP from the edges of the strongbox. An ample spacing of at

least 4DVOP was also allowed between test sites to avoid overlapping of the zones of

influence from each failure mechanism.

UNDRAINED SHEAR STRENGTH AND SENSITIVITY 4.4

OF SAMPLE

Figure 4.3 depicts the strength derived from the initial T-bar penetrations, su-in during

tests T-bar_NC1 and T-bar_NC2 (see Table 4.2), which were performed before and

after the VOP tests. By making use of Equation 4.1, the profile of soil strength (su) can

be derived from the T-bar penetration resistance, qT-bar:

T baru

T bar

qs

N

4.1

In order to infer the soil strength, the common practice is to adopt a constant bearing

factor, NT-bar, of 10.5 (which reflects an intermediate surface roughness condition)

4.8

corresponding to a deep flow-round mechanism (Randolph and Houlsby 1984; Stewart

and Randolph 1991; Martin and Randolph 2006). In reality, qT-bar will be affected by

both soil buoyancy and a near-surface heave failure mechanism (rather than flow-

round), leading to significantly lower NT-bar before the deep failure mechanism is

mobilised. Therefore, the two su-in profiles shown in Figure 4.3 were corrected for both

soil buoyancy and surface heave effects according to the procedure outlined by White et

al. (2010). Since the soil sample is normally consolidated, the normalised strength ratio,

su/γ′DT-bar (where γ′ = soil effective unit weight, DT-bar = T-bar diameter) is low. This

causes the full-flow failure mechanism around the T-bar to be achieved at an average

depth of just 1.5DT-bar. At depths below the mudline, z, the su-in profiles can be

represented by the following equation:

u in sus k z (kPa) 4.2

where the best fit strength gradient, ksu, is 0.09 kPa/mm at model scale, which

corresponds to 0.9 kPa/m at prototype scale.

The full cyclic T-bar penetration and extraction profiles are plotted in Figure 4.4(a). It

should be noted that near the mudline, the NT-bar shallow correction procedure of White

et al. (2010) can only be used for the penetration phase as an extraction process involves

a different failure mechanism. To avoid unnecessary complexity, a constant NT-bar factor

of 10.5 is adopted for the cyclic interpretation, because the cyclic test depths are located

in the deep-failure mechanism zones, where NT-bar = 10.5 applies. The degradation in

strength resulting from continuous cycles of penetration and extraction can be

quantified using the degradation factor, defined as the shear strength at a particular

cycle, su-n, normalised by the initial T-bar penetration strength, su-in. These strength

values are extracted from the mid-depth of the T-bar cyclic test distance. The resulting

decrease in degradation factor with cycle number derived from all the T-bar tests is

shown in Figure 4.4(b). The cycle number, n, is calculated as recommended by

Randolph et al. (2007). The first penetration is designated as n = 0.25 and the n for

subsequent extractions and penetrations is cumulatively calculated in increments of 0.5

(n = 0.75, 1.25, etc.). From the average degradation curves in Figure 4.4(b), the

4.9

sensitivity, St, can be estimated from the inverse of the average degradation factors once

the degradation curve has stabilised (at about n > 8.25 – 10.75). The resulting post-

cyclic St is approximately 2.3.

Generally, as depicted in Figure 4.3 and Figure 4.4, no significant differences in

strength profiles and sensitivities between T-bar tests are observed. This indicates good

sample consistency throughout the testing period.

RECONSOLIDATION DURING CYCLIC VOP TEST 4.5

The key results from this centrifuge test series are the load-displacement responses of

the VOP, which vary with test embedment, velocity and cyclic distance. These three

parameters will ultimately affect the rate of dissipation of positive excess pore pressure

relative to the number of cycles performed, as discussed in Section 4.6 to Section 4.8.

Two examples of the full VOP load-displacement profiles for zT = 30 mm (where zT is

the VOP test embedment) are shown in Figure 4.5. Tests VOP_NC3 (v = 3 mm/s) and

VOP_NC0330 (v = 0.3 mm/s) involved a total of 9.75 and 26.75 cycles, respectively. In

these figures, the horizontal displacement, u, is normalised with the VOP diameter,

DVOP. It should be noted that the cycle numbering system follows that for the T-bar

cyclic test (Section 4.4). The first VOP horizontal forward sweeps (n = 0.25) in both

tests show distinct peaks after which the resistance stabilises at a displacement of 2 –

3DVOP. The peak loads for all the tests are generally about 1.5 – 2 times the steady

loads.

By using these steady loads, Sahdi (2012) showed that the deep failure mechanism is

mobilised at a very shallow depth of 1DVOP and beyond because of the low strength

ratio, su/γ′DVOP (where γ′ is the unit weight and DVOP is the VOP diameter), in this

sample. Owing to the dominating effect of the flow-round mechanism (as opposed to

any transition from a two-sided to one-sided shallow wedge failure), the transition from

peak to steady load is governed by softening of the soil mobilised around the VOP. This

response is expected because this clay is not an elastic-perfectly plastic material in

undrained conditions, but softens with accumulated strain. This feature of behaviour is

4.10

well-established from analysis of the T-bar (Einav and Randolph 2005; Zhou and

Randolph 2007; Zhou and Randolph 2009b).

Interestingly, while cyclic test VOP_NC3 shows a typical decrease in the load-

displacement profile akin to the T-bar cyclic test (Figure 4.4(a)), test VOP_0330 shows

a distinctly different cyclic load-displacement profile (Figure 4.5(b)). After the first

VOP forward sweep, the load continuously decreases from cycle number 0.75 to 4.25,

after which it increases steadily, surpassing even the initial steady load after many

cycles.

The evolution of the load with increasing cycles of forward and backward sweeps for

each test is depicted more clearly in Figure 4.6. Here, the load at the mid-cycle position

in each cycle, Fh-n, is normalised by the load from the first mid-cycle VOP forward

sweep (n = 0.25), Fh-0.25. For ease of identification, the dotted lines represent the tests at

an embedment of 30 mm and the results for tests at an embedment of 45 mm are plotted

as solid lines. A dotted and solid line of similar colour indicates tests at the same

velocity. For comparison, the average T-bar degradation curve (Figure 4.4 (b)) is also

included in the figure. Similar soil remoulding response to that from the T-bar is

observed for tests at v ≥ 3 mm/s regardless of embedment depths. These results are

expected to overlie each other since the flow-round mechanism dominates the resistance

in both the T-bar and the VOP, so the resulting resistance patterns are a function only of

the soil remoulding behaviour (Sahdi 2012).

However, different profiles of load-cycle number are evident for the tests at v ≤ 1 mm/s.

The resulting ‘remoulded’ Fh-n/Fh-0.25 inferred from the two cyclic tests at v = 1 mm/s

(tests VOP_NC1 and VOP_NC2) appears to be higher with decreasing embedment, and

a slight increase in Fh-n/Fh-0.25 when n > 7.75 can be seen in these two data. This

increasing trend of Fh-n/Fh-0.25 with cycle number is much more pronounced when the

test velocity is decreased to 0.3 mm/s (tests VOP_NC0330 and VOP_NC0345), where

the rate of increase in Fh-n/Fh-0.25 with cycle number (beyond 4.25) amplified with

decreasing embedment.

The centrifuge tests reported in this paper involved repeated cyclic failure as the VOP is

pushed to a given soil location. The varying test parameters as outlined in Section 4.3.3

4.11

and Table 4.2 were the VOP cyclic test distance, velocity and test embedment. These

test parameters point directly to the reconsolidation response of the soil at a given

location after disturbance, since the cyclic test distance and velocity affect the

reconsolidation time between each occasion that the VOP passes the mid-cycle position.

Also, the embedment may be directly linked to the drainage path length, which controls

the rate of pore pressure dissipation.

Having presented some of the experimental results, the remaining portion of this paper

details the attempts to replicate analytically the cyclic response shown in Figure 4.6

using the simplified remoulding and reconsolidation framework of White and Hodder

(2010).

FRAMEWORK FOR BACK ANALYSIS 4.6

Strength analysis considering drainage condition 4.6.1

It is important to understand the soil drainage condition during a VOP pass, so that the

correct strength can be interpreted from the resulting resistance. The VOP total load

may increase due to both viscous and partial drainage effects, where the latter effect is

potentially greater. The degree of drainage, or consolidation, is controlled by the non-

dimensional velocity, V, which is defined as:

VOP

v

vDV

c 4.3

where v is the VOP horizontal velocity and cv is the coefficient of consolidation of the

soil. Figure 4.7 presents the variation of the measured average steady load across the

range of tested V. In the figure, the measured average steady load from the first forward

sweep (n = 0.25), Fhs, is normalised with that measured at v = 1 mm/s, Fhs-ref. In order to

calculate V, a value of cv = 2.6 m2/year is adopted (Table 4.1). Also plotted in the figure

are two backbone curves (neglecting viscous effects) derived using two separate T-bar

4.12

variable rate studies of Watson and Suemasa (2000, unpublished) and House et al.

(2001). These curves can be represented as:

hs

d

hs ref

F ba

F 1 cV

4.4

where values of a, b, c and d can be found in Table 4.3. The magnitude of a + b is the

ratio of the drained to undrained resistance. Although there is only limited data, it

appears from Figure 4.7 that the small increase in the Fhs/Fhs-ref ratio at v = 0.3 mm/s (V

= 17) falls within the Watson and Suemasa trendline, which may indicate partially

drained shearing condition as the VOP is pushed horizontally. However, due to the

absence of data at V < 17, and considering the fact that the partial drainage effect (if

any) is minor (based on the Watson and Suemasa’s trendline), undrained conditions are

assumed to govern within a single cycle at v = 0.3 mm/s for the subsequent back

analysis of the VOP data. Beyond V = 57 (v > 1 mm/s), it can be seen that the

normalised VOP resistance increases owing to viscous effects (e.g. Biscontin and

Pestana 2001; Boukpeti et al. 2012).

If the VOP shearing process is assumed to be completely undrained, Fh-n (VOP load

extracted from the mid-cycle position at a certain cycle number) can be calculated using

the convenient total stress approach commonly used for a laterally loaded pile (e.g.

Matlock 1970; Randolph and Houlsby 1984; Murff and Hamilton 1993):

Tz

h n h d VOP u op

0

F N D s dz 4.5

where Nh-d is the bearing capacity factor, and zT is the total VOP embedment depth. At

velocities of 0.3 and 1 mm/s, the operative shear strength, su-op at a particular soil depth,

z, is assumed to equal the su-in from the T-bar tests (Equation 4.2). Because the flow-

round mechanism dominates over the VOP length (Section 4.5), a constant Nh-d = 10.5

is used. Since the base area of the VOP is small relative to the shaft, any soil resistance

arising from the tip shearing is neglected.

4.13

Strength evolution due to repeated disturbance and reconsolidation 4.6.2

Besides quantifying the change in su-op due to remoulding, consideration of the effects of

reconsolidation on the su-op is equally important. These effects can be accounted for

using a form of the framework outlined by White and Hodder (2010). In essence, this

framework links the soil strength degradation to the accumulation of positive excess

pore pressure, and the recovery of soil strength is linked to the dissipation of that excess

pore pressure.

To idealise the stress path of a representative soil element near the VOP, consider the

effects as the VOP is swept forward and backward as shown schematically in Figure

4.8(a). It is assumed that the soil element is initially on the ‘wet’ side of the critical state

line. As the soil element is sheared during the first VOP forward push, positive excess

pore pressure will be generated, causing the in-situ vertical effective stress, σ′vo, to

decrease. The VOP then moves further away from S to the cyclic distance limit and

subsequently returns back to shear element S further. If the elapsed time between

multiple shearing events is small, i.e. the cyclic velocity of the VOP is fast enough,

negligible reconsolidation occurs between the first and last shearing events, the whole

process can be treated as undrained, and the soil will be fully remoulded after a certain

number of cycles. However, if the elapsed time during the process is long, i.e. the

velocity is small and the cyclic distance large, the soil may reconsolidate during the

cyclic process, including between individual shearing events, leading to an increase in

the VOP resistance due to strengthening of the soil.

Figure 4.8(b) represents an idealisation of the effective stress path of soil element S in

the specific volume (v) – vertical effective stress (σ'v) space during a VOP cyclic

episode based on the same mechanism postulated by White and Hodder (2010). For

convenience, the effect of stress anisotropy is neglected and the soil strength is assumed

to be governed solely by the change in the vertical effective stress.

In the undisturbed, normally consolidated state, S will lie on the normal consolidation

line (NCL) in v - σ'v space (point A in Figure 4.8(b)). The in-situ specific volume can be

calculated from Equation 4.6:

4.14

NCL voN ln ' 4.6

where NNCL is the specific volume on the NCL at σ'v = 1 kPa and λ is the slope of the

NCL.

The model is based on two failure lines in specific volume - effective stress space. The

critical state line (parallel to the NCL with slope λ) is assumed to be reached when the

soil element first fails (in the case of a T-bar or VOP test, this corresponds to the n =

0.25 shearing event). During subsequent undrained cycles, the stress point migrates to a

second failure line, defined as the remoulded state line (RSL).

During a particular VOP cycle number, the positive excess pore pressure results in a

current vertical effective stress, σ′vn (subscript n denotes cycle number), which is a

proportion of the stress at the critical state line corresponding to the current specific

volume, v:

vn vCSL nσ' Rσ' 4.7

where σ′vCSL-n is the vertical effective stress on the CSL at the current n, and so is linked

to the current specific volume by:

Γ

λvCSL nσ' e

v

4.8

where Γ is the CSL specific volume at σ′v = 1 kPa.

The parameter R controls the position of the failure stress, σ′vn between the CSL and the

RSL and can be written as (Einav and Randolph 2005):

953(n 0.25)/N

t t

1 1R 1 e

S S

4.9

4.15

where St is the soil sensitivity and N95 is the number of VOP cycles to cause 95%

degradation of the initial resistance when negligible reconsolidation occurs during a

cyclic episode. In this way, the progress of the failure stress point from the CSL to the

RSL is linked to the degradation in resistance seen in undrained cyclic T-bar or VOP

tests.

During the first VOP forward sweep, it is assumed that the state of S reaches the CSL

(A to B in Figure 4.8(b)), implying that R = 1 (since n = 0.25). Before the next VOP

shearing event, as the VOP moves away from S (Figure 4.8(a)), excess pore pressure

(with a magnitude of σ'vo - σ'vn) may dissipate and σ′vn increases by ∆σ′v (cyclic process

not fully undrained):

v vo vn' 'σ U(σ σ' ) 4.10

where U is the degree of excess pore pressure dissipation, which is derived from the

elapsed time. During reconsolidation, densification of S occurs, leading to a decrease in

specific volume, ∆v (point B to C):

v

vn

vnσ' σ'κln

σ'

v 4.11

where the magnitude of ∆v depends on the reconsolidation slope κ. However, if the time

between two passes of the VOP is short, negligible reconsolidation occurs and the

effective stress path proceeds from B to B' on the remoulded state line (RSL) –

complete soil strength degradation. It is assumed that the RSL represents the lowest

stress state of the soil element at any given v, and the stress spacing ratio between the

RSL and the CSL is equal to the soil sensitivity, St.

As the VOP returns on the next cycle ((Figure 4.8(a)), the effective stress at failure

moves to a position between the CSL and the RSL which depends on the cumulative

number of cycles (Equation 4.9), to position E as depicted in Figure 4.8(b). To calculate

the stress at E (via Equation 4.7), the failure stress on the CSL – point D in Figure

4.8(b) (Equation 4.8) at a decreased specific volume (due to reconsolidation i.e. – point

4.16

B to C in Figure 4.8(b)) is first determined. The failure stress at D is then factored down

by R (Equation 4.9) according to the current cycle number – giving the stress state at E

(Figure 4.8(b)). This process of disturbance and reconsolidation continues until the end

of the VOP cyclic episode. At any given σ′vn during shearing, the operative strength,

su-op can be calculated from the friction factor, μ as:

u op vns μσ' 4.12

Ultimately, the VOP resistance can be calculated using Equation 4.5.

Dissipation of excess pore water pressure 4.6.3

As can be seen in Figure 4.6, the rate of increase in Fh-n/Fh-0.25 for tests at v ≤ 1mm/s

(after n > 4.25) is higher with decreasing test embedment at a given test velocity and

cycle distance. This suggests that the dissipation of positive excess pore pressure

primarily occurs vertically, and is controlled by the total VOP embedment depth, zT.

During the first VOP horizontal sweep, the distribution of excess pore pressure can be

assumed to be triangular in shape, increasing linearly from the mudline, and equal to

σ′vo - σ′v0.25. If dissipation of the excess pore pressure is assumed to occur only at the

mudline-free water boundary, the remaining excess pore pressure (ue-tc) at a given depth

below the mudline (z) up to the total VOP embedment depth (zT) can be estimated using

the Fourier-series solution of Terzaghi and Frohlich (1936):

2 2

ezTe tc 2 2

x 1 T

8u xπ xπz x π Tu sin sin exp

π 2 2z 4x

4.13

where x is the number of Fourier-series terms and uezT is the excess pore pressure at z =

zT. The elapsed reconsolidation time, tc, is expressed as a dimensionless time factor, T:

v c

2

T

c tT

z 4.14

4.17

The isochrones of ue-tc normalised by uezT after the first VOP forward sweep at depths

up to zT are plotted in Figure 4.9. The resulting degree of consolidation, U at a given

depth is then:

e tc

ez-t0

uU 1

u

4.15

where uez-t0 is the excess pore pressure generated at a particular depth below the

mudline. For convenience, U calculated from the first VOP shearing and

reconsolidation (n = 0.25) event is assumed to apply throughout the cyclic episode.

As a comparison with the vertical dissipation model outlined above, the lateral

dissipation model proposed by Bolton (1979) can also be used to estimate U (Equation

4.15). Hodder et al. (2012) used this same consolidation solution to estimate the degree

of excess pore pressure dissipation after a cyclic T-bar remoulding episode, achieving

good agreement with their experimental data. In this solution, a rectangular excess pore

pressure distribution is assumed (Figure 4.10(a) – presented as a slice of the VOP-soil

plan section at any depth z) where the maximum magnitude is equal to σ′vo - σ′vn. The

excess pore pressure section is assumed to extend to 2DVOP laterally from the centre of

the VOP. This is analogous to the extent of the ‘influence zone’ of a T-bar flow-round

failure mechanism (Zhou and Randolph 2007; Zhou and Randolph 2009b; Hodder et al.

2010; Hodder et al. 2012). Dissipation occurs laterally on two sides of the rectangular

excess pore pressure columns (two-way drainage). The resulting U with T is plotted in

Figure 4.10 (b), where T is defined as:

v c

2

VOP

c tT

D 4.16

Also plotted in Figure 4.10 (b) is the dissipation of excess pore pressure predicted using

the Terzaghi and Frohlich (1936) solution with increasing time factor T (Equation 4.14)

at normalised depths (z/zT) of 0.25, 0.75 and 1. It can be seen that for z/zT ≤ 0.75, the

Terzaghi and Frohlich (1936) solution predicts a slower consolidation rate compared to

4.18

that of Bolton (1979) for T ≤ 0.1. However, full consolidation occurs much faster

(T ~ 2) following the predictions of Terzaghi and Frohlich (1936).

It should be noted that the 2 methods to estimate the degree of consolidation outlined in

this section can only be used to predict the change in specific volume, v (Equations

4.10-4.11) at a particular depth z below the mudline during reconsolidation. The

operative strength, su-op during a subsequent VOP pass can then be determined based on

the current v (Equations 4.7, 4.8 and 4.12) which can be ultimately used to estimate the

integrated VOP resistance (Fh-n) via Equation 4.5.

REMOULDING-RECONSOLIDATION PARAMETERS 4.7

The parameters used to apply this model to the VOP tests include NNCL, κ, λ, and cv,

which have been introduced previously in Table 4.1. The remaining model parameters

required are presented in Table 4.4. This section summarises the methodology used to

derive these remaining parameters.

Failure state parameters 4.7.1

During the first VOP forward sweep, the soil fails at the CSL, and su-op is equal to

μσ'vCSL-0.25 (Equation 4.12). By combining equations 4.6, 4.8 and 4.12, the normally

consolidated strength ratio, (su/σ′v)nc can be written as follows:

NCL

NCL

ΓΓ Nλ

u vCSL 0.25 λN

v v0 λnc

s μσ' μe μe

σ' σ'e

v

v

4.17

where Γ is the specific volume at a vertical effective stress σ'v of 1 kPa on the CSL. The

(su/σ′v)nc shown in Table 4.1 relates to the intact soil strength which is not identical to

the strength derived from the initial T-bar penetration, su-in. This is usually somewhat

lower than the intact value – corresponding to that inferred from element testing –

owing to the significantly higher level of strain softening in the soil around the T-bar

probe (Einav and Randolph 2005; Zhou and Randolph 2007; Zhou and Randolph

4.19

2009b). The discrepancy arises since the intact strength measured in the laboratory

corresponds to n = 0, whereas the initial T-bar penetration is linked to a level of

remoulding corresponding to n = 0.25 (see Equation 4.9). Assuming that the T-bar su-in

is about 40% lower than the intact shear strength (Teh 2007; Hodder et al. 2010; Sahdi

et al. 2010), the operative (su/σ′v)nc is taken as 0.15. The friction factor, μ, used to

calculate the su-op at any given σ′vn (Equation 4.12) on the CSL and below, is assumed to

equal to that reported in White and Hodder (2010) for the T-bar penetrometer (μ = 0.7).

The CSL parameter, Γ can now be estimated as:

1uNCL

v nc

sΓ N λln μ

σ'

4.18

The resulting Γ is 3.29.

Soil sensitivity and degradation rate parameter 4.7.2

Parameters St and N95 control the magnitude of soil damage at a given VOP pass

(Equation 4.9). The Fh-n/Fh-0.25 degradation test data at v > 1 mm/s (no reconsolidation)

were used to derive these parameters by fitting the calculated Fh-n/Fh-0.25 (Equation 4.9)

to that measured. This is achieved by optimising St and N95 using the least square

method. The best fit St and N95 are 2.3 and 2.5 respectively.

COMPARISON OF MODEL WITH TEST DATA 4.8

Comparisons between the measured horizontal loads at a given cycle number, Fh-n with

those predicted using the model framework outlined in previous sections for VOP tests

pushed at v = 1 and 0.3 mm/s are shown in Figure 4.11 (30 mm embedment) and Figure

4.12 (45 mm embedment). It should be noted that the measured Fh-n were extracted from

the mid-cycle distance of each test. In these figures, the performance of the model using

the two different excess pore pressure dissipation models is demonstrated.

4.20

It is clear that the framework which incorporates the lateral dissipation model of

Bolton (1979) to estimate the effective stress recovery, significantly overpredicts Fh-n

for all the data. The time factors, T for tests with velocities of 1 mm/s (VOP_NC1 and

VOP_NC2) and 0.3 mm/s (VOP_NC0330 and VOP_0345) are 0.148 and 0.247

corresponding to a degree of consolidation U of 22% and 28% respectively. The U

predicted using the Bolton (1979) approach applies throughout the VOP length and is

believed to cause the calculated su-op to increase at a significantly higher rate resulting in

overestimations of Fh-n. Moreover, Bolton’s lateral dissipation solution assumes that a

free draining layer is present in front of the excess pore pressure rectangular column. In

reality, the model soil layer does not contain this infinitely permeable layer and

therefore may not be applicable to the cyclic VOP tests reported in this paper. However,

the Bolton model may work in prototype conditions where coarse and fine grained soils

are horizontally stratified. In this case, the course layer provides a one-way drainage

path in front of the VOP.

An improved estimate is evident for each test when the model utilises the one-way

vertical dissipation model of Terzaghi and Frohlich (1936). This indicates that the

dissipation of excess pore pressure is predominantly vertical.

The model agrees well with the measured data for tests at v = 1 mm/s (VOP_NC1 and

VOP_NC2). The dimensionless elapsed time factors, T for reconsolidations to occur are

T = 0.0037 and 0.0016 in tests VOP_NC1 and VOP_NC2, during which, dissipations of

only U = 6.8% and 4.6% at each deepest test embedment (zT) occur, reflecting the

increasing zT of 30 mm and 45 mm respectively. For comparison purposes, the VOP

resistance - cycle number model response when no allowance for reconsolidation

between two VOP passes is plotted in Figure 4.11(a) and Figure 4.12(a) each. As

expected, the predicted remoulded Fh-n is lower for each case.

The performance of the model when compared to the test data at v = 0.3 mm/s (Figure

4.11(b) and Figure 4.12(b)) is strongly influenced by the reconsolidation element of the

model. The general trend of initial softening, followed by gentle hardening, is captured

well by the model using Terzaghi and Frohlich’s vertical dissipation. However, using

the base case input parameters, the model undepredicts the measured Fh-n for tests

4.21

VOP_NC0330 and VOP_NC0345 by about 30% at each last cycle number. Again, for

comparison purposes, the model response when no allowance for reconsolidation is

plotted in the corresponding figures.

There are two likely explanations for this discrepancy. As mentioned in Section 4.6.3, U

for each reconsolidation event is estimated based on the triangular excess pore pressure

distribution generated during cycle number, n = 0.25. In fact, as repeated events of

disturbance and reconsolidation increase, this distribution is no longer triangular thus,

decreasing the accuracy of the model if consolidation occurs at a higher rate (as seen for

tests VOP_NC0330 and VOP_NC0345).

Alternatively, the assumed parameters that control dissipation and strengthening may be

inaccurate. For example, a higher reconsolidation slope κ may be more appropriate.

This is in accordance to the experimental findings using triaxial and simple shear tests

(Ohara and Matsuda 1988; Yasuhara and Andersen 1991; Hyde et al. 2007) where

reconsolidation slopes of 1.5κ up to magnitudes comparable to the slope of the NCL or

CSL – λ – were documented in these studies. Notwithstanding the differences in the

nature of the testing methods (where the VOP causes the soil to fail beyond the CSL

compared to that in cyclic element tests), the sensitivity of the model behaviour to this

parameter is illustrated in Figure 4.11and Figure 4.12 where additional response based

on vertical dissipation using a higher (by a factor 2) value of κ is illustrated. The

agreement is now excellent for the tests at v = 0.3 mm/s (VOP_NC0330 and

VOP_NC0345), but the model predicts a slightly faster rate of increase in Fh-n for the

tests at v = 1 mm/s (VOP_NC1 and VOP_NC2).

Nevertheless, the framework is shown to be able to capture the underlying trend of VOP

resistance evolution during a cyclic event when enough time for reconsolidation in

between shearing events is permitted.

The simulated stress path in the specific volume – effective stress (v - σ'v) space of a soil

element at the deepest test depth of 45 mm below the mudline for test VOP_NC0345 is

shown in Figure 4.13. As depicted, all stress paths are limited by the RSL, reflecting the

maximum sensitivity, St of 2.3. At each reconsolidation event, the effective stress

recovers where the cumulative recovery at the end of the simulated cyclic event (n =

4.22

35.75) resulted in a decrease in v of about 20% with respect to the original in-situ v. The

decrease in v will inevitably cause a reduction in the soil moisture content, which

ultimately results in surface settlement. Photographic evidence of soil surface settlement

is provided in Figure 4.14, corroborating this back-analysis.

STEADY STATE CYCLIC RESISTANCE 4.9

It is interesting to investigate the effects of further events of disturbance and

reconsolidation on the simulated stress path. For this, the stress paths of a soil element

at the deepest embedment, zT = 45 mm for test VOP_NC0345 is extrapolated up to n =

400.25 (Figure 4.13). The resulting evolution of the σ'vn of the same soil element at an

arbitrary VOP pass normalised by σ'v0.25 (first pass) for the complete cyclic episode (up

to n = 400.25) is shown in Figure 4.15. Using the model assumptions, if the degree of

reconsolidation is sufficient, the effective stress will eventually converge to the original

in-situ vertical effective stress, σ'vo magnitude.

No excess pore pressure is developed at this steady state, which lies on the RSL. At this

stage, a drained cyclic effective stress is achieved, σ'vcyc-drained and further failure at any

rate of shearing does not generate excess pore pressure (White and Cathie 2010).

Beyond this, the ratio of σ'vcyc-drained/σ'v0.25 reaches a plateau of 4.67 (Figure 4.15).

This is directly analogous to the drained to undrained resistance ratio during a

monotonic shearing (Figure 4.7), which is within the range documented for

penetrometer and spudcan penetration in fine-grained soil of 3.1 – 5.5 (Watson and

Suemasa 2000; House et al. 2001; Randolph and Hope 2004; Chung et al. 2006;

Barbosa-Cruz 2007; Oliveira et al. 2011). Since soil has no tendency to contract or

dilate at this stage and any failure mechanism around the VOP must occur at constant

volume, the total stress approach (Equation 4.5) is assumed to be equally applicable

here, thus resulting in a VOP normalised resistance at zT equal to 4.67.

4.23

CONCLUSIONS 4.10

The results of a series of centrifuge tests utilising the novel vertically oriented

penetrometer (VOP) to perform cyclic horizontal tests have been presented in this

paper. In this test series, the VOP was embedded at 30 and 45 mm depth (6.4 – 9.5 VOP

diameters), and the horizontal cyclic velocities were 0.3, 1, 3, 10 and 30 mm/s for each

test embedment. For test velocities of 3, 10 and 30 mm/s, similar load degradation

profiles to that inferred from a T-bar cyclic test are observed, owing to the intense soil

disturbance, where ultimately a steady remoulded state was reached. However, for tests

velocities slower than 1 mm/s, it was found that the VOP resistance rises after the first

few cycles, reflecting the dissipation of excess pore pressures generated during each

cycle, leading to reconsolidation. The rate of increase in the VOP resistance with cycle

number rises with decreasing embedment – reflecting the vertical direction of drainage

to the mudline – and decreasing velocity – reflecting the dissipation time between

cycles. The VOP resistance even surpasses the initial value after many cycles for test

velocity of 0.3 mm/s, illustrating that the strengthening effect of reconsolidation can

eclipse the weakening effect of remoulding.

The simple framework of White and Hodder (2010) which combines the effects

disturbance and reconsolidation is used to back-analyse the centrifuge data. By

incorporating a one-dimensional consolidation solution of a hydraulic fill into this

framework (to estimate the degree of consolidation in between two VOP passes), good

predictions of the increasing measured VOP resistance can be made.

The test data shown in this paper shows how the increase in soil strength during a cyclic

shearing episode can be quantified, allowing for the competing effects of remoulding

and reconsolidation. The data and the framework used in the back analysis are relevant

to other cyclic processes found in offshore engineering, including the cyclic response of

pipelines and piles, as well as the re-installation of spudcan foundations. The simple

framework outlined in this paper and the VOP as a novel soil characterisation tool, offer

a basis to estimate the significant changes in soil strength that can occur during cyclic

loading events that span a timeframe comparable to the consolidation process.

4.24


This research is supported by the Minerals and Energy Research Institute of Western

Australia, and by BP, BHP Billiton, Chevron, Petrobras, Shell and Woodside. The

financial support of all the participants is gratefully acknowledged. The authors are

grateful to James Schneider, Don Herley, Phil Hortin and Shane De Catania for their

technical assistance. The first author is also grateful for the financial support received

from the Ministry of Higher Education Malaysia (MOHE) and Universiti Malaysia

Sarawak (UNIMAS).

4.25

Table 4.1. Properties of UWA kaolin

Property Value Reference

Specific gravity ,Gs 2.6

Boukpeti et al. (2012) Plastic limit, PL 28 %

Liquid limit, LL 58.4 %

Coefficient of consolidation (at σ'v = 10 – 20 kPa), cv

(m2/year)

2.6 Stewart (1991)

Normal consolidation line (NCL) or Critical state line

(CSL) slope, λ 0.281

Richardson (2007, unpublished) Specific volume at σ'v = 1 kPa on NCL, NNCL 3.722

Unload-reload slope, κ 0.06

Normally consolidated strength ratio, (su/σ'v)nc – from

direct simple shear test 0.25 Lehane et al. (2009)

Table 4.2. VOP and T-bar test details

Test

reference

VOP

embedment,

zT (mm)

VOP

horizontal

displacement,

u (mm)

T-bar total

vertical

displacement,

(mm)

VOP

horizontal

cyclic

distance

(mm)

T-bar

vertical

cyclic

depth

(mm)

Test

velocity,

v (mm/s)

Number

of cycles,

n

T-bar_NC1 - - 100 - 18-48 1 10.75

VOP_NC1 30 40 - 0-40 - 1 9.75

VOP_NC2 45 40 - 0-40 - 1 9.75

VOP_NC3 30 40 - 0-40 - 3 9.75

VOP_NC4 45 40 - 0-40 - 3 9.75

VOP_NC5 30 100 - 0-100 - 10 9.75

VOP_NC6 45 100 - 0-100 - 10 9.75

VOP_NC0330 30 20 - 0-20 - 0.3 26.75

VOP_NC0345 45 20 - 0-20 - 0.3 35.75

VOP_NC7 30 100 - 0-100 - 30 9.75

VOP_NC8 45 100 - 0-100 - 30 9.75

T-bar_NC2 - - 60 - 29-49 1 10.75

4.26

Table 4.3. Constants derived from two sets of T-bar variable rate tests

Parameter (for

Equation 4.4)

Value and reference

House et al. (2001) Watson and Suemasa (2000, unpublished)

a 1 1

b 2.77 2.77

c 4.28 0.57

d 1.25 1.45

Table 4.4. Framework parameters

Parameter Value

Operative normally consolidated strength ratio, (su/σ'v)nc 0.15

Friction factor, μ White and Hodder (2010) 0.7

CSL specific volume at σ'v = 1 kPa, Γ 3.29

Soil sensitivity, St 2.3

Rate of strength degradation, N95 2.5

4.27

(a)

(b)

(c)

(d)

(e)

Figure 4.1. Examples of events involving gross disturbance and reconsolidation of soil: (a) pile

subjected to cyclic lateral loads; (b) spudcan reinstallation; (c) cyclic motion of steel catenary riser;

(d) pipeline lateral buckling induced by temperature effects; (e) submarine slide

Slope failure

Debris flow and

turbidity current

Deposition

4.28

Figure 4.2. (a) Photo of VOP; (b) Schematics of VOP

Figure 4.3. Shear strength profile from initial T-bar penetration

0

10

20

30

40

50

60

70

0 2 4 6

Dep

th b

elo

w m

ud

lin

e, z (

mm

)


T-bar_NC1

T-bar_NC2

su-in = 0.09zT-bar kPa

N = 100 g

4.29

(a)

(b)

Figure 4.4. T-bar cyclic tests results in NC sample: (a) Cyclic tests profiles; (b) T-bar strength

degradation factor extracted from mid-point of cyclic test distance

0

20

40

60

80

100

-8 -6 -4 -2 0 2 4 6 8 10

Dep

th b

elo

w m

ud

lin

e, z (

mm

)


T-bar_NC1

T-bar_NC2

N = 100 g

0.2

0.4

0.6

0.8

1

0 2 4 6 8 10 12

Deg

ra

da

tion

fa

cto

r, s

u-n

/su

-in

Cycle number

T-bar_NC1

T-bar_NC2

Average

St = 2.3

4.30

(a)

(b)

Figure 4.5. Load-displacement response for test with zT = 30 mm: (a) Test VOP_NC3 (v = 3 mm/s);

(b) Test VOP_NC0330 (v = 0.3 mm/s)

-2

0

2

4

0 1 2 3 4 5 6 7 8 9

Ho

riz

on

tal lo

ad

, F

h(N

)

Displacement, u/DVOP

n = 0.25

n > 0.25

-4

-2

0

2

4

0 1 2 3 4 5

Ho

riz

on

tal lo

ad

, F

h(N

)

Displacement, u/DVOP

n = 0.25

n = 0.75 - 4.25

n > 4.25

4.31

Figure 4.6. Evolution of VOP current to initial load ratio with increasing cycle number

Figure 4.7. Normalised steady load across different non-dimensional velocities

0.2

0.4

0.6

0.8

1

1.2

1.4

0.1 1 10 100

Fh

-n /

Fh

-0.2

5

Cycle number

VOP_NC0330 (v = 0.3 mm/s) VOP_NC0345 (v = 0.3 mm/s) VOP_NC1 (v = 1 mm/s)

VOP_NC2 (v = 1 mm/s) VOP_NC3 (v = 3 mm/s) VOP_NC4 (v = 3 mm/s)

VOP_NC5 (v = 10 mm/s) VOP_NC6 (v = 10 mm/s) VOP_NC7 (v = 30 mm/s)

VOP_NC8 (v = 30 mm/s) T-bar

0.5

1

1.5

2

2.5

3

3.5

4

0.01 0.1 1 10 100 1000 10000

No

rm

ali

sed

ste

ad

y l

oa

d, F

hs/

Fh

s-re

f

V = vDVOP/cv

Current data

House et al. (2001)

Watson and Suemasa (2000, unpublished)

4.32

(a)

(b)

Figure 4.8. (a) Shearing and reconsolidation of soil element S; (b) Idealised stress path of soil

element S in v - σ'v space

Sz

zT

σʹvo = γ'z

Cyclic test distance

Elapsed

consolidation

time, tc

v mm/s

N

Γ

Sp

ecif

ic v

olu

me,

v

Vertical effective stress, σ'v (log scale)

A

DE

λ

λ

κ

BC

σ'vn σ'vn + ∆σ'v

∆v

∆ σ'v

σ'vo

NCL

CSL

RSL

B'

4.33

Figure 4.9. Isochrones of excess pore pressure (Terzaghi and Frohlich 1936)

0

0.2

0.4

0.6

0.8

1

0 0.2 0.4 0.6 0.8 1

z/z T

ue-tc / uezT

T = 0.5T = 0.8

T = 2

T = 0.25 T = 0.1

T = 0.05

T = 0

4.34

VOP

σ'vo - σ'vn

2D

VO

P

(a)

(b)

Figure 4.10. (a) Rectangular pore pressure distribution during a VOP pass in plan view following

Bolton (1979); (b) Comparison of the lateral (Bolton 1979) and the vertical (Terzaghi and Frohlich

1936) excess pore pressure dissipation models

0

0.2

0.4

0.6

0.8

1

0.001 0.01 0.1 1 10 100 1000

Deg

ree

of

con

soli

da

tio

n, U

T = cvtc/DVOP2 or T = cvtc/zT

2

Terzaghi and

Frohlich (1936)

z/zT = 0.25

z/zT = 0.75

z/zT = 1

Bolton (1979)

4.35

(a)

(b)

Figure 4.11. Comparison between model and VOP centrifuge data for tests at 30 mm embedment

and velocities of: (a) 1 mm/s (test VOP_NC1); (b) 0.3 mm/s (test VOP_0330)

0

1

2

3

4

0 2 4 6 8 10

Ho

riz

on

tal lo

ad

, F

h-n

(N)

Cycle number

VOP_NC1

Bolton

Terzaghi & Frohlich

Terzaghi & Frohlich (2k)

No reconsolidation

(2κ)

0

1

2

3

4

0 5 10 15 20 25 30

Ho

riz

on

tal lo

ad

, F

h-n

(N)

Cycle number

VOP_0330

Bolton

Terzaghi & Frohlich


No reconsolidation

(2κ)

VOP_NC0330

4.36

(a)

(b)

Figure 4.12. Comparison between model and VOP centrifuge data for tests at 45 mm embedment

and velocities of: (a) 1 mm/s (test VOP_NC2); (b) 0.3 mm/s (test VOP_0345)

0

2

4

6

8

0 2 4 6 8 10

Ho

riz

on

tal lo

ad

, F

h-n

(N)

Cycle number

VOP_NC2

Bolton's dissipation

Terzaghi & Frohlich


No reconsolidation

(2κ)

0

2

4

6

8

10

0 10 20 30 40

Ho

riz

on

tal lo

ad

, F

h-n

(N)

Cycle number

VOP_0345

Bolton

Terzaghi & Frohlich

Terzaghi & Frohlich

No reconsolidation

(2κ)

VOP_NC0345

4.37

Figure 4.13. Stress path of soil element at z = zT for test VOP_NC0345

2

2.1

2.2

2.3

2.4

2.5

2.6

2.7

2.8

2.9

3

3.5 35

Sp

ecif

ic v

olu

me,

v

Vertical effective stress, σ'v (kPa)

Simulated test stress path

NCL

CSL

RSL

n = 36.25 to 400.25

27 kPa

σ'v0.25 (4.79 kPa) σ'vo

σ'vcyc-drained

4.38

Figure 4.14. Settlement of soil surface at VOP test sites

Figure 4.15. Extrapolated evolution of the normalised effective stress at z = zT based on test

VOP_NC0345

VOP

Settlement of

test area

0

1

2

3

4

5

0.1 1 10 100 1000

σ' v

n /

σ' v

-0.2

5

Cycle number

4.67

5.1

CHAPTER 5. THE STRENGTH PROPERTIES OF

ULTRA-SOFT KAOLIN CLAY

FOREWORD

This chapter is specifically devoted to understanding the cyclic undrained strength

behaviour of clay across a wide range of liquidity indices - much higher than that found

in the literature. This aspect is important for cyclic pipeline-surficial soil and

submarine slide-pipeline interactions where the moisture content of the soil often

exceeds its liquid limit. Correlations between the strength degradation parameters for

the exponential decay equation – shown previously in Chapters 2 (Equation 2.2) and 4

(Equation 4.9) and the index properties of clay are explored.

After progressing from using the T-bar penetrometer to usage of the VOP in chapters 2

to 4, this chapter reverts to the former tool. This is done deliberately because the T-bar

is geometrically and mechanically similar to a section of offshore pipelines, therefore

allowing the proposed empirical formulas to be potentially utilised for various pipeline-

soil interactions that involves an exponential decay of the soil undrained strength. Apart

from addressing Objective 2 (see Section 1.2.2), this chapter also serves as a prelude to

Chapter 6. It documents thoroughly the method used for interpreting the very small

magnitude of undrained strengths of the samples outlined in the next chapter.

5.2

ABSTRACT 5.1

Geotechnical design considerations for offshore pipelines, piled foundations and

submarine slides involve assessment of the strength of fine-grained soils, and the

degradation of that strength with disturbance and remoulding. For slides and pipelines,

the relevant strength may be very low, relating to near-surface soils and high levels of

remoulding. It is well-known that soft soils exhibit a loss of strength when disturbed,

but the development of this brittleness through the consolidation process has not been

examined previously, and it is not clear how the degradation properties vary with

consolidation and strength. This paper describes a series of centrifuge tests on kaolin

samples consolidated from slurries with an initial voids ratio of 4.0. This provides a

convenient method of investigating the evolution of soil strength degradation behaviour,

and its connection with the changing water content, effective stress and strength during

large-strain consolidation. A total of 81 cyclic T-bar tests were conducted in samples

with shear strengths ranging from 0.08 to 1.7 kPa (reflecting the various stages of

consolidation). Large-strain consolidation numerical analyses were used to assist the

interpretation of the T-bar results. The cyclic T-bar test data showed that the soil

ductility (a parameter controlling the rate of strength degradation) can be correlated

directly with the liquidity index of kaolin. The ductility-liquidity index data obtained

from other T-bar field and centrifuge tests on various natural and reconstituted clay

samples shows good agreement with that of the current data, leading to a linear

correlation between soil ductility and liquidity index. The proposed ductility-liquidity

index is coupled with a previously published sensitivity-liquidity index relationship for

natural clays to produce a T-bar cyclic strain softening model. In turn, because the

sensitivity is a function of the liquidity index, the intact soil strength is linked to the

remoulded strength obtained from laboratory (e.g. fall cone or miniature vane test), and

simple index tests. The resulting correlations provide new insights into the source and

characteristics of the strength degradation behaviour of ultra-soft soils. These provide an

improved basis to incorporate softening effects into simulations of submarine slide

runout and models for soil-structure interactions that involve intense remoulding.

5.3

INTRODUCTION 5.2

Importance of quantifying strength degradation 5.2.1

The study of the soil strength degradation is an important aspect for many geotechnical

analyses of subsea structures for offshore hydrocarbon developments. The cyclic nature

of both environmental and operational forces on subsea pipelines and piled foundations

requires proper quantification of the behaviour of soil under intense cyclic shearing.

Similarly, the behaviour of submarine slides during runout, which involves intense soil

remoulding within the slide material, also requires an understanding of this behaviour.

Cyclic penetration and extraction tests using either the ball or T-bar penetrometer offer

a reliable method to quantify both the sensitivity of the soil and the rate at which it

degrades to the fully remoulded steady state (Yafrate and DeJong 2005; Boylan et al.

2007; Randolph et al. 2007; Gaudin and White 2009; Yafrate et al. 2009; White and

Hodder 2010).

Lately, new methods that link the soil degradation, as observed from penetrometers, to

soil-structure interactions and the behaviour of submarine slides have been developed.

Cheuk and White (2011) developed a model to predict the embedment of pipelines

during laying process – where the laying process involves cyclic motion of the pipeline

due to vessel motion and hydrodynamic loading. Their model utilises yield envelopes

and a flow rule to predict the trajectory of the pipeline. The size of the yield envelope is

controlled by the bearing capacity, which varies with the operative shear strength of the

surrounding soil. In turn, the soil strength changes during the laying process because of

the cyclic motion of the pipeline which remoulds the soil. They proposed that the cycle-

by-cycle change in the strength of the soil around the pipeline can be linked to a

conventional cyclic T-bar test owing to the similar cylindrical geometry of the T-bar.

Other cyclic soil-structure interaction behaviours that may be predicted (following

development of more robust models) using cyclic T-bar tests includes estimating the

degradation of pile-soil (p-y) lateral stiffness (Zhang et al. 2011) and the change in

riser-seabed stiffness at the touchdown zone after episodes of remoulding and

reconsolidation (Hodder et al. 2009). Besides applications for cyclic soil-structure

5.4

interactions, predictions of submarine slide runout behaviour using a depth averaged

numerical model (Boylan and White 2011) can also be calibrated to incorporate strength

softening using cyclic T-bar strength degradation data.

In general, the strength degradation observed during a cyclic T-bar penetrometer test

can be captured using the following strain softening law (Einav and Randolph 2005):

953(n 0.25)/Nu n

u in t t

s 1 11 e

s S S

5.1

where St is the apparent sensitivity of the soil (as measured from a T-bar penetrometer)

defined as the ratio of the strength measured from the first T-bar penetration to the fully

remoulded strength measured after 10 to 20 cycles of remoulding, depending of the

nature of the soil. su-n is the current measured strength during a particular cycle number,

n. The cyclic numbering system, as established by Randolph et al. (2007), is shown

schematically in Figure 5.1. If the soil element completely passes through the full-flow

zone during penetration, the cycle number increases by 0.5. A reversing event

(extraction) causes the cycle number to increase to 1. Since the full-flow mechanism is

symmetrical above and below the T-bar, the measured resistance is assumed to be a

function of the strength at the mid-height of the T-bar. Therefore, the first penetration is

taken as cycle number 0.25 (half way between cycle number 0 to 0.5) and the

subsequent extraction is taken as cycle number 0.75 (half-way between cycle numbers

0.5 and 1). The parameter N95 in Equation 5.1 is the number of T-bar cycles required to

achieve 95% strength reduction from the initial to the remoulded strengths. Parameters

St and N95 defines the shape of a T-bar strength degradation curve, where the former

dictates the fully remoulded resistance around the penetrometer whereas the latter

controls the rate at which the strength degrades to this steady state (i.e. the ductility of

the soil). These two parameters are used in models to predict the as laid embedment of

pipelines (Cheuk and White 2011) and to numerically predict a submarine slide runout

(Boylan and White 2011) as discussed earlier.

It should be noted that the parameters in Equation 5.1 are normally derived assuming

that the strength can be back-calculated from the T-bar resistance using an ideal elastic-

5.5

perfectly plastic bearing factor of 10.5 - representing a bar of intermediate roughness

(Randolph and Houlsby 1984; Stewart and Randolph 1991; Martin and Randolph 2006).

In actual fact, the bearing factors for both intact and remoulded soils are different

because of the combined effects of strain-softening and strain rate, ultimately leading to

different values of the true sensitivity, St-r as compared to the apparent T-bar sensitivity,

St (Yafrate et al. 2009; Zhou and Randolph 2009a; Low et al. 2010; DeJong et al. 2011).

This aspect will be discussed further in latter sections.

Consolidation at very low stresses and through large-strains 5.2.2

The strategy adopted in this paper to study the strength evolution and strength

degradation of ultra-soft clay involves self-weight consolidation of samples of kaolin

slurry, whilst concurrently testing the strength properties. At this low strength level, the

soil stiffness and permeability are highly non-linear, and consolidation involves large-

strains.

To back-analyse the changing strength it has been necessary to model this consolidation

process, taking correct account of large-strain consolidation effects. The study of large-

strain consolidation of soil originally deposited from a slurry has long been a topic of

interest of many researchers. Previous work applications concerning large-strain

consolidation focused mainly on mine tailings (e.g. Toh 1992; Seneviratne et al. 1996)

and land reclamation (e.g. Bo et al. 2005). The geotechnical characteristics of a soil,

such as voids ratio, shear strength and stiffness, change significantly as it consolidates

from a slurry to a soft clay.

Previous experiments have attempted to quantify the changes in undrained shear

strength of a consolidating soil (e.g. Toh 1992; Robinson et al. 2003) using the vane

shear, cone or T-bar penetrometer test. However, these tests were conducted in samples

at relatively high strengths compared to the present study, with the weakest detected

strength corresponding to approximately 1 – 2.5 kPa. Other studies have attempted to

characterise the early stages of soil consolidation through X-ray measurement of the

changing density, although these provide no direct evidence of the changing strength

(Been and Sills 1981).

5.6

Objectives 5.2.3

The first motivation of this paper is to present a method for interpreting the changing

strength of a sample as it is consolidating from a slurry, especially in the early stages of

consolidation. Since the T-bar tests were performed in underconsolidated ultra-soft

kaolin samples, a particular corrective procedure was used to interpret the measured

resistance. A complementary numerical large-strain consolidation model was then used

to validate this corrective procedure. This paper also shows the value of performing a

non-linear numerical simulation of the progressive consolidation of the entire bed of

soil, incorporating non-linear relationships between permeability, stiffness and effective

stress level. This highlights the limitations of standard consolidation solutions, in which

the mechanical properties are assumed to be invariant.

The second motivation of this study is to reveal the strength degradation characteristics

of ultra-soft kaolin – at different strengths, reflecting different levels of consolidation –

through cyclic T-bar tests. The aim here is to provide a better understanding of the soil

parameters that govern strength degradation, which may lead to improved models for

cyclic soil-pipeline interactions and submarine slide runout simulations. Links between

the strength degradation properties observed during cyclic T-bar tests and conventional

soil index properties are explored. These links are then compared to that obtained from

other centrifuge and field T-bar cyclic test studies.

PROPERTIES OF KAOLIN AFFECTING CYCLIC T-BAR 5.3

RESISTANCE

Overview 5.3.1

This section briefly summarises some of the important mechanical properties of UWA

kaolin employed in the centrifuge tests reported in this paper. These properties are used

in the numerical back analysis described in Section 5.6, which facilitates proper

interpretation of the T-bar test results.

5.7

Index properties 5.3.2

Boukpeti et al. (2012) determined the liquid limit (LL) and plastic limit (PL) of UWA

kaolin using testing procedures in accordance to AS 1289.3.9 (1991) and AS 1289.3.2.1

(1995) respectively. The LL and PL values are shown in Table 5.1.

Compressibility and permeability 5.3.3

During self-weight consolidation of kaolin slurry, the excess pore pressure dissipates,

resulting in an increase in the in-situ vertical effective stress, σvo. This in turn, causes

the voids ratio of the soil, ev, to decrease according to:

v c c 10 voe e C log σ' 5.2

where ec is the voids ratio at σ'vo = 1 kPa, and Cc is the compression index.

At low σvo, Equation 5.2 may not be applicable (Carrier et al. 1983; Toh 1992;

Seneviratne et al. 1996). An alternative ev-σvo relationship that can represent the large

range of voids ratios associated with centrifuge consolidation of a sample originally

from a slurry state may be written as:

1

v 1

b

voe a (σ' ) 5.3

Equation 5.3 is commonly used to characterise mineral wastes (Carrier et al. 1983) and

mine tailings (Seneviratne et al. 1996). Using data from moisture content samples

obtained from fully consolidated samples of UWA kaolin, in both the beam centrifuge

(White et al. 2008, unpublished) and the drum centrifuge (Wu 2008) (Figure 5.2), the

best fit a1 and b1 values are shown in Table 5.1.

The rate of consolidation is a function of the permeability-voids ratio relationship

(kv-ev). The kv-ev relationship for UWA kaolin, obtained from Rowe cell tests

(Richardson 2007, unpublished) (Figure 5.3) can be expressed in the form commonly

used for mineral wastes (Carrier et al. 1983):

5.8

2b

vv 2

v

ek a

1 e

(m/day) 5.4

The best fit a2 and b2 values are presented in Table 5.1.

Undrained strength ratio 5.3.4

Assuming that one-dimensional conditions apply during consolidation in the centrifuge,

the undrained shear strength of clay, su, will vary with depth according to the

relationship proposed by Ladd et al. (1977) and Wroth (1984):

0.8uu vo

v nc

ss σ' OCR

σ'

5.5

where (su/σv)nc is the normally consolidated undrained strength ratio and OCR is the

overconsolidation ratio. In the case of a consolidating sample, OCR can be taken as

unity. The value of (su/σv)nc shown in Table 5.1 is deduced from simple shear tests

(Lehane et al. 2009).

TEST METHODOLOGY 5.4

Overview of test setup 5.4.1

The centrifuge testing programme involved the consolidation of 13 samples in the UWA

drum centrifuge (Stewart et al. 1998). The test setup is illustrated schematically in

Figure 5.4. Each sample was characterised through a series of T-bar penetrometer tests

performed at intervals throughout consolidation, at different locations within the drum

centrifuge channel. No drainage layer was installed at the base of the drum channel, so

pore water was only allowed to drain from the top of the sample. In each sample, the

change in pore pressure as consolidation progressed was monitored using a pore

pressure transducer (PPT) placed at the base of the drum channel.

5.9

Sample preparation 5.4.2

Each sample was prepared by first mixing dry kaolin powder with water to achieve a

moisture content of 155%. Because this experimental programme requires that all 13

samples have the same consolidation response, the amounts of water and dry kaolin

were carefully measured. Each slurry was then mixed in a barrel mixer for at least 3

hours. Moisture content measurements of the well-mixed slurries (for all 13 samples)

indicated satisfactory consistencies with moisture contents in the range 154%-156%

(compared to a targeted moisture content of 155%).

After filling the slurry into the drum centrifuge channel to a target height of 184 mm,

each sample was spun at 40 g. Both the slurry filling process and subsequent sample

settlements, ∆ht, were monitored using a scale and a video camera (see Figure 5.4 and

Figure 5.5). Pictures of the sample surface were used to derive changes in the sample

height with time. The water level was maintained at 200 mm (height limit of the

centrifuge channel) throughout testing in each sample.

Minimising stress error in consolidating soil 5.4.3

In order to minimise the stress error within a centrifuge soil model (owing to the non-

linear acceleration field produced in the centrifuge), it is common practice to adopt an

effective radius, Re as the distance between the central axis of the centrifuge to one-third

depth position within the model (Schofield 1980). However, this recommendation is

based on soil model with negligible settlement during the testing period, and may not be

applicable to the current test programme owing to the marked changes in sample height

during large-strain consolidation. Minimisation of the stress errors allows the results

from centrifuge tests in a consolidating sample to be compared with numerical

simulations of the prototype event reliably as will be discussed in later sections. This

section outlines the steps employed to minimise the stress errors within a consolidating

soil following the approach of Toh (1992).

The relative total stress error, Er within a centrifuge soil model can be calculated as

(Toh 1992):

5.10

z

v m v p

0r z

v p

0

σ σ dz

E

σ dz

5.6

where z is the depth below the current mudline, σv-m is the vertical stress in the model

soil and σv-p is the prototype vertical stress. σv-m can be calculated from:

v m t

e

ρNgz zσ R

R 2

5.7

where N is the ratio of the centrifuge centripetal acceleration to Earth’s gravity, g (taken

as 9.81 m/s2), ρ is the soil density, Re is the effective radius and Rt is the distance from

the centrifuge central axis to the current mudline of the consolidating sample. In turn,

σv-p can be simply calculated as:

v pσ Nρgz 5.8

By substituting Equation 5.7 and Equation 5.8 into Equation 5.6, and setting z = ht

(where ht is the current sample height), Er can be written as follows:

t tr

e e

R hE 1

R 3R 5.9

The profiles of σv-m - σv-p (assuming ρ = 1400 kg/m3) with depth, when Re is fixed at

one third (measured from the mudline) of the initial height of the sample (Re = Rt-i +

1/3ht-i, where Rt-i is the distance from the central axis to the initial mudline level and ht-i

is the initial sample height), for the conditions of 0% and 50% settlements, are depicted

in Figure 5.6(a) and Figure 5.6(b), respectively. Here, the dimensions related to the

centrifuge testing reported in this paper are used, where the drum centrifuge channel

depth and radius to the bottom of the channel are 0.2 m and 0.6 m respectively. ht-i is

taken as 0.184 m (the same as the test condition). For the case of no settlement (Figure

5.11

5.6(a)), the σv-m - σv-p is zero at 2/3ht-i below the initial mudline, which is similar to

Schofield (1980). After 50% settlement (Figure 5.6(b)), it can be seen that σv-m exceeds

σv-p at all depths. This will cause a stress error in the model sample which may affect

any numerical comparison with the centrifuge data.

The stress error profile may be compared more conveniently by plotting Er against the

consolidation strain, ∆ht/ht-i (where ∆ht is the sample settlement), the magnitude of Er

changes with increasing sample settlements. This is illustrated in Figure 5.7 where the

changes in Er with sample consolidation strain, ∆ht/ht-i for Re = Rt-i + 1/3ht-i, Rt-i + 1/2ht-i

and Rt-i + 2/3ht-i are plotted based on the dimensions of the UWA drum centrifuge with

ht-i of 0.184 m (and therefore Rt-i = 0.416 m).

As shown, if Re is located at Rt-i + 1/3ht-i (Schofield 1980), the decreasing sample height

will cause the overstress error to increase. When the sample is at 50% of its initial

height, this will cause an overstress of approximately 13%. The plot of Er-∆ht/ht-i with

Re fixed at Rt-i + 2/3ht-i shows that the maximum understress is more than 11% at the

start of consolidation. Although the error decreases at higher ∆ht/ht-i (reduces to zero at

50% sample settlement), the majority of the centrifuge consolidation time at ∆ht/ht-i <

50% will have relatively large stress errors. From Figure 5.7, it is evident that the Re

should be located at Rt-i + 1/2ht-i with a maximum Er of approximately ±6% if ∆ht/ht-i is

expected to vary from 0-50%.

The parametric study shown in this section is limited to the UWA drum centrifuge.

Obviously, modelling large-strain consolidation in a centrifuge with larger radius will

further reduce Er, albeit still being at its minimum when Re = Rt-i + 1/2ht-i (if ∆ht/ht-i is

expected to vary from 0-50%). It should be noted that the ρ is assumed to be constant in

the calculations of Er (which is not valid for large-strain cases). However, Toh (1992)

reported good agreement between the centrifuge results with large-strain consolidation

analyses (assuming constant N with depth) when Re is fixed at Rt-i + 1/2ht-i thus,

supporting this simplified assumption of constant ρ. A more complicated numerical

treatment of this subject as reported by Fox et al. (2005), also validates this simplified

parametric study.

5.12

Based on the optimised Re reported in this section, all 13 tests were conducted by

setting Re = Rt-i + 1/2ht-i.

T-bar tests 5.4.4

A T-bar penetrometer was used to provide a continuous penetration resistance profile

within each consolidating sample. A larger bar measuring 10 × 40 mm (diameter (DT-bar)

× length (LT-bar)) was used instead of the usual smaller bar (typically 5 × 20 mm)

commonly utilised in centrifuge tests. This was to provide better resolution of the

penetration resistance, especially in samples at very low degrees of consolidation. A set

of axial strain gauges located immediately above the T-bar is used to measure the

penetration resistance, qT-bar, which can be subsequently converted to the undrained

shear strength of the soil (su) via the following equation:

T baru

T bar

qs

N

5.10

where NT-bar is the T-bar bearing factor, which is usually taken as 10.5 for a T-bar with

an intermediate roughness (assuming ideal elastic-perfectly plastic soil) during which a

full-flow mechanism takes place at failure (Randolph and Houlsby 1984; Martin and

Randolph 2006).

A total of 81 T-bar tests were performed, with a penetration rate of 2 mm/s used in all

cases. Details of the T-bar tests conducted at different degrees of consolidation, U%, are

provided in Table 5.2. U% is calculated as follows:

t

u

hU% 100

h

5.11

where ∆hu is the total settlement of the sample. As mentioned in later sections, ∆hu was

found to be 73 mm (40% of the initial height of 184 mm).

5.13

A minimum distance of 3.5DT-bar (measured from the centre of the previous to the

centre of the next test regions) was allowed between two T-bar tests to minimise

interference between test sites. Penetrations were conducted to normalised depths,

z/DT-bar, ranging from 8 to 13.1. All T-bar tests involved multiple penetration and

extraction cycles to measure the sensitivity of the soil. In Sample 1, the T-bar tests and

sample settlement observations were continued until no further settlement was

detectable. In all other tests, only partial consolidation had occurred by the end of the

test, although in some cases this amounted to U > 95%. Although no T-bar tests were

conducted in samples EET and Sample 2 because of technical problems, the changes in

sample heights were monitored (see Figure 5.8).

CENTRIFUGE TEST RESULTS 5.5

All centrifuge test results in this paper are presented in model scale units unless stated

otherwise.

Average degree of consolidation, U% 5.5.1

An image processing software, AnalyzingDigitalImages (Pickle 2008) was used to

process the photos taken during sample consolidation. Figure 5.8 shows the evolution of

the average degree of consolidation, U% (Equation 5.11), for all 13 tests. The total

settlement, ∆hu is taken as the ultimate settlement of Sample 1, which was 73 mm, from

an initial sample height of 184 mm. From Figure 5.8, it can be seen that the U% profiles

for all the tests fall within a unique and very narrow band. This confirms that the

changes in the sample properties as consolidation progresses were similar for all the

tests, and further confirms the consistency of the sample preparation method.

Correction of T-bar resistance using cyclic data 5.5.2

In all the T-bar tests, cycles of penetration and extraction were performed. Besides

investigating the strength degradation behaviour, these cyclic tests were used to correct

the shear strength derived from the initial T-bar penetration, su-in, at a depth of 50 mm

below the soil surface, especially for tests in samples of low U%. A cyclic correction

5.14

example is shown in Figure 5.9 from test T4-1 in Sample 4 at U ~ 67%. The

uncorrected cyclic data appears to have a strong tensile skew, which resulted in an

oscillatory strength degradation pattern as shown in Figure 5.10(a).

The strength degradation factor is defined as the shear strength during a particular cycle

number of penetration or extraction phase, su-n, normalised by the initial strength

derived from the first T-bar penetration, su-in. The cyclic numbering system follows that

discussed in Section 5.2.1. Note that the strength used to derive the degradation factors

in Figure 5.10 are taken at a distance of 50 mm from the soil surface to the mid-height

of the T-bar.

The cyclic T-bar penetration and extraction response was used to correct the data as

recommended by Randolph et al. (2007), Boylan et al. (2007) and Yafrate et al. (2009).

A correction factor comprising an adjustment that varies linearly with depth is applied

until the remoulded extraction and penetration resistance is symmetrical about the zero

line, at least at the depth of interest (Figure 5.9). This then leads to a smooth strength

degradation pattern where the sample is completely remoulded after only ~ 2 cycles, as

depicted in Figure 5.10(b).

This corrective method is not entirely satisfactory because the response remains

asymmetrical especially at shallow depths. The factors which contribute to the resultant

resistance (qT-bar) are shown schematically in Figure 5.11 to Figure 5.13. As the T-bar

enters the soil – which is denser than water – a sudden increase in the measured

resistance, qT-bar is evident due to the increasing buoyancy, qb. Furthermore, as the T-bar

penetrates into deeper and denser soil, this buoyancy force will increase, counteracting

the self-weight of the T-bar, qw (which itself increases slightly as the bar is moved to a

greater radius within the centrifuge channel).

In addition, the lateral pressure on the axial strain gauged region results in an apparent

tensile contribution (qL) to the total resistance (see Figure 5.11 to Figure 5.13). This is

due to Poisson’s ratio effects at the shaft (which are complicated by the composite

action of the aluminium shaft and the epoxy resin used to coat the gauge).

5.15

These effects are difficult to quantify, but the net effect on the measured uncorrected qT-

bar profile can be seen when the qT-bar changes from compressive to tensile during the

first penetration at a depth of around 16 mm (Figure 5.9). A simple linear corrective

offset with an intercept at the soil surface (Figure 5.13) may be more appropriate

compared to the cyclic correction as outlined above. However, besides the difficulty in

quantifying the effects of qL, estimating the qb in a consolidating soil is complicated

because the effective unit weight changes with both time (due to consolidation) and

depth (due to the radial variation in g-level). Since qb, qL and qw act in the same

directions regardless of the current movement of the T-bar (penetration or extraction),

the cyclic corrective method is sufficient to estimate the strength at a particular depth (in

this paper: 50 mm below the soil surface) by taking advantage of symmetry. This

isolates the soil strength component (qsoil), which is the only component of resistance

that always acts to oppose the current direction of T-bar motion.

VALIDATION OF CYCLIC CORRECTION 5.6

General numerical validation procedure 5.6.1

To validate the T-bar correction method, and to provide additional data of the soil

parameters during consolidation, the measurements from different stages of

consolidation are compared with those derived from numerical analyses of the

consolidation process using a single set of material parameters, applicable across the

entire consolidation process.

The classical one-dimensional consolidation solution of Terzaghi and Frohlich (1936),

which assumes small-strain conditions cannot yield accurate results when applied to

large-strain consolidation cases (Lee and Sills 1981; Fox and Berles 1997). To simulate

the large-strain consolidation of the centrifuge samples, a finite element programme,

MinTaCo (Mine Tailings Consolidation) was used to back analyse the shear strength

changes of the consolidating sample. MinTaCo was originally developed to model one-

dimensional consolidation of tailings using the large-strain consolidation equation

proposed by Gibson et al. (1967) and an updated Lagrangian coordinate system. A

complete description of MinTaCo is provided by Seneviratne et al. (1996).

5.16

To run the programme, a number of input parameters are required. The programme may

be utilised to model large-strain consolidation in the centrifuge by adjusting any linear

dimensions to correspond to the prototype scale. The programme captures the non-linear

compression and permeability behaviour using the following parameters (only the

parameters required for the current analyses are listed):

1. Initial sample height, ht-i (in this case, ht-i = 7.36 m (prototype scale))

2. Drainage boundary conditions (can be set as either one-way or two-way).

3. The specific gravity of the soil, Gs (taken as 2.6 for the UWA kaolin).

4. Voids ratio upon sedimentation, eo (at the onset of consolidation).

5. ev-σvo (voids ratio – in-situ vertical effective stress) relationship as shown in

Equation 5.3.

6. kv-ev (permeability – voids ratio) relationship as shown in Equation 5.4.

Calibration of numerical analysis 5.6.2

All of the MinTaCo simulations were conducted with the assumption that the

gravitational field within the sample is constant at Ng. In the drum centrifuge tests, no

drainage was allowed at the bottom of the samples. Therefore, the soil surface is the

only drainage boundary in the MinTaCo simulations, leading to one-way drainage. As

depicted in Figure 5.2, the a1 and b1 parameters (see Equation 5.3 and Table 5.1) were

derived based on measurements of voids ratios relevant to the current centrifuge

consolidation tests (large voids ratios). However, the kv-ev parameters (a2 and b2 –

Equation 5.4 and Table 5.1) were derived from Rowe cell tests at a small and limited

range of voids ratios (< 2). This may cause the numerically simulated consolidation

response to be slower than that in the centrifuge tests. To investigate this aspect further,

2 cases of MinTaCo simulations were conducted (named Case x_drum). The soil

parameters used in each case are summarised in Table 5.3(a). The details of each

simulation case are described briefly as follows:

1. Case 1_drum - the a1 and b1 parameters (see Equation 5.3 and Table 5.1)

correspond to those obtained from centrifuge tests while the a2 and b2 parameters

(see Equation 5.4 and Table 5.1) were obtained from the Rowe cell test. e0 was

5.17

set to 4 (initial moisture content, wi = 155%). As shown in Figure 5.14, although

the final height of the sample is predicted well, the rate of sample settlement of

Case 1_drum is much slower than that measured in all 13 samples. The a2 and b2

parameters derived from the Rowe cell may not reflect the correct kv-ev

relationship at large voids ratios since the Rowe cell test (Figure 5.3) reported in

this paper was conducted at voids ratios ranging only from 1 - 1.8. The kv-ev

relationship can be alternatively determined by matching the consolidation

response obtained from numerical analysis to that measured (e.g. Williams and

Tanaka 1991; Toh 1992). As such, the permeability parameters were back-

calculated in the subsequent simulation case.

2. Case 2_drum - All the input parameters of Case 1_drum were maintained except

the kv-ev parameters. In this simulation, a2 and b2 were varied until the simulated

sample height-time profile matched the drum centrifuge measurements. A good

match is achieved using the parameters a2 and b2 of 0.00015 and 4 respectively

(Figure 5.14). The kv-ev relationship calculated from these back-calculated

parameters (via Equation 5.4) is plotted in Figure 5.3 as a comparison to that

obtained from the Rowe cell test.

Validating calibrated numerical parameters 5.6.3

Following the approach taken by Toh (1992), the validity of the a2 and b2 parameters for

the permeability - voids ratio relationship (Equation 5.4) can be assessed by comparing

the simulated sample height-time profile with that measured in the case of a large-strain

consolidation with two-way drainage condition.

White et al. (2008, unpublished) conducted a large-strain consolidation test on a kaolin

slurry (eo = 3.12) in the UWA beam centrifuge at 100 g with drainage allowed at the top

and bottom of the sample (two-way drainage). The sample had an initial height of

210 mm and the settlement was monitored using a laser proximity sensor. The change in

sample height as consolidation progressed is plotted in Figure 5.15. During testing, the

beam centrifuge was stopped at approximately 450 minutes after the start of

consolidation. Unfortunately, after the centrifuge restart, the laser sensor gave

5.18

misleading readings, and therefore, the data after 450 minutes of consolidation have

been omitted from Figure 5.15.

For the MinTaCo simulation, the same a1, b1, a2 and b2 values were used as for Case

2_drum, with eo = 3.12 as the initial voids ratio (Case 1_beam – Table 5.3(b)). The

simulated rate of sample height decrease (consolidation rate) is in good agreement with

the measured profile (Figure 5.15). This validates the a2 and b2 parameters back-

calculated from the drum consolidation tests. Also plotted in Figure 5.15 is the Case

2_beam simulation response where the Rowe cell derived a2 and b2 parameters were

used (Table 5.3(b)). As expected, MinTaCo predicted a much slower consolidation rate.

Base excess pore pressure 5.6.4

The excess pore pressures measured at the base of the drum centrifuge channel in

Sample 4, 5 and 6 together with the predicted excess pore pressures from MinTaCo

analyses Case 1_drum and Case 2_drum are plotted in Figure 5.16. The excess pore

pressure data for the remaining tests are omitted due to measurement disturbances

arising from halting the centrifuge tool-table whilst the channel continued spinning.

Similar to the predicted sample height-time profile, the dissipation of the excess pore

pressure is much slower for Case 1_drum compared to the measured ones. The close

agreement of Case 2_drum with the measured excess pore pressures in the three tests

support the use of the back-calculated a2 and b2 permeability parameters. The initial

excess pore pressures measured in Sample 4, 5 and 6 ranged between 23-24 kPa. These

are consistent with the theoretical initial excess pore pressure of 23 kPa (because eo =

4).

Evolution of undrained shear strength and voids ratio 5.6.5

Figure 5.17 depicts the last two shear strength profiles (T1-15 and T1-16) obtained from

the first T-bar penetration (su-in), in Sample 1. These tests were conducted 7 and 9 days

after the start of consolidation. The two shear strength profiles have negligible

differences, indicating that the sample was effectively at full primary consolidation after

7 days. It can be seen that multiplying the predicted fully consolidated vertical effective

5.19

stress, σvo, obtained from Case 2_drum simulation with the normally consolidated

strength ratio, (su/σv)nc = 0.17 resulted in a good match with the T-bar data. Although

this strength ratio is consistent to those reported by Hodder et al. (2010) and Sahdi et al.

(2010) (both conducted centrifuge T-bar tests on the UWA kaolin), this strength ratio is

lower than the typical (su/σv)nc = 0.25 obtained from element test (Section 5.3.4). The

strength derived from T-bar test is usually somewhat lower than the intact value –

corresponding to that inferred from element testing – owing to the significantly higher

level of strain softening in the soil around the T-bar probe (Einav and Randolph 2005;

Zhou and Randolph 2007; Zhou and Randolph 2009b). Disregarding the disparity

between the operative strain rates of the T-bar and laboratory test, the softening effect is

essentially negligible in the latter (Randolph et al. 2007), and therefore the derived

strength is higher than that obtained from a T-bar – leading to higher (su/σv)nc.

The su-in for all 81 T-bar tests are compared to the strength evolution predicted from

MinTaCo Case 2_drum simulation in Figure 5.18. Both measured and predicted

strengths are taken from a depth of 50 mm below the surface. The predicted strengths

were calculated by multiplying the predicted σvo by (su/σv)nc = 0.17. Generally, the

predicted strength-time profile is reasonably close to that measured from T-bar tests.

This validates the cyclic correction procedure.

By combining Equation 5.3 and Equation 5.5 and substituting OCR = 1 (for a

consolidating soil) and (su/σv)nc = 0.17, the changes in voids ratio, ev, as the soil

consolidates may be calculated as follow:

1b

u inv 1

se a

0.17

5.12

The predicted voids ratio-time profile (Case 2_drum) is compared with that calculated

from su-in (Equation 5.12) in Figure 5.19. Similar to the su-in-time profile, MinTaCo

predicts a slightly faster consolidation rate (reflected by the higher rate of decrease in

voids ratios) between 800 and 2500 minutes after the start of consolidation.

5.20

EFFECT OF WATER CONTENT ON SENSITIVITY, 5.7

DUCTILITY AND REMOULDED STRENGTH

Sensitivity and ductility 5.7.1

To investigate the evolution of sensitivity, St and ductility, N95 with consolidation, the

degradation curves (similar to that shown in Figure 5.10(b)) for all 81 cyclic T-bar tests

were fitted with Equation 5.1 by optimising the St and N95 parameters. Two examples of

the optimised degradation curves are depicted in Figure 5.20 for tests in U = 51% and

100%. The optimised St and N95 parameters are plotted against the increase in average

degree of consolidation, U%, in Figure 5.21.

There is a tendency for relatively high and somewhat scattered St at U < 60% (Figure

5.21(a)) and this may be attributed to the reduced accuracy of the T-bar tests at low U%.

Accordingly, the scatter in St appears to decrease at higher U%. Generally, the St-U%

data points are tightly bunched with a mean St ~ 2.0. This suggests that, for

reconstituted kaolin, the St is not affected by the water content. This is in contrast with

natural clays, which may contain interparticle cementation or exists as metastable fabric

and often show trends of increasing sensitivity with liquidity index (Mitchell and Soga

2005).

Although some degree of scatter can also be seen in the N95-U% plot (Figure 5.21(b)), a

trend of increasing ductility (N95) with increasing U% is evident. To reflect the effects

of the decrease in water content within the soil voids as consolidation progresses, N95 is

plotted with the corresponding liquidity index, LI in Figure 5.22. The corresponding

water content used to determine LI is calculated from the voids ratio-initial strength

relationship via Equation 5.12. As more water is expelled from the voids (decreasing

LI), the soil densifies resulting in more interparticle contacts. This increases the soil

strength and stiffness, thus requiring more T-bar cycles to remould the soil sample (N95

increases). The increase in N95 with decreasing LI can be represented by the following

linear fit (Figure 5.22):

5.21

95 d dN LI 5.13

This relationship is deduced from all the 81 T-bar cyclic tests in samples of different

U%. The resulting least square analysis gave a reasonable coefficient of determination,

R2, of 0.80. The best fit values of αd and βd are shown in Table 5.4.

Also plotted in Figure 5.22 are supplementary data from previous field and centrifuge

T-bar cyclic tests. The field data were obtained from those reported by Yafrate and

DeJong (2005) (Louiseville, Gloucester and OnsØy clays) and Boylan et al. (2007)

(Bothkennar clay). The data for West African clay was obtained from centrifuge T-bar

cyclic test (White et al. 2008, unpublished). In these tests, the N95 was fitted to each

measured strength degradation data using Equation 5.1. It can be seen that both field

and centrifuge data follow the N95-LI relationship fitted to the present study (kaolin)

reasonably well.

Remoulded shear strength 5.7.2

The relationship of the remoulded strength, su-rem, with the corresponding LI (calculated

based on the moisture content derived from the voids ratio – Equation 5.12) is shown in

Figure 5.23. It should be noted that each remoulded strength is the average of the last

three T-bar cyclic strengths (once the degradation curve becomes relatively steady). The

data is fitted using a power function of the form (R2 = 0.97):

r β

u rem rs α (LI) 5.14

where the best fit αr and βr constants are shown in Table 5.4. Also plotted in Figure 5.23

are the su-rem-LI trend lines reported by Leroueil et al. (1983) and Boukpeti et al. (2012)

(including their experimental data) for fully consolidated soils. The former derived the

su-rem-LI relationship based on more than 11 natural clay samples with LI of ~ 0.4 - 2.5

using the fall cone test while the latter used the vane shear and viscometer tests to derive

the su-rem-LI relationship of UWA kaolin (LI ~ 0.8 – 6). Leroueil et al. (1983) fitted their

data using the following relationship:

5.22

u rem 2

1s

(LI 0.21)

5.15

Boukpeti et al. (2012) conducted both vane shear and viscometer tests on remoulded

UWA kaolin samples and fitted their data with a similar power function (Equation 5.14)

where their best fit αr and βr values shown in Table 5.4. As seen in Figure 5.23, the

trend line reported by Leroueil et al. (1983) agrees well with the vane shear data.

The su-rem derived from the T-bar tests (see Figure 5.23) are consistently slightly higher

than those reported by Leroueil et al. (1983) and Boukpeti et al. (2012) at LI < 3.5

although the data points appear to converge with the viscometer data at LI > 3.5. Field

tests reported by Yafrate et al. (2009) and Low et al. (2010) showed similar higher

T-bar su-rem (if a nominal NT-bar = 10.5 is used to convert the remoulded T-bar resistance,

qrem) values when compared to other testing methods (vane, fall cone, etc). Zhou and

Randolph (2009a) conducted numerical simulation of T-bar cyclic test and found that

the width of the failure mechanism evolves during cyclic T-bar penetration and

extraction. Consequently, they postulated that the remoulded bearing factor, Nrem,

should be higher than the nominal T-bar bearing factor, NT-bar = 10.5.

To explore this possible explanation, the remoulded T-bar resistance from all the 81

tests were correlated to the su-rem derived for the same voids ratio from both the

viscometer (LI > 3.0) and vane data (LI < 3.0) obtained from Boukpeti et al. (2012).

The reference su-rem associated with the vane and viscometer data were calculated based

on two trend lines (Equation 5.14) fitted to the vane and viscometer data using separate

αr and βr values (Table 5.4), respectively. The calculated su-rem using these separate

relationships were then used to derive the remoulded bearing factors associated with the

vane (Nrem-vane) and viscometer (Nrem-visco), which are plotted in Figure 5.24. Generally

the Nrem-vane data falls within the documented field Nrem-vane range of 11.38 – 16.90 (Low

et al. 2010). The Nrem-visco ranges from 9.70-16.12. This validates the assumption of a

Nrem being slightly higher than NT-bar.

5.23

CORRELATIONS FOR STRENGTH DEGRADATION 5.8

As shown in Equation 5.1, the exponential strength degradation model is restrictive in

the sense that the soil ductility, N95, and 1/St (St – soil apparent sensitivity) parameters

must be fitted to a full sequence of cyclic T-bar (or ball) test, i.e. when cycles are

conducted until the fully remoulded state is reached. Yafrate et al. (2009) showed that

the strength degradation can be predicted by correlating the N95 and 1/St parameters

with the first sequence of T-bar extraction and penetration resistance ratio. Previously,

no attempts has been made to link the N95 parameter with any geotechnical properties.

The N95-LI relationship established in the previous section may be used as a basis to

predict the strength degradation curve.

As already seen from the centrifuge T-bar tests on the UWA kaolin, the operative

strength ratio (su/σ'v)nc of 0.17 appears to be smaller than that of the ‘true’ value inferred

from simple shear test. This implies that for the reconstituted UWA kaolin, the

penetration resistance during the first penetration, qin, is affected by strain softening.

This can be compensated by dividing the qin with a bearing factor, Nint, which is less

than the ideal rigid plastic factor, NT-bar of 10.5 to obtain the intact strength, su-int.

However, for some types of clays, viscous effect may dominate, leading to higher Nint

(Zhou and Randolph 2007; Zhou and Randolph 2009b; Low et al. 2010; DeJong et al.

2011). Apart from that, the Nrem magnitude may be higher than NT-bar due to the slightly

different T-bar failure mechanism width in remoulded soils (Zhou and Randolph

2009a), leading to lower magnitudes of the ‘true’ remoulded strength and therefore

higher ‘true’ sensitivity, St-r (higher than the apparent T-bar sensitivity, St) as proved to

be the case also for the UWA kaolin (Figure 5.24). It is for this reason that Equation 5.1

should be recast as:

953(n 0.25)/Nu n

u int t r t r

s 1 11 e

s S S

5.16

Based on comprehensive laboratory and field data (Skempton and Northey 1953;

Bjerrum 1954; Bjerrum and Simons 1960; Bjerrum 1967), the St-r (derived from tests

5.24

other than penetrometers) for many natural clays increases exponentially with liquidity

index (LI), following the relationship suggested by Wood (1990):

kLI

t rS e 5.17

where a factor k of 2 is recommended, although there is a tendency of the data to plot

within the upper and lower limits represented by k = 3 and k = 1 respectively (see

Figure 5.25). Additional St-r vs. LI plots obtained from tests on Louiseville (Leroueil et

al. 2003), Gloucester (Lo et al. 1976), OnsØy (Lunne et al. 2003) , Bothkennar (Nash et

al. 1992) and Burswood (Low et al. 2011) clays are included in Figure 5.25. These St-r

are either calculated as the ratio of the intact to remoulded strength from the field vane

test or as the ratio of the vane intact to remoulded strength deduced from the fall cone

test. It can be seen that, most of these additional data plot within the boundary of k = 1

and 3, where, except for the Louiseville data, the Gloucester, Bothkennar, Burswood

and OnsØy data plot pleasingly close to the k = 2 trend line. Since it is very likely that

the sensitivity of reconstituted clays (without any additions of chemicals that may

induce particle bonding, particle reorientation, etc.) does not vary with water content

(see Figure 5.21(a)), Equation 5.17 should perhaps be applied for natural clays only. On

the other hand, natural clays often have higher peak strengths due to fabric or particle

bonding elements, which are destroyed during the first penetrometer pass or the first

vane shear rotation, resulting in higher sensitivity than reconstituted clays.

Equation 5.17 may be used to estimate the intact shear strength, su-int provided that the

remoulded strength, su-rem is known (Wood 1990):

kLI

u int u rems e s 5.18

Besides using Equation 5.15, Wroth and Wood (1978) proposed that the remoulded

strength for clays with LI ≤ 1, su-rem may be estimated as:

4.6LI

u rems 170e 5.19

5.25

However, with the availability of disturbed samples, it may be more reliable to estimate

su-rem from the fall cone of miniature vane tests.

It follows that by combining Equations 5.13, 5.17 and 5.18 into Equation 5.16, the

decrease in the current strength, su-n from su-int can be estimated based solely on simple

index tests and determination of the remoulded strength, su-rem of disturbed samples:

d d3(n 0.25)/( )kLI LI

u n u rem u rems s s (e 1)e

5.20

The response of the degradation curve predicted by the above relationship is highly

dependent on the accuracy of determining su-rem and LI and the choice of the k

parameter (1 – 3).

PERFORMANCE OF CORRELATIONS 5.9

Figure 5.26(a) and Figure 5.27(a) depict the tests data at depths of 13 – 14 m and 3.85 –

4.7 m, respectively, obtained from previous T-bar cyclic tests in Burswood clay (NGI-

COFS 2006). Since the field test results were recorded in terms of raw T-bar resistance,

the equation that represents the degradation in resistance can be written as:

953(n 0.25)/N

n in

t t

1 1q q 1 e

S S

5.21

The first T-bar penetration resistance, qin is innately related to the intact strength, su-int

and can be converted to the latter using a suitable bearing factor, Nint (DeJong et al.

2011):

int 3

t r

6.5N 12

S1

10

5.22

5.26

where the ‘true’ sensitivity, St-r can be estimated using Equation 5.17. Because the

apparent T-bar sensitivity, St may not be equal to St-r, Equation 5.23 (Yafrate et al.

2009) can be used to estimate the former:

0.714 0.714kLI

t t rS S e 5.23

Using Nint, qin can now be estimated from the remoulded strength (Equation 5.18):

kLI

in u rem intq s e N 5.24

Substituting Equations 5.13, 5.23 and 5.24 into Equation 5.21, the T-bar current cyclic

resistance can now be estimated as:

d d3(n 0.25)/( LI )kLI

n u rem int 0.714kLI 0.714kLI

1 1q s e N 1 e

e e

5.25

The predicted qn (Equation 5.25 ) based on su-rem and LI quoted by Low et al. (2011),

is compared to the measured cyclic T-bar resistance in Figure 5.26(a) and Figure

5.27(a). It should be noted that the su-rem values used in these estimations are based on

the reported fall cone tests. The sensitivity of the predicted qn to the k parameter (1, 2

and 3) is demonstrated. It can be seen that by incorporating the proposed k value of 2

(Wood 1990) in Equation 5.25 , resulted in good agreement between the predicted and

measured T-bar resistance degradation for both cases.

The predicted strength degradations using Equation 5.20 for the T-bar cyclic tests at

both cyclic test depths are shown in Figure 5.26(b) and Figure 5.27(b). Here, the

measured in-situ intact and remoulded vane strengths are plotted for comparison. The

agreement between the su-rem inferred from the vane and fall cone tests for these two

tests is evident resulting in a reasonably good predictions of the intact strength as

compared to the intact vane strength.

5.27

CONCLUSIONS 5.10

This paper presents a centrifuge testing programme to study the evolution of the

geotechnical properties of consolidating kaolin samples. A total of 81 cyclic T-bar tests

were performed in 13 samples of UWA kaolin with undrained shear strengths that spans

from 0.08 to 1.7 kPa at 50 mm depth below the soil surface. T-bar cyclic tests were

performed to assess the strength degradation behaviour of consolidating ultra-soft kaolin

samples.

To obtain reliable measurements of the shear strengths of ultra-soft kaolin samples, a

simple method that utilises the current strength from T-bar cyclic tests were used to

correct the data. The corrected shear strengths agree well with those derived from large-

strain consolidation analyses. The back-calculated undrained strength ratio is close to

those derived from other centrifuge T-bar tests on the UWA kaolin but, lower than

typically obtained from element tests because of strain-softening.

From cyclic T-bar tests, the strength degradation parameters, namely the soil ductility,

sensitivity and remoulded shear strength were determined. The soil sensitivity is found

to be independent of the decrease in water content. The ductility and remoulded strength

correlate well with the normalised water content in the form of the liquidity index. This

paper also outlines the discrepancies between the remoulded strength inferred from a T-

bar test with those obtained from the vane shear, viscometer and fall-cone methods.

It follows that the soil strength degradation response can be linked to the index

properties of a soil sample by including the ductility-liquidity index and remoulded

strength-liquidity index relationships in the degradation model originally proposed by

Einav and Randolph (2005). This type of correlation is explored using published data

from other soils, with encouraging results.

The resulting correlations, and the consequent evolution of strength and strength

degradation properties, provide an improved basis for simulating this behaviour. These

correlations have potential applications to the simulation of submarine slide runout, and

soil-pipeline interactions that involve intense remoulding.

5.28






the CSIRO Wealth from Oceans Flagship Cluster on Subsea Pipelines. The authors are

also grateful to Bart Thompson, Phil Hortin, Shane De Catania and Tuarn Brown for

their technical assistance. Professor Martin Fahey’s assistance with MinTaCo is

gratefully acknowledged. The first author is also grateful for the financial support

received from the Ministry of Higher Education Malaysia (MOHE) and Universiti

Malaysia Sarawak (UNIMAS).

5.29

Table 5.1. Geotechnical properties of UWA kaolin

Property

Value Reference

Liquid Limit 58.40% Boukpeti et al. (2012)

Plastic Limit 28%

aa1 3.396

White et al. (2008, unpublished); Wu (2008) ab1 -0.238

ba2 0.000076

Richardson (2007, unpublished) bb2 3.468

Normally consolidated undrained

strength ratio, (su/σv)nc

0.25 Lehane et al. (2009)

a ev-σvo relationship (Equation 5.3)

b kv-ev relationship (Equation 5.4)

5.30

Table 5.2. Details of T-bar test programme

Sample

name

Test

name

Time after

start of

consolidation

(min)

Sample

U%

Initial

penetration

depth, z/DT-bar

Cyclic

depth,

z/DT-bar

Cycle

number

Sample 1 T1-1 182 24.39 9.5 4-7 9.75

Sample 1 T1-2 353 40.2 9.5 4-7 9.75

Sample 1 T1-3 478 49.88 9.5 4-7 4.75

Sample 1 T1-4 1268 91.03 9.5 4-7 9.75

Sample 1 T1-5 1333 92.31 9.5 4-7 9.75

Sample 1 T1-6 1473 94.3 9.5 4-7 9.75

Sample 1 T1-7 1762 95.62 9.5 4-7 9.75

Sample 1 T1-8 2003 97.03 9.5 4-7 9.75

Sample 1 T1-9 2720 98.82 9.5 4-7 9.75

Sample 1 T1-10 3112 98.99 9.5 4-7 9.75

Sample 1 T1-11 3463 99.08 9.5 4-7 9.75

Sample 1 T1-12 4234 99.08 9.5 4-7 9.75

Sample 1 T1-13 4902 99.96 9.5 4-7 9.75

Sample 1 T1-14 6292 99.99 9.5 4-7 9.75

Sample 1 T1-15 10116 100 9.5 4-7 9.75

Sample 1 T1-16 12732 100 9.5 4-7 9.75

Sample 3 T3-1 482 50.09 9.5 4-7 9.75

Sample 3 T3-2 1515 94.49 8.3 4-7 9.75

Sample 3 T3-3 1828 96.01 8.5 4-7 9.75

Sample 3 T3-4 2833 98.87 8.5 4-7 9.75

Sample 4 T4-1 802 66.76 10.5 4-7 9.75

Sample 4 T4-2 1162 85.51 9.5 4-7 9.75

Sample 4 T4-3 1418 93.52 9.1 4-7 9.75

Sample 5 T5-1 326 37.70 9.5 4-7 9.75

Sample 5 T5-2 382 42.44 9.5 4-7 9.75

Sample 5 T5-3 407 44.38 9.9 4-7 9.75

Sample 6 T6-1 209 26.88 9.5 4-7 7.75

Sample 6 T6-2 275 32.99 9.5 4-7 7.75

Sample 6 T6-3 492 50.61 11.8 4-7 9.75

Sample 6 T6-4 1171 85.98 9.1 4-7 9.75

Sample 6 T6-5 1259 90.56 8.9 4-7 9.75

Sample 7 T7-1 317 36.87 9.5 4-7 9.75

Sample 7 T7-2 378 42.13 12.5 4-7 9.75

Sample 7 T7-3 928 73.32 10.3 4-7 9.75

Sample 7 T7-4 943 74.10 9.9 4-7 9.75

Sample 7 T7-5 1055 79.94 9.7 4-7 9.75

Sample 7 T7-6 1684 95.26 8.6 4-7 9.75

Sample 7 T7-7 1699 95.33 8.6 4-7 9.75

5.31

Table 5.2. Details of T-bar test programme (continued)

Sample

name

Test

name

Time after

start of

consolidation

(min)

Sample

U%

Initial

penetration

depth, z/DT-bar

Cyclic

depth,

z/DT-bar

Cycle

number

Sample 8 T8-1 355 40.35 10.2 4-7 9.75

Sample 8 T8-2 370 41.51 12.5 4-7 7.75

Sample 8 T8-3 880 70.82 9.7 4-7 7.75

Sample 8 T8-4 895 71.60 9.9 4-7 7.75

Sample 8 T8-5 1102 82.39 9.3 4-7 9.75

Sample 8 T8-6 1116 83.11 9.1 4-7 6.75

Sample 8 T8-7 1126 83.64 9.4 4-7 7.75

Sample 8 T8-8 1608 94.91 8.3 4-7 7.75

Sample 8 T8-9 1811 95.91 8.3 4-7 9.75

Sample 8 T8-10 1822 95.97 8.5 4-7 9.75

Sample 9 T9-1 332 38.25 12.3 4-7 9.75

Sample 9 T9-2 376 41.98 12.3 4-7 9.75

Sample 9 T9-3 390 43.06 12.3 4-7 7.75

Sample 9 T9-4 989 76.50 9.5 4-7 7.75

Sample 9 T9-5 999 77.02 9.6 4-7 9.75

Sample 9 T9-6 1083 81.40 9.3 4-7 9.75

Sample 9 T9-7 1373 92.88 8.9 4-7 9.75

Sample 9 T9-8 1385 93.05 8.7 4-7 9.75

Sample 9 T9-9 1599 94.87 8.4 4-7 9.75

Sample 9 T9-10 1770 95.67 8.5 4-7 9.75

Sample 9 T9-11 2744 98.83 8 4-7 9.75

Sample 10 T10-1 315 36.68 13 4-7 6.75

Sample 10 T10-2 382 42.44 12.2 4-7 8.75

Sample 10 T10-3 541 53.16 11.7 4-7 7.75

Sample 10 T10-4 557 54.00 11.7 4-7 8.75

Sample 10 T10-5 1206 87.80 9.1 4-7 8.75

Sample 10 T10-6 1216 88.32 9.1 4-7 7.75

Sample 10 T10-7 1316 91.98 9 4-7 7.75

Sample 10 T10-8 2684 98.73 8.1 4-7 9.75

Sample 11 T11-1 376 41.98 12.6 4-7 6.75

Sample 11 T11-2 387 42.83 12.5 4-7 6.75

Sample 11 T11-3 1204 87.70 9 4-7 9.75

Sample 11 T11-4 1214 88.22 9.2 4-7 8.75

Sample 11 T11-5 1574 94.76 8.5 4-7 9.75

Sample 11 T11-6 1587 94.82 8.5 4-7 7.75

Sample 11 T11-7 1597 94.86 8.5 4-7 7.75

Sample 12 T12-1 258 31.41 13.1 4-7 6.75

Sample 12 T12-2 386 42.75 12.2 4-7 5.75

Sample 12 T12-3 534 52.80 11.5 4-7 6.75

Sample 12 T12-4 553 53.79 11.3 4-7 6.75

Sample 12 T12-5 1264 90.82 9.1 4-7 9.75

Sample 12 T12-6 1592 94.84 8.8 4-7 9.75

Sample 12 T12-7 1605 94.90 8.8 4-7 9.75

5.32

Table 5.3. Parameters for MinTaCo simulation cases: (a) Drum centrifuge consolidation; (b) Beam

centrifuge consolidation

Simulation

Name

Compressibility

parameters

Permeability

parameters Initial

water

content,

wi (%)

Initial

voids

ratio,

eo

Drainage

condition a1 b1 a2 b2

Case 1_drum 3.396a -0.238

a 0.000076

b 3.468

b 155 4 One-way

Case 2_drum 3.396a -0.238

a 0.00015

c 4.0

c 155 4 One-way

(a)

Simulation

Name

Compressibility

parameters

Permeability

parameters

Initial

water

content,

wi (%)

Initial

voids

ratio,

eo

Drainage

condition

a1 b1 a2 b2

Case 1_beam 3.396a -0.238

a 0.00015

c 4.0

c 120 3.12 Two-way

Case 2_beam 3.396a -0.238

a 0.000076

b 3.468

b 120 3.12 Two-way

(b)

a Calculated from moisture content cores from drum centrifuge test (Wu 2008) and beam centrifuge test

(White et al. 2008, unpublished) b Based on Rowe cell test reported by Richardson (2007, unpublished)

c Back-analysis using MinTaCo

Table 5.4. Best fit parameters for strength degradation curve, N95 and su-rem determined through

different testing methods

Coefficient

T-bar data

(present

study)

Combined vane

shear & viscometer

data (Boukpeti et al.

2012)

Vane shear

data (Boukpeti

et al. 2012)

Viscometer data

(Boukpeti et al.

2012)

αd – Eq. 5.13 -0.66 - - -

βd – Eq. 5.13 3.49 - - -

αr (kPa) – Eq. 5.14 3.235 1.71 1.7 1.168

βr – Eq. 5.14 3.055 2.64 2.532 2.41

5.33

Sample surface

z

T-bar

Penetration

Extractionn = 0.5

n = 0 n = 1

n = 0.25

n = 0.75

Figure 5.1. Cycle number notations (Randolph et al. 2007)

Figure 5.2. ev-σ′vo relationship of UWA kaolin

0.0

1.0

2.0

3.0

4.0

0.1 1 10 100 1000

Vo

ids

rati

o, e

v

Vertical effective stress, σ'v

Eq. 5.3 (a1 = 3.396, b1 = -0.238)

White et al. (2008, unpublished)

Wu (2008)

5.34

Figure 5.3. kv-ev relationship of UWA kaolin

0

0.0002

0.0004

0.0006

0.5 1 1.5 2

Per

mea

bil

ity, k

v(m

/da

y)

Voids ratio, ev

Richardson (2007, unpublished)

Eq. 5.4 (a2 = 0.000076, b2 = 3.468)

MinTaCo back-analysis: Eq. 5.4

(a2 = 0.00015, b2 = 4.0)

5.35

Figure 5.4. Cross sectional test setup in the UWA drum centrifuge channel

Figure 5.5. Scale used to monitor sample filling and settlement

5

0 mm10 20

30 40 50

60 70 80 90

100 110 120

130 140 150

160 170

180 190

200 mm

300 mm

Initial height of kaolin

slurry = 184 mm

Drum centrifuge channel

10 × 40 mm T-bar

Water level maintained

at 200 mm

200 mm scale for measuring

sample settlement (Figure 5.5)

Base PPT

Cyclic T-bar test

On-board camera to

monitor scale

Lower channelUpper channel

Ng

184 mm mark

Lower

channel

5.36

(a)

(b)

Figure 5.6. Stress error within a consolidating sample: (a) no settlement (initial condition); (b)

settlement of 50% of the initial sample height

0.2 m

ht-i = 0.184 m

1/3ht-i

Re = Rt-i + 1/3ht-i

Rt-i

z

Sample

Drum channel

Centre of rotation

0

40

80

120

160

200

-3 -2 -1 0 1 2 3 4 5 6 7

Sa

mp

le d

epth

, z

(mm

)

σv-m - σv-p (kPa)

ht = 0.184 m

N = 40 g

Zero

error0.6 m

0.2 m

ht = 0.092 m

Re = Rt-i + 1/3ht-i

Rt

z

Sample

Drum channel

Centre of rotation

0

20

40

60

80

100

-1 0 1 2 3 4 5 6 7 8 9

Sam

ple

dep

th, z

(mm

)σv-m - σv-p (kPa)

N = 40 g

ht = 0.092 m

Zero

error

0.6 m

5.37

Figure 5.7. Determination of the optimum effective radius

Figure 5.8. Average degree of consolidation, U%, of centrifuge test samples (Note: U% is calculated

based on each sample settlement (Eq. 5.11)

-15

-10

-5

0

5

10

15

0 0.1 0.2 0.3 0.4 0.5 0.6

Rel

ati

ve

stre

ss e

rro

r, E

r(%

)

Consolidation strain, ∆ht/ht-i

Re = Rt-i + 1/3ht-i

Re = Rt-i + 1/2ht-i

Re = Rt-i + 2/3ht-i

0

20

40

60

80

100

120

1 10 100 1000 10000 100000

Deg

ree

of

co

nso

lid

ati

on

, U%

Time (mins)

Sample 1 Sample EET

Sample 2 Sample 3

Sample 4 Sample 5

Sample 6 Sample 7

Sample 8 Sample 9

Sample 10 Sample 11

Sample 12

5.38

Figure 5.9. Cyclic penetration and extraction data for T-bar test T4-1 (U ~ 67%)

(a)

(b)

Figure 5.10. Cyclic strength degradation derived from T-bar test T4-1 (U ~ 67%)

0

20

40

60

80

100

120

140

-1 -0.5 0 0.5 1D

ep

th, z(m

m)


After correction

Before correction

Zero line

N = 40 g

0

1

2

3

4

5

0 2 4 6 8 10 12

Deg

ra

da

tion

fa

cto

r, s

u-n

/ s u

-in

Cycle number

Before correction

0

0.2

0.4

0.6

0.8

1

0 2 4 6 8 10 12

Deg

ra

da

tion

fa

cto

r, s

u-n

/ s u

-in

Cycle number

After correction

5.39

Figure 5.11. Pressure components acting on a T-bar

qw

qsoil

qb

Crossbar

qL qL

Axial gaugeShaft

5.40

Figure 5.12. Factors affecting T-bar resistance in underconsolidated samples

Figure 5.13. Resultant T-bar resistance (qT-bar) in underconsolidated samples

qsoil

qT-bar

Dep

th, z

qb

qw

qL

Axial gauge

level

CompressiveTensile

Extraction

qT-bar

Dep

th, z

qsoil + qb + qw + qL

Simple linear adjustment

Penetration

Dominated by

buoyancy

Dominated by T-

bar self-weight and

pressure on axial

gauge

5.41

Figure 5.14. Comparison of the measured and predicted sample height (drum centrifuge data)

Figure 5.15. Comparison of the measured and predicted sample height (beam centrifuge data)

80

100

120

140

160

180

200

0.1 1 10 100 1000 10000 100000

Sa

mp

le h

eig

ht (m

m)

Time (mins)

Case 1_drum Case 2_drum

Sample 1 EET

Sample 2 Sample 3

Sample 4 Sample 5

Sample 6 Sample 7

Sample 8 Sample 9

Sample 10 Sample11

Sample 12

Optimised k-e parameters

k-e parameters from Rowe cell

80

100

120

140

160

180

200

220

1 10 100 1000 10000

Sa

mp

le h

eig

ht (m

m)

Time (mins)

Case 1_beam

Case 2_beam

White et al. (2008, unpublished)

Optimised k-e parameters (for drum

centrifuge tests)


kv-ev parameters

from Rowe cell

Optimised kv-ev

parameters

kv-ev parameters

from Rowe cell

Optimised kv-ev

parameters (from

current drum

centrifuge tests)

5.42

Figure 5.16. Evolution of base excess pore pressure

Figure 5.17. Comparison of the fully consolidated strengths (Sample 1) from T-bar tests (T1-15 and

T1-16) with the predicted strengths obtained from Case 2_drum simulation

0

5

10

15

20

25

0.1 1 10 100 1000 10000 100000

Ba

se e

xce

ss p

ore

pre

ssu

re (

kP

a)

Time (mins)

Case 1_drum

Case 2_drum

Sample 4

Sample 5

Sample 6


Optimised k-e parameters

0

20

40

60

80

100

0 1 2 3 4 5

Pen

etra

tio

n d

epth

, z (

mm

)

Initial undrained shear strength, su-in (kPa)

T1-15 (t = 10116 mins)

T1-16 (t = 12732 mins)

(su/σ'v)nc = 0.17(su/σ'v)nc = 0.25

Fully consolidated strength (Case 2_drum)

kv-ev parameters

from Rowe cell

Optimised kv-ev

parameters

5.43

Figure 5.18. Increase of undrained shear strength at 50 mm below the soil surface

Figure 5.19. Decrease of voids ratio at 50 mm below soil surface

0.01

0.1

1

10

100 1000 10000 100000

Un

dra

ined

sh

ear

stre

ng

th (k

Pa

)

Time (mins)

(su/σ'v)nc = 0.17

su-in (T-bar)

Case2_drum

1

2

3

4

5

100 1000 10000 100000

Vo

id r

ati

o, e v

Time (mins)

Eq. 5.12

Case 2_drum

Vo

ids

rati

o, e v

5.44

(a)

(b)

Figure 5.20. Cyclic degradation of soil sample: (a) U = 51%; (b) U = 100%

0

0.4

0.8

1.2

0.1 1 10

Deg

rad

ati

on

fa

cto

r, s

u-n

/su

-in

Cycle number

Fit to data: N95 = 1.02, 1/St = 0.45

Test T6-3 (U = 51%)

0

0.4

0.8

1.2

0.1 1 10

Deg

rad

ati

on

fa

cto

r, s

u-n

/su

-in

Cycle number

Fit to data: N95 = 2.82, 1/St = 0.43

Test T1-16 (U = 100%)

5.45

(a)

(b)

Figure 5.21. Strength degradation parameters: (a) soil sensitivity, St; (b) soil ductility, N95

0

1

2

3

0 20 40 60 80 100 120

Sen

siti

vit

y, S

t

Average degree of consolidation, U%

0

1

2

3

4

0 20 40 60 80 100 120

Du

ctil

ity, N

95

Average degree of consolidation, U%

5.46

Figure 5.22. Relationship of N95 from T-bar tests and liquidity index (LI)

Figure 5.23. Interrelationship between su-rem and LI

0

1

2

3

4

0 1 2 3 4 5

Du

ctil

ity,

N9

5

Liquidity index, LI

Kaolin data

Louiseville

Gloucester

Bothkennar (13 - 14 m)

West African clay

Onsoy

Burswood (4.3 m)Eq. 5.13

OnsØy

0.01

0.1

1

10

0 1 2 3 4 5 6 7 8

Rem

ou

lded

sh

ear

stre

ng

th, s

u-r

em

Liquidity index, LI

T-bar data

T-bar data fit (Eq. 5.14, Table 5.4)

Vane data (Boukpeti et al. 2012)

Vane fit (Eq. 5.14, Table 5.4)

Viscometer data (Boukpeti et al. 2012)

Viscometer data fit (Eq. 5.14, Table 5.4)

Vane + viscometer data fit (Boukpeti et al. 2012) - Eq. 5.14, Table 5.4

Leroueil et al. (1983) - Eq. 5.15

5.47

Figure 5.24. T-bar bearing factors (Nrem) in remoulded soil

Figure 5.25. Relationship of sensitivity with liquidity index (LI) for natural clays

8

10

12

14

16

18

0 1 2 3 4 5

Nrem

Liquidity index, LI

Range of Nrem-vane

(Low et al. 2010)

Nrem-vane

Nrem-visco

1

10

100

0 0.5 1 1.5 2 2.5 3 3.5 4

Sen

siti

vit

y, S

t-r

Liquidity index, LI

Best fit (Wood 1990) - k = 2

Lower bound (Wood 1990) - k = 1

Upper bound (Wood 1990) - k = 3

Louiseville (Fall cone)

Gloucester (Vane)

Onsoy (5 - 24 m - Vane)

Bothkennar (3 - 4 m Vane)

Burswood (3 - 14m Vane)

Burswood (3 - 14 m Fall cone)

OnsØy (5 - 24 m Vane)

5.48

(a)

(b)

Figure 5.26. (a) Comparison of measured (test at 13 – 14 m) and predicted T-bar resistance

degradation for Burswood clay; (b) predicted strength degradation

(a)

(b)

Figure 5.27. (a) Comparison of measured (test at 3.85 – 4.7 m) and predicted T-bar resistance

degradation for Burswood clay; (b) predicted strength degradation

0

200

400

600

800

0 2 4 6 8 10 12

Pen

etr

ati

on

resi

sta

nce

, q

n(k

Pa

)

Cycle number

k = 1

k = 2

k = 3

T-bar (13 - 14 m)

0

20

40

60

80

100

120

0 2 4 6 8 10 12S

hea

r s

tren

gth

, s u

-n(k

Pa)

Cycle number

k = 1

k = 2

k = 3

Vane (intact)

Vane (remoulded)

0

50

100

150

200

0 1 2 3 4 5 6

Pen

etr

ati

on

resi

sta

nce

, q

n(k

Pa

)

Cycle number

k = 1

k = 2

k = 3

T-bar (3.85 - 4.7 m)

0

10

20

30

40

50

0 1 2 3 4 5 6

Sh

ea

r s

tren

gth

, s u

-n(k

Pa)

Cycle number

k = 1

k = 2

k = 3

Vane (intact)

Vane (remoulded)

6.1

CHAPTER 6. CENTRIFUGE MODELLING OF ACTIVE

SLIDE-PIPELINE LOADING IN SOFT CLAY

FOREWORD

The effect of strain rate (Objective 1 – see Section 1.2.1) are explored further in this

chapter. This chapter is a continuation from Chapter 3, where it extends the viscous

regime previously considered, to encompass inertial effects by laterally translating a

model pipe at high velocities in consolidating samples within the geotechnical

centrifuge. The rate of straining in the experiments documented here was augmented by

up to 33 fold compared to the strain rates in the VOP experiments in Chapter 3. The

range of soil strengths considered in this chapter is also much wider than that in

Chapter 3, using the strength models developed in Chapter 5. The wide range of

strengths and strain rates considered are important to understand the behaviour of

submarine slide-pipeline interaction during the later stages of a slide, where the

combination of a very high velocity and very low strength warrants the consideration of

inertial effects. The chapter ends with a capstone outcome of this thesis: a general

model for impact forces on pipelines across the solid-fluid boundary, verified by

experiments.

6.2

ABSTRACT 6.1

Submarine slides pose a serious threat to the viability of pipelines in the proximity of

the continental slope. This paper describes the results of a centrifuge testing programme

aimed at studying the impact forces exerted by a submarine slide on an offshore

pipeline. This was achieved by dragging a model pipe at varying velocities, in fine-

grained soil at various degrees of consolidation, hence exhibiting properties spanning

from the fluid to the geotechnical domains. To simulate the high strain rates experienced

by the soil during a submarine slide, tests were conducted at relative pipe-soil velocities

of up to 4.2 m/s. The changing density and shear strength of the samples were back

calculated from T-bar penetrometer test results. A hybrid approach combining

geotechnical and fluid mechanics-based components of horizontal drag resistance was

developed. This approach links the density and strength of the samples to the resistance

on the pipe. Besides fitting the present observations, the method provides an improved

reinterpretation of similar data from the literature. Data gathered from the two pore

pressure transducers on the model pipe indicated the existence of vortex shedding at

high non-Newtonian Reynolds number. It was further demonstrated that the hybrid

approach developed to estimate the horizontal drag force can also be used to estimate

the vertical lift force induced by vortex shedding.

INTRODUCTION 6.2

With the expansion of the oil and gas industry into deeper waters, too deep to build rigid

platforms founded on the seabed, there is now a greater reliance on subsea infrastructure

to extract hydrocarbon resources. Export pipelines are used to convey the resources to

shore. These export pipelines can extend to over 500 km and the viability of these

developments is influenced by the security of the pipeline against potential damage

from geohazards along the pipeline route. One of the most damaging forms of

geohazards is submarine slides. Compared to subaerial slides, submarine slides have

greater mobility with run-out distances of more than 100 km (Locat and Lee 2002) and

involve larger volumes of failed material. As a result, they pose serious threats to the

safety of nearby pipelines on the continental shelf.

6.3

Existing methods to quantify the slide forces exerted on a pipeline can be divided into

the geotechnical and fluid mechanics methods. At the onset of a submarine slope

failure, the failed mass travels downslope initially at low velocity (compared to the

more advanced stages of a slide) and possesses geotechnical properties close to the

intact parent (pre-failure) soil mass. As such, the slide horizontal drag pressure qH can

be estimated from the operative undrained shear strength of the soil su-op using a

conventional geotechnical bearing capacity factor NH as follows:

H H u opq N s 6.1

However, qH is a function of the slide velocity due to the effect of strain rate and thus

slide velocity on su-op (Zhu and Randolph 2011). To capture this effect, rather than

enhancing NH (Georgiadis 1991; Zakeri et al. 2011), a more straight forward approach

is to use a single NH factor – reflecting that bearing factors are essentially a function of

the problem geometry, not the soil properties – and to account for the enhanced qH

imposed by a slide at high velocities on a pipeline by adjusting su-op for strain rate via

(e.g. Biscontin and Pestana 2001; Boukpeti et al. 2012):

m

u op u ref

ref

γs s

γ

6.2

where γ is the shear strain rate, su-ref is the reference undrained shear strength at a

reference strain rate ref .γ Zhu and Randolph (2011) showed that γ may be taken as

equal to v/Dpipe provided that the shear thinning parameter m is less than 0.3. Boukpeti

et al. (2012) further showed that m is independent of the soil density.

As the failed mass travels further downslope, remoulding of the soil and interaction with

the surrounding water takes place. This causes a decrease in the shear strength (or

mobilised shear stress) of the slide material (now known as a debris flow) compared to

the original intact pre-failure slope. Debris flow can travel up to velocities of 7 to

30 m/s (Bjerrum 1971; Imran et al. 2001; Canals et al. 2004; De Blasio et al. 2004b).

Although a debris flow has low shear strength, the density of the slide material is

6.4

sufficiently high to cause damage to a pipeline installation located in the path of the

debris flow. Due to the reliance on the slide material operative shear strength, the

geotechnical approach (Equation 6.1) is inadequate on its own to estimate the slide

impact force on a pipeline when inertial drag forces – which arise from the density of

the flow, rather than its strength – are not negligible.

A common approach to assess the impact load from a debris flow is to start from a fluid

drag perspective and characterise the flow as a non-Newtonian fluid. The slide impact

force is linked to the slide material inertia (combined effects of slide density ρ and

velocity v) via a drag coefficient CD as shown below:

2

H D

1q C ρv

2 6.3

CD is a function of the non-Newtonian Reynolds number Renon-Newtonian which is defined

as:

2

non-Newtonian

vRe

6.4

where τ is the mobilised shear stress within the slide material. For a shear thinning

viscoplastic debris flow, τ varies non-linearly with γ and can be quantified by the

Herschel-Bulkley model (Locat and Demers 1988) in the form:

m

yτ τ ηγ 6.5

where τy is the yield stress, η is the viscosity and m is a shear thinning parameter

(similar to Equation 6.2).

The drag factor, CD, can then be represented by (Zakeri et al. 2008; Zakeri 2009; Zakeri

et al. 2009):

6.5

D 1.25

non Newtonian

17.5C 1.4

Re

6.6

The above equation was calibrated from experimental tests in the flume (Zakeri et al.

2008) coupled with Computational Fluid Dynamics (Zakeri et al. 2009) simulating

debris flow impact on a suspended pipeline.

The two methods outlined above fail to capture the full behaviour of the passive slide-

pipeline interaction over a wide range of relative pipe-soil velocity v and su-op. While the

geotechnical approach (Equation 6.1) neglects the inertial drag of debris flows of low

su-op travelling at high velocities, the fluid mechanics method links the drag force

directly to the material density, even if the resistance is principally due to the strength of

the material.

The ultimate aim of this paper is to unify both the geotechnical and fluid mechanics

principles (denoted as hybrid approach from hereon) in order to provide a method of

quantifying the forces of a submarine slide on a pipeline regardless of the parameters

(strain rate, density, strength) that control the magnitude of the slide force. To achieve

this, an extensive centrifuge testing programme was undertaken, which involved

horizontally translating a model pipe through soil samples with shear strengths and

densities that span from those relevant to the conventional geotechnical to fluid

mechanics domains.

Another aspect of slide impact loading that is not adequately recognised is the vertical

lift force induced at high Renon-Newtonian. This force component cannot be overlooked: it

has been measured as varying from 25 – 96% of the horizontal impact force (Zakeri et

al. 2008; Zakeri et al. 2009). However, no systematic method for quantifying the

vertical force was provided. As a secondary aim of this paper, the flume test data of

Zakeri et al. (2008) coupled with the on-pipe pore pressure measurements from the

present centrifuge study were used to improve the understanding of this often-neglected

impact load component.

6.6

EXPERIMENTAL APPARATUS 6.3

UWA drum centrifuge 6.3.1

The force exerted by a debris flow impact on a pipeline may be simulated by releasing a

volume of soil at high velocities, simulating a slide event (Zakeri et al. 2008; Boylan et

al. 2012). However, it is difficult to control the test parameters, especially the slide

velocity, and the volume of soil may also be limited. A more straightforward approach

is to move a model pipe at a fixed velocity in a soil sample (active loading). Other test

parameters such as the soil density and undrained shear strength can easily be varied by

conducting tests at different stages of consolidation.

The drum centrifuge at the University of Western Australia (UWA) (Figure 6.1) is an

ideal tool for pipe drag testing because of its large plan area where tests can be

conducted. At model scale, the channel containment area has a width of 300 mm

(measured vertically) and a depth of 200 mm (measured radially). The UWA drum

centrifuge has a full diameter (measured to the base of the channel) of 1.2 m and can

rotate at up to 850 rpm. The tool-table has actuators to enable movement in three

directions (vertical, radial and circumferential) and can be stopped independently from

the channel (maintaining the stress level in the sample). This allows the experimental

tools to be changed over during a test program. A complete description of the UWA

drum centrifuge, as established in 1998, is provided by Stewart et al. (1998).

The tool-table circumferential movement relative to the rotating channel is controlled by

the Dynaserv servo motor. The Dynaserv is capable of reaching a maximum relative

angular velocity and acceleration of 540°/s and 1800°/s2 respectively, which correspond

to a velocity of 4.71 m/s and an acceleration of 15.7 m/s2 at a radius of 0.5 m. However,

the original servo controller bottlenecked this capability causing the Dynaserv to

respond to the specified high acceleration and velocity inaccurately. In order to achieve

the high relative velocities and accelerations required for the test programme, a new

PCI-7344 motion controller card (manufactured by National Instruments®

) (Figure 6.2)

was installed. It is capable of handling a PID (proportional - integral - derivative) loop

update rate of 60 µs, hence achieving a better control of the tool-table motion. A new

6.7

LabView control interface was also developed (Figure 6.3) to allow the user to pre-

program specific testing variables such as the total displacement, test position, test

velocity and acceleration.

Following rigorous calibrations of the tool-table velocity-displacement response,

suitable P, I and D constants were experimentally optimised as 1, 5 and 150,

respectively, at angular velocities between 10°/s to 540°/s and 5, 3, 100 for test

velocities below 10°/s.

T-bar penetrometer 6.3.2

A T-bar penetrometer (Figure 6.4) was used to provide continuous shear strength

profiles within the consolidating samples. To provide better resolution of the soil

resistance in samples at low degrees of consolidation, a larger T-bar measuring 10 × 40

mm (diameter DT-bar × length LT-bar) was used instead of the usual standard size of 5 ×

20 mm. The axial strain gauge located immediately above the T-bar records the vertical

penetration resistance qT-bar, which can be converted to the undrained shear strength su

as follows:

T baru

T bar

qs

N

6.7

where NT-bar is the bearing capacity factor, with a value of 10.5 adopted here as

suggested by Stewart and Randolph (1991), reflecting an intermediate surface

roughness consistent with lower and upper bound plasticity solutions (Randolph and

Houlsby 1984; Martin and Randolph 2006).

Model pipeline 6.3.3

A model pipeline, with a diameter Dpipe of 20 mm, was manufactured for the test

programme (Figure 6.5). The model pipe features three pairs of extension pieces of

different lengths (Lpipe) which can be fitted to the threaded sections at both model pipe

ends. This permits testing with model pipe aspect ratios Lpipe/Dpipe of 6, 8 and 10.

6.8

For testing, the model pipe is connected to a loading arm monitoring both vertical and

horizontal loads. Horizontal loads on the pipe are measured by electrical comparison of

the response from the two sets of bending strain gauges fixed at different positions on

the upper part of the loading arm. A set of axial strain gauges attached just above the

model pipe is used to measure the vertical resistance. During vertical penetration of the

model pipe into the soil, the whole assembly resembles a T-bar penetrometer, and as

such, the measured vertical penetration resistance qpipe can then be converted into su, by

using Equation 6.7 (substituting qT-bar = qpipe and using the same bearing factor of 10.5).

Two pore pressure transducers (PPTs) are installed on the mid-section of the pipe. One

is installed on the front face and the other on the opposite side of the model pipe (i.e. at

the 3 o’clock and 9 o’clock positions). These PPTs are used to detect pore pressure

differences associated with the soil flow from the front to the rear of the model pipe.

SAMPLE PREPARATION 6.4

The entire experimental programme included a total of 13 samples. To prepare each

sample, dry kaolin powder was mixed with water to form a slurry with a targeted

moisture content of 155%. Special care was taken to ensure consistency of the ratio of

water to dry kaolin for all 13 samples. Each kaolin slurry was mixed for at least 3 hours,

which is sufficient to ensure homogeneity. Moisture content measurements of the 13

samples taken after mixing indicated satisfactory sample consistencies with differences

within ± 1% of the targeted water content.

For each test series, the slurry was transferred to the drum centrifuge channel whilst it

was spinning at a speed equivalent to 40 g acceleration. Assuming sedimentation occurs

immediately, the initial voids ratio of the each sample, eo, was 4 (calculated based on a

specific gravity Gs = 2.6).

6.9

TEST PROCEDURE 6.5

Figure 6.6 depicts the test setup within the drum centrifuge channel. In the subsequent

sections, all test parameters and results are presented in model scale unless stated

otherwise. General information regarding the tests is as follows:

1. All centrifuge tests were performed at a centrifuge acceleration of 40 g and no

drainage layer was installed at the bottom of each sample (creating one-way

drainage to the soil surface). This was done to slow the sample consolidation so

that a number of tests (pipe or T-bar tests) can be completed with minimal

change in degree of consolidation U%, which is defined in terms of the sample

settlement:

t

u

hU% 100

h

6.8

where ∆ht is the settlement at a particular time t and ∆hu is the final settlement

when the sample is fully consolidated.

2. For each test, the water level was maintained at a height of 200 mm, which is the

full height of the drum channel. This was achieved by allowing a continuous

discharge of water by overflow from the top of the channel. This was maintained

to avoid the influence of evaporation, thus ensuring consistent consolidation

responses between all 13 samples.

3. The initial sample height was 184 mm and sample settlements were monitored

using a scale placed on the lower side of the channel. Pictures captured by an on-

board video camera were used to derive the changes in sample height with time.

The progress of consolidation during each test was also monitored using a pore

pressure transducer installed at the base of the drum channel.

4. To determine the shear strength su of the samples, T-bar tests were performed at

a penetration rate of 2 mm/s. Since a 10 mm diameter T-bar was used in all the

tests, the penetration velocity of 2 mm/s was chosen to match the strain rate of

0.2 s-1

of a conventional small-scale T-bar with a typical diameter of 5 mm and a

penetration rate of 1 mm/s. To avoid boundary effects on the T-bar results, a

6.10

minimum spacing of at least 20 mm between the T-bar edges and the centrifuge

channel was adopted, since the T-bar failure mechanism (or influence zone)

extends to about 2 diameters (Einav and Randolph 2005; Martin and Randolph

2006; Zhou and Randolph 2009a). All T-bar tests involved multiple penetration

and extraction cycles.

5. Prior to the drag forces being measured through horizontal translation at a

specified velocity, the model pipe was penetrated vertically into the soil at a rate

of 3 mm/s (strain rate of 0.15 s-1

) to a normalised embedment z/Dpipe of 2.5.

Here, z is the distance of the mid-section of the pipe to the soil surface and Dpipe

is the pipe diameter. This embedment depth was chosen to avoid boundary

effects arising from the base of the drum channel, which would be most

significant when the soil was fully consolidated. Ideally, the model pipe should

be penetrated at a rate of 4 mm/s in order to match the strain rate of the T-bar

tests (0.2 s-1

). This was not possible as the maximum radial velocity of the tool-

table actuator is limited to 3 mm/s. However, shear strength data derived from

the model pipe matches the su derived from the T-bar closely (Section 6.6.2),

highlighting the minor effect of these small differences in strain rate.

6. After pipe embedment, the model pipe was translated horizontally at a specified

velocity v for a normalised test distance u/Dpipe (where u is the total horizontal

displacement).

An overview of the test programme is presented in Figure 6.8. The programme involved

13 soil samples in which 81 T-bar tests and 37 pipe tests were performed. The pipe tests

were associated with a further 31 so-called shaft tests, where only the connecting shaft

was embedded at a depth equivalent to the shaft length exposed to the soil during pipe

tests (40 mm) and translated laterally. The purpose of these tests was to measure

independently the forces acting on the connecting shaft under the same conditions as the

pipe tests, so the forces acting on the pipe can be isolated.

It is acknowledged that the above method of accounting for the load component of the

shaft does not take into account the additional shearing component at the base of the

shaft. However, assuming a friction factor of 1 (fully rough shaft base), it is estimated

that this shear component only causes a negligible error of 0.1%.

6.11

The testing programme is divided into four sections related to (i) investigation of the

consolidation characteristics of the clay (Section 6.5.1), (ii) pipe tests performed in a

fully consolidated sample (Section 6.5.2), (iii) investigation of end effects on the forces

acting on the pipe (Section 6.5.3), and (iv) pipe tests performed in consolidating

samples (Section 6.5.4).

Investigation of kaolin consolidation behaviour - Sample 1 6.5.1

The test series in Sample 1 was performed to provide insight into the consolidation

behaviour of kaolin. A total of 16 T-bar cyclic tests were conducted in this test series.

The time of test, depth of initial penetration, number of penetration and extraction

cycles as well as cyclic depth for each T-bar test are shown in Table 6.1. The T-bar

cyclic numbering system is based on the recommendation of Randolph et al. (2007)

where the first penetration is taken as cycle number 0.25 and the subsequent extraction

as 0.75, representing the average number of cycles experienced by the soil within the

failure mechanism.

Sample height changes were monitored by video means as outlined in Section 6.5. The

undrained shear strength derived from the T-bar tests and sample settlement

measurements was used to plan the other pipe tests in samples with U < 100%.

(Sections 6.5.3 and 6.5.4).

Pipe tests in fully consolidated sample - Sample 1 6.5.2

Approximately 8 days after the start of consolidation (when full primary consolidation

had been achieved), four pipe tests and their respective shaft tests were performed. The

test details for the pipe and shaft tests are shown in Table 6.2. A 20 × 120 mm pipe was

used in this test series. In stiffer soil samples, this model pipe length to diameter ratio

(aspect ratio) Lpipe/Dpipe is considered sufficiently high to avoid end effects. This has

been confirmed from T-bar tests (which can be regarded to be similar to a model pipe)

conducted by Chung (2005) and Watson (1999) using T-bars with different aspect

ratios.

6.12

End effects tests (EET) - Sample EET 6.5.3

It is widely accepted that end effects (reflecting non-plane strain flow) are significant

for model pipe experiments in Newtonian fluids (Farell and Fedeniuk 1988; Park and

Lee 2000; Afgan et al. 2007). In order to ensure plane strain flow for tests using a model

pipe with a finite length in Newtonian fluids, researchers either install end plates

(e.g. Farell and Fedeniuk 1988) or resort to using model pipe aspect ratios of up to 45

(Gouda 1975).

For model pipe testing at high velocities in non-Newtonian fluids (such as a fluid-like

kaolin slurry), there are no published studies about the implications of using a model

pipe with a finite length. Therefore, this test series was specifically tailored to study the

effects of using different model pipe Lpipe/Dpipe ratios during high velocity testing in

kaolin slurry.

The end effects test series was conducted in Sample EET, with the testing details

provided in Table 6.3 All tests (labelled EET-x) were performed in a single sample

initially consolidated to U = 10%. For each test, the model pipe was first penetrated

vertically (at 3 mm/s) into the soil sample to an embedment ratio z/Dpipe of 2.5. This was

subsequently followed by successive cycles of forward and backward sweeps as shown

in Figure 6.9. The cyclic numbering system follows that of the T-bar tests (Section

6.5.1). Cycles 0.25 – 1.75, 2.25 – 3.75 and 4.25 – 5.75 were conducted at velocities of

4, 2.75 and 1.88 m/s respectively. The same test sequence was repeated for the different

Lpipe/Dpipe ratios of 6 (EET - 1,2,3), 8 (EET - 4,5,6), and 10 (EET - 7,8,9). This cyclic

regime was imposed to ensure sample consistency, thus ensuring that all model pipe

tests were conducted in soil exhibiting the same remoulding state (see Section 6.7).

Tests at the same velocities, displacements and sequence were repeated with just the

connecting shaft, to determine the contribution of the shaft to the total load. Although

the whole test sequence took 2 hours to complete, the sample degree of consolidation

U% (and therefore the soil properties) are considered to be unchanged owing to the

intense soil remoulding during cyclic pipe displacements, which disrupted the

consolidation progress. This assumption is justified as the sample height was observed

to be constant throughout the testing period.

6.13

The 20 × 120 mm pipe was chosen for further tests at U < 100% (Section 6.5.4). The

logic for choosing this pipe dimension is explained further in Section 6.7.

Tests in samples at U < 100% (Samples 2 –12) 6.5.4

This test series involved the remaining 11 samples. In total, 24 model pipe tests were

performed using a 20 × 120 mm pipe in samples at U ~ 10 – 95%. For each model pipe

test, a shaft test was also performed to assess the influence of the shaft on the total

force. The changing properties (density ρ and strength su) as the samples consolidated

were characterised by T-bar tests. Details of the T-bar, model pipe and shaft tests are

provided in Table 6.4 and Table 6.5. It should be noted that during the course of testing

in Sample 2, the T-bar penetrometer malfunctioned and therefore the changes in su as

consolidation progressed could not be quantified directly. However, the sample height

changes were similar to the pattern followed by the other tests as depicted in Figure

6.10. Therefore, the changes in su and ρ were regarded as the same.

RESULTS: CONSOLIDATION BEHAVIOUR OF KAOLIN 6.6

Sample settlement 6.6.1

An image processing software, AnalyzingDigitalImages (Pickle 2008) was used to

determine the changes in sample heights from the photos taken during testing. Figure

6.10 and Figure 6.11 show the evolution of sample heights and the corresponding

average degrees of consolidation U% for all 13 samples respectively. The U% is

calculated via Equation 6.8.

It can be seen that both the sample height-time and U%-time profiles fall within a

narrow band. This confirms that the changes in the sample properties as consolidation

progressed were similar for all the tests, and validates the sample preparation method.

6.14

Evolution of shear strength and density with consolidation time 6.6.2

The shear strengths extracted from all the initial T-bar penetrations su-in at a depth of z =

2.5Dpipe (i.e. 50 mm below the soil surface) are combined in Figure 6.12. At very low

U%, where the soil strengths are very small, these T-bar su-in values were corrected

using the cyclic tests as outlined in Sahdi (2012). For comparison, the values of su-in

inferred from the initial vertical penetrations of the model pipe tests are also plotted in

Figure 6.12. It should be noted that only values of su-in obtained from model pipe tests in

samples at U > 65% are included in Figure 6.12 owing to the inability of the axial strain

gauges on the model pipe to detect lower soil resistance accurately. This limitation is

due to the relatively large shaft area where the axial strain gauges are attached.

However, the su-in values inferred from both the T-bar and pipe tests generally follow

the same trend with minimal scatter.

In a separate study, Sahdi (2012) conducted large-strain consolidation analyses

simulating the consolidation process reported in this paper. Analyses were conducted

using a finite element programme, MinTaCo (Seneviratne et al. 1996) to estimate the

increase in shear strength with consolidation time. Their results are compared to those

measured from the T-bar and pipe tests in Figure 6.12. The predicted shear strength was

obtained via the following equation:

uu in ov

v nc

ss σ'

σ'

6.9

where σvo is the predicted vertical effective stress and (su/σv)nc is the normally

consolidated strength ratio. Good agreement was achieved when σvo were multiplied

with (su/σv)nc = 0.17. This strength ratio is consistent to other centrifuge T-bar tests on

UWA kaolin reported by Hodder et al. (2010) and Sahdi et al. (2010), but is somewhat

lower than normally inferred from element test (Lehane et al. 2009). This may be

attributed to strain softening as the T-bar or pipe is penetrated into the soil (Einav and

Randolph 2005; Zhou and Randolph 2009a; Zhou and Randolph 2009b). However, this

back calculated strength ratio is assumed to represent the operative strength ratio in the

6.15

vicinity of the pipe, as the pipe and T-bar are most likely to experience the same

magnitude of strain softening because of their similar geometry.

For convenience, the operative shear strength in the vicinity of the pipe at a particular

test U% is approximated by the following expression (R2 = 0.98) (Figure 6.13):

u in 2

1s

17.65 0.23U 0.00055U

6.10

which fits the measured results closely. The decrease in voids ratio, ev, as the soil

consolidates can be calculated using the following equation:

1

o

b

v 1 ve a σ' 6.11

For the UWA kaolin, the constants a1 and b1 are taken as 3.396 and -0.238 respectively

(Sahdi 2012). This power function is commonly employed in consolidation cases such

as this, involving samples with high voids ratios (Carrier et al. 1983; Toh 1992;

Seneviratne et al. 1996).

It is possible to estimate the voids ratio (ev) directly from su-in (Equation 6.10) by

substituting Equation 6.9 into Equation 6.11, which yields:

1b

u1 u in

v nc

se a s

σ'

6.12

The soil density ρ is then:

s v w

v

G e ρρ

1 e

6.13

where ρw is the density of water (taken as 1000 kg/m3).

6.16

These relationships provide values of both strength and density, to be used in the

proposed hybrid method of estimating the slide impact forces developed from back-

analysing the measured pipe-soil resistance.

RESULTS: END EFFECTS TESTS (EET) 6.7

Figure 6.14 presents the results of the EET tests in terms of the horizontal pressure on

the pipe qH, calculated as the lateral force on the pipe H divided by the projected area,

LpipeDpipe. As depicted in Figure 6.9, the lateral forces for model pipe tests with aspect

ratios Lpipe/Dpipe = 6, 8, and 10 were extracted from cycle number 1.75 (tests at 4 m/s),

3.75 (tests at 2.75 m/s) and 5.75 (tests at 1.88 m/s) to ensure that the sample is

remoulded (initially at a degree of consolidation U = 10%).

As depicted in Figure 6.14, the pressure profiles fall within a narrow band with the

exception of the tests at a velocity of 4 m/s, in which the pressure on the model pipe of

Lpipe/Dpipe = 6 is approximately 6% and 9% higher than that of Lpipe/Dpipe = 8 and 10

respectively. Generally, these results show that end effects for model pipe tests in kaolin

slurry are not as significant as for tests in Newtonian fluids. Although a wide range of

aspect ratios was covered, no systematic influence of end effects has been observed. It is

acknowledged that these results should ideally be compared with test data of a model

pipe that is completely free from end effects (i.e. pure plane strain flow). This is

however, impossible in the UWA drum centrifuge channel in its current configuration.

A model pipe of Lpipe/Dpipe = 6 was chosen for subsequent tests, as a smaller model pipe

would allow T-bar tests to be performed near each model pipe test site in the adjacent

test zone that is uninfluenced by the model pipe.

RESULTS: HORIZONTAL FORCE 6.8

Overview 6.8.1

Figure 6.15 and Figure 6.16 show two typical examples of the velocity-displacement

and horizontal load-displacement profiles (with the horizontal load from the shaft

6.17

subtracted to the measured load as mentioned in Section 6.5). For subsequent analyses,

only the steady velocities and horizontal force values are considered. These values are

extracted, ignoring the initial build-up and decay of both force and velocity, and any

initial peak in resistance. An overview of the net horizontal pressure on the pipe qH at

the corresponding velocity v for all the tests is shown in Figure 6.17. No unique trend is

observed in the qH-v plot, indicating the need to normalise the variables on the x and y

axes. This normalisation is examined in the following subsections.

Viscous effects 6.8.2

Figure 6.18 shows the variation in the normalised resistance qH/qH-ref with increasing

normalised strain rate, refγ / γ . Here, the qH-ref is the horizontal pressure at a reference

strain rate ref , where the test strain rate, is assumed to be v/Dpipe (Zhu and Randolph

2011). ref corresponds to the pipe test at v = 0.004 m/s (test 1PS-2), which has the same

strain rate as the T-bar tests (0.2 s-1

). It should be noted that only the tests in Sample 1

(fully consolidated) are used here, which is shown later to imply that inertial drag

effects are negligible. The following power law relationship for the increase in qH/qH-ref

fits the results with good agreement:

m

H

H ref ref

q γ

q γ

6.14

with a best fit shear thinning parameter m = 0.1, which is within the documented range

of 0.05 – 0.17 (Biscontin and Pestana 2001; Jeong et al. 2009; Boukpeti et al. 2012)

typical for fine-grained soils.

In the absence of inertial effects, the increase in qH can be attributed to an increase in

the operative shear strength su-op of about 26% per log cycle (which corresponds to m =

0.1) as shown in Equation 6.2. The relationship of qH with su-op for all the tests is shown

in Figure 6.19. Since m is independent of the water content (Boukpeti et al. 2012), m =

0.1 is assumed to be sufficient to represent su-op in the vicinity of the pipe during testing

at high γ regardless of the degree of consolidation. To derive su-op, the reference

6.18

strength su-ref is calculated from Equation 6.10 where su-ref is taken as equal to su-in. At

su-op higher than ~ 0.7 kPa, the relationship appears to be linear where the slope is equal

to the lateral bearing capacity factor NH. However, no linear trend is observed at su-op

lower than ~ 0.7 kPa, indicating that the qH is not entirely a function of su-op in ultra-soft

fluid-like soil. To reconcile this discrepancy in fluid-like conditions, the first step is to

revert to a fluid mechanics approach of assessing the resistance via a drag factor linked

to the fluid density.

Inertial drag coefficient (fluid mechanics approach) 6.8.3

Using the approach of Zakeri et al. (2008), Zakeri et al. (2009) and Zakeri (2009), the

drag coefficient CD (Equation 6.3) back-calculated from each pipe test is plotted against

the non-Newtonian Reynolds number Renon-Newtonian (Equation 6.4) in Figure 6.20. The

Renon-Newtonian is derived by taking su-op = τ. The derived CD spans a huge range,

exceeding 1000000 at the lowest Renon-Newtonian.

The large range of CD and in particular the inverse dependence on Renon-Newtonian, arises

because the fluid mechanics approach lumps the shear strength and inertia contribution

into a single drag coefficient (i.e. CD), therefore masking the relative influence of

strength and inertia. In the creeping flow region (plastic to viscous regime), CD is more

appropriately a function of the soil strength (Deglo De Besses et al. 2003; Zhu and

Randolph 2011). This indicates the need to separate the resistance into two components

to capture both the inertia-dominated and strength-dominated domains. Such a unified

framework is examined in the next subsection.

Hybrid approach 6.8.4

To begin deriving the hybrid interpretation, the lateral resistance qH for all the tests can

first be expressed using the normalised lateral bearing capacity factor NH:

HH

u op

qN

s

6.15

6.19

The relationship between NH and Renon-Newtonian is shown in Figure 6.21. The following

hybrid relationship superposes separate drag and bearing components:

2

H D H u op

1q C ρv N s

2 6.16

From regression analysis, the best fit values of CD and NH are 1.06 and 7.35 respectively

(R2 = 0.99). At low Renon-Newtonian, where su-op dominates, qH is equal to NHsu-op. When

the velocity is high enough and the magnitude of su-op is very low, NH increases linearly

with Renon-Newtonian, highlighting the dominance of inertia (first term of Equation 6.16)

over the strength component (second term of Equation 6.16). The relative contributions

of the strength and inertia terms is also illustrated on Figure 6.19 by the dotted line at

NH = 7.35. The increased resistance above this line corresponds to the added force from

the inertial term.

The best fit NH of 7.35 is somewhat lower than the range of 9.14 – 11.94 (Randolph and

Houlsby 1984), which applies to deeply embedded cylinders. This difference is due to

the embedment of z = 2.5Dpipe (from the top of the soil surface to the pipe mid-height),

which is not sufficient to establish a full-flow failure mechanism. Oliveira et al. (2010)

developed an approximate analytical model to estimate the NH which is applicable at z

≤ 3.5Dpipe:

1

H

pipe

zN 5 tan 0.5

D

6.17

The resulting shallow NH estimated from Equation 6.17 is 6.30, which is compared to

the centrifuge data in Figure 6.21.

Also plotted in Figure 6.21 are test data obtained from both flume tests (Zakeri et al.

2008) and centrifuge tests conducted at 30 g (Zakeri et al. 2011). These results have

been recalculated from the previously published values into the form of the normalised

resistance NH. Both sets of experiments were conducted by launching a slide material

towards a stationary suspended model pipe (passive loading). The Zakeri et al. (2008)

6.20

tests used clay-sand slurries in which inertial effects dominated. The Zakeri et al. (2011)

tests used solid blocks of kaolin clay, where the strength component dominated. These

test data are tabulated in Table 6.6 (flume data) and Table 6.7 (centrifuge data). It

should be noted that not all the data in the two tables are documented in Zakeri et al.

(2008) and Zakeri et al. (2011), but some additional values have been subsequently

derived.

For each of the flume test data (Table 6.6), Dpipe is calculated from γ in the Herschel-

Bulkley equation by deducing the shear stress τ from Renon-Newtonian and making use of

the test velocity v (since γ = v/Dpipe). The pipe length Lpipe is assumed to be equal to the

width of the test flume (0.2 m) and su-op = τ. qH is then calculated from the horizontal

force H divided by the pipe projected area (= LPipe × Dpipe).

In the tabulated data for the centrifuge test (Table 6.7), su-op of the clay block is

calculated using Equation 6.2. This is done by taking the reference strength su-ref and the

reference strain rates refγ from the T-bar tests (DT-bar of 7.5 mm pushed at 3 mm/s) and

assuming m = 0.1 (since kaolin was also used in the centrifuge test of Zakeri et al.

(2011).

It may be seen from Figure 6.21 that the fitted line through the flume and centrifuge

data follows the same trend as the present centrifuge data best fit line, albeit with a

higher CD of 1.43. This is to be expected, as the pipe in the flume test was suspended

and free from near surface effects. However, at low Renon-Newtonian, despite the model

pipe being suspended (where a full-flow mechanism is expected to occur) in the

centrifuge experiments, the NH is only 1% higher than that in the present study. This is

perhaps partly attributable to the uncertainty in the magnitude of m, which may be

lower. Overall, both trend lines for the flume and centrifuge tests (Zakeri et al. 2008;

Zakeri et al. 2011) and the present study show that the inertial term in Equation 6.16

dominates when Renon-Newtonian > ~3.

Equation 6.16 is similar to that established by Randolph and White (2012) where a

linear relationship is obtained when CD (derived from Equation 6.3) is plotted against

1/Renon-Newtonian. However, this approach also yields a very wide range of CD as seen in

Figure 6.20. By plotting NH on the y-axis, the influence of fluid drag can be easily

6.21

demonstrated, which is reflected by the sudden increase in NH beyond a constant value

where the purely soil mechanics framework is no longer applicable.

RESULTS: PORE WATER PRESSURE RESPONSES 6.9

Figure 6.22 depicts the evolution of the normalised total pore pressure difference,

(uf - ub)/su-op with the non-Newtonian Reynolds number Renon-Newtonian. In this case, uf is

the total pore pressure measured from the front PPT facing the soil flow, ub is the total

pore pressure measured from the back PPT facing the wake and su-op is the operative

strength enhanced for viscous effects via Equation 6.2. Interestingly, the trend of

constant normalised total pore pressure ratio at Renon-Newtonian < 3 and the subsequent

increase at Renon-Newtonian > 3 is the same as previously seen for the bearing factor (NH)

vs Renon-Newtonian plot in Figure 6.21.

The factors that contribute to the changes in (uf - ub)/su-op when Renon-Newtonian > 3 are not

straightforward to identify in the absence of direct observations of the flow pattern

around the model pipe. However, as these changes occur in fluid-like kaolin samples

(high Renon-Newtonian), some inference may be drawn from the boundary layer separation

concept common to Newtonian fluid flow around a cylinder at high Newtonian

Reynolds number, ReNewtonian. This concept is depicted schematically in Figure 6.23. At

high ReNewtonian, the boundary layer within which viscous effects are important becomes

thin. Beyond this layer, the flow of fluid may be treated as inviscid. Owing to viscous

effects, the fluid particle experiences a loss in kinetic energy when flowing from

zones/points 1 to 3 and the boundary layer separates from the pipe surface at zone/point

2, causing a region of low pressure behind the pipe (decreasing at higher ReNewtonian). An

important phenomenon that may arise from this separation is vortex shedding at the

boundary shear layer, which generally occurs at around ReNewtonian ≥ 40 (e.g. Sumer and

Fredsøe 1997).

The sudden increase in the (uf - ub)/su-op ratio for the present centrifuge test data at Renon-

Newtonian > 3, could be caused by vortex shedding, where the back side of the pipe facing

the wake experiences a pressure decrease. Vortex shedding causes uneven pressure

distributions to develop at the top and bottom surfaces of the pipe. Consequently, an

6.22

oscillating vertical lift force is generated (Zakeri et al. 2008; Zakeri et al. 2009).

Attempts to investigate the oscillatory vertical forces were made but, the current

arrangement of the axial gauge (Figure 6.5) is susceptible to bending strains induced

from the horizontal drag forces. Therefore, the axial readings generated when the model

pipe was displaced horizontally, are not reliable. However, evidences of the vortex

shedding phenomenon in non-Newtonian flow past a pipe are documented both

experimentally (Zakeri et al. 2008) and numerically (Zakeri et al. 2009). The present

study suggests that the onset of vortex shedding is at Renon-Newtonian ~ 3 – beyond which

the inertial fluid drag term also dominates (see Section 6.8.4) – whereas the flume

experiments (Zakeri et al. 2008) and the numerical simulations (Zakeri et al. 2009)

showed comparable but slightly different onsets of vortex shedding at Renon-Newtonian ~

20 and 130 respectively. Given the wide range of Renon-Newtonian under consideration, the

different pipe embedment depths, and the different methods used to (i) identify vortex

shedding and (ii) define the relevant shear stress, these discrepancies are small.

ESTIMATION OF MEAN VERTICAL LIFT FORCE 6.10

Despite the unsuccessful attempts to quantify the vertical lift forces in the present study,

an analysis to estimate this force based on the drag coefficient may be developed from

the flume test data reported by Zakeri et al. (2008) as tabulated in Table 6.6. It should be

noted that for each data, the mean vertical uplift force VL is calculated based on the

average value of the oscillating vertical force range. The data is reinterpreted by

normalising the mean vertical uplift pressure (qV-L) by the su-op (assuming τ = su-op) and

plotted against the Renon-Newtonian in Figure 6.24. A linear relationship is evident, which

takes the form of:

2

V L L

1q C ρv

2 6.18

where the best fit gives CL = 0.95. Using the results shown in Figure 6.24, Equation

6.18 may be interpreted as follows:

6.23

1. At Renon-Newtonian < 20, the vertical uplift force is minimal. This is supported by

other model pipe centrifuge experiments reported by Oliveira et al. (2010) and

Zakeri et al. (2011).

2. At Renon-Newtonian ≥ 20, a non-zero mean vertical pressure is exerted on the pipe

which is a function of the inertial term that is characterised by CL = 0.95. The

contribution of the strength component is negligible, owing to the low su-op.

The above method can only be used to estimate the mean vertical force, but, not the

time dependent oscillating vertical force. However, the data tabulated in Table 6.6 (with

the exception of test at Renon-Newtonian = 62.2) suggests that the oscillating component of

the vertical force has a magnitude that is generally 5 – 25% of the mean vertical force.

CONCLUSIONS 6.11

The resistance of a buried model pipe moving horizontally in a stationary soil was

investigated by means of centrifuge model tests, exploring the influences of soil

strength and density, and the relative pipe-soil velocity. These results are analogous to

the passing of a submarine slide across a stationary pipeline, and the resulting

calculation method is applicable to this problem.

A model pipe 20 mm in diameter and 120 mm long was displaced horizontally at an

embedment ratio of 2.5 pipe diameters (measured from the soil surface to the mid-depth

of the pipe). Displacements were performed at velocities ranging from 0.004 to 4.2 m/s,

in soil samples with shear strengths ranging from 0.08 to 1.7 kPa – reflecting the

different degrees of consolidation. These tests were performed in 13 drum centrifuge

soil samples prepared using an identical self-weight consolidation procedure.

The centrifuge tests results demonstrate that:

1. The evolution of the soil density and shear strength inferred from the T-bar and

model pipe tests for all soil samples fell within a narrow band and showed the

same increasing trend as consolidation progressed. This demonstrates consistent

behaviour across all of the samples, thus allowing the soil parameters critical for

the interpretation of the pipe test results to be deduced reliably.

6.24

2. Test results from lateral movement using different model pipe length to diameter

ratios, pushed at velocities of 1.88, 2.75 and 4 m/s in ultra-soft kaolin slurry

showed that end effects are insignificant.

3. A hybrid approach to assess the pipe-soil loading, unifying both the soil

mechanics and the fluid mechanics approaches, was established and calibrated

from the model pipe test results. This model shows that for a non-Newtonian

Reynolds number (Renon-Newtonian) lower than 3, pipe-soil interaction is dominated

by the strength component and may be characterised by a bearing factor NH of

7.35 for an embedment level of 2.5 pipe diameters. At different levels of

embedment, a different factor is required, as can be derived from conventional

plasticity solutions. For Renon-Newtonian higher than 3, the inertial drag term

increases linearly with Renon-Newtonian and can be accounted for, in this case, by

using a constant mean drag coefficient CD of 1.06. The flume test data of Zakeri

et al. (2008) and centrifuge data reported by Zakeri et al. (2011) was

reinterpreted in light of the hybrid approach and similarly show that the Renon-

Newtonian = 3 represents a boundary below which the strength component would

dominate the drag resistance.

4. Interpretation of the pore pressure data gathered from the pore pressure

transducers located in the front (facing the upstream soil) and the back (facing

the wake) suggests the presence of vortex shedding at Renon-Newtonian > 3. This

observation is similar to the behaviour seen in other studies.

5. The mean vertical lift forces induced by vortex shedding documented by Zakeri

et al. (2008) was reanalysed using the hybrid approach similar to that applied for

the horizontal drag resistance data. It is found that the mean vertical lift is solely

governed by the inertial term.

This study provides experimental validation of a new approach to assess the impact

forces on subsea pipelines that are subjected to flows ranging across the solid-fluid

boundary. The approach is simple and theoretically robust from both a fluid mechanics

and solid mechanics perspective.

6.25






the CSIRO Wealth from Oceans Flagship Cluster on Subsea Pipelines. The authors are

also grateful to Bart Thompson, Phil Hortin, Shane De Catania, Dave Jones, Tuarn

Brown, David Yong and Khin Seint for their technical assistances and suggestions. The

first author is also grateful for the financial support received from the Ministry of

Higher Education Malaysia (MOHE) and Universiti Malaysia Sarawak (UNIMAS).

6.26

Table 6.1. Test programme in Sample 1: T-bar tests

Test

name Time (min)

Sample

U%

Initial

penetration

depth, z/DT-bar

Cyclic

depth,

z/DT-bar

Cycle

number

T1-1 182 24.39 9.5 4-7 9.75

T1-2 353 40.2 9.5 4-7 9.75

T1-3 478 49.88 9.5 4-7 4.75

T1-4 1268 91.03 9.5 4-7 9.75

T1-5 1333 92.31 9.5 4-7 9.75

T1-6 1473 94.3 9.5 4-7 9.75

T1-7 1762 95.62 9.5 4-7 9.75

T1-8 2003 97.03 9.5 4-7 9.75

T1-9 2720 98.82 9.5 4-7 9.75

T1-10 3112 98.99 9.5 4-7 9.75

T1-11 3463 99.08 9.5 4-7 9.75

T1-12 4234 99.08 9.5 4-7 9.75

T1-13 4902 99.96 9.5 4-7 9.75

T1-14 6292 99.99 9.5 4-7 9.75

T1-15 10116 100 9.5 4-7 9.75

T1-16 12732 100 9.5 4-7 9.75

Table 6.2. Test programme in Sample 1: Model pipe/shaft tests

Test

name Type of test

Sample

U%

Test

velocity

(m/s)

Test distance u/Dpipe Embedment ratio, z/Dpipe

1-PS1 Model pipe test 100 0.005 5.5 2.5




1-S1 Shaft test for 1-PS1 100 0.005 5.5 *




* Shaft embedded to a depth of 40 mm (measured from the tip of the shaft to the soil surface)

6.27

Table 6.3. Test programme in Sample EET: End effects test

Test

name

Model

pipe

aspect

ratio,

Lpipe/Dpipe

Embedment

ratio, z/Dpipe

Sample

U% Type of test

Test

velocity

(m/s)

Cyclic

distance,

u/Dpipe

Cycle

number

EET-1 6 2.5 10 Model pipe 4 132.5 0.25 - 1.75

EET-2 6 2.5 10 Model pipe 2.75 86 2.25 - 3.75

EET-3 6 2.5 10 Model pipe 1.88 52.2 4.25 - 5.75

EET-4 8 2.5 10 Model pipe 4 132.5 0.25 - 1.75

EET-5 8 2.5 10 Model pipe 2.75 86 2.25 - 3.75

EET-6 8 2.5 10 Model pipe 1.9 52.2 4.25 - 5.75

EET 7 10 2.5 10 Model pipe 4 132.5 0.25 - 1.75

EET-8 10 2.5 10 Model pipe 2.75 86 2.25 - 3.75

EET- 9 10 2.5 10 Model pipe 1.88 52.2 4.25 - 5.75

EET-S1 - * 10 Shaft test 4 132.5 0.25 - 1.75

EET-S2 - * 10 Shaft test 2.75 86 2.25 - 3.75

EET-S3 - * 10 Shaft test 1.88 52.2 4.25 - 5.75


6.28

Table 6.4. Test programme in Samples 3 - 12: T-bar tests

Sample Test

name Time (min)

Sample

U%

Initial penetration

depth, z/DT-bar

Cyclic depth,

z/DT-bar

Cycle

number

Sample 3 T3-1 482 50.09 9.5 4-7 9.75 Sample 3 T3-2 1515 94.49 8.3 4-7 9.75 Sample 3 T3-3 1828 96.01 8.5 4-7 9.75 Sample 3 T3-4 2833 98.87 8.5 4-7 9.75 Sample 4 T4-1 802 66.76 10.5 4-7 9.75 Sample 4 T4-2 1162 85.51 9.5 4-7 9.75 Sample 4 T4-3 1418 93.52 9.1 4-7 9.75 Sample 5 T5-1 326 37.70 9.5 4-7 9.75 Sample 5 T5-2 382 42.44 9.5 4-7 9.75 Sample 5 T5-3 407 44.38 9.9 4-7 9.75 Sample 6 T6-1 209 26.88 9.5 4-7 7.75 Sample 6 T6-2 275 32.99 9.5 4-7 7.75 Sample 6 T6-3 492 50.61 11.8 4-7 9.75 Sample 6 T6-4 1171 85.98 9.1 4-7 9.75 Sample 6 T6-5 1259 90.56 8.9 4-7 9.75 Sample 7 T7-1 317 36.87 9.5 4-7 9.75 Sample 7 T7-2 378 42.13 12.5 4-7 9.75 Sample 7 T7-3 928 73.32 10.3 4-7 9.75 Sample 7 T7-4 943 74.10 9.9 4-7 9.75 Sample 7 T7-5 1055 79.94 9.7 4-7 9.75 Sample 7 T7-6 1684 95.26 8.6 4-7 9.75 Sample 7 T7-7 1699 95.33 8.6 4-7 9.75 Sample 8 T8-1 355 40.35 10.2 4-7 9.75 Sample 8 T8-2 370 41.51 12.5 4-7 7.75 Sample 8 T8-3 880 70.82 9.7 4-7 7.75 Sample 8 T8-4 895 71.60 9.9 4-7 7.75 Sample 8 T8-5 1102 82.39 9.3 4-7 9.75 Sample 8 T8-6 1116 83.11 9.1 4-7 6.75 Sample 8 T8-7 1126 83.64 9.4 4-7 7.75 Sample 8 T8-8 1608 94.91 8.3 4-7 7.75 Sample 8 T8-9 1811 95.91 8.3 4-7 9.75 Sample 8 T8-10 1822 95.97 8.5 4-7 9.75 Sample 9 T9-1 332 38.25 12.3 4-7 9.75 Sample 9 T9-2 376 41.98 12.3 4-7 9.75 Sample 9 T9-3 390 43.06 12.3 4-7 7.75 Sample 9 T9-4 989 76.50 9.5 4-7 7.75 Sample 9 T9-5 999 77.02 9.6 4-7 9.75 Sample 9 T9-6 1083 81.40 9.3 4-7 9.75 Sample 9 T9-7 1373 92.88 8.9 4-7 9.75 Sample 9 T9-8 1385 93.05 8.7 4-7 9.75 Sample 9 T9-9 1599 94.87 8.4 4-7 9.75 Sample 9 T9-10 1770 95.67 8.5 4-7 9.75 Sample 9 T9-11 2744 98.83 8 4-7 9.75

Sample 10 T10-1 315 36.68 13 4-7 6.75 Sample 10 T10-2 382 42.44 12.2 4-7 8.75 Sample 10 T10-3 541 53.16 11.7 4-7 7.75 Sample 10 T10-4 557 54.00 11.7 4-7 8.75 Sample 10 T10-5 1206 87.80 9.1 4-7 8.75 Sample 10 T10-6 1216 88.32 9.1 4-7 7.75 Sample 10 T10-7 1316 91.98 9 4-7 7.75 Sample 10 T10-8 2684 98.73 8.1 4-7 9.75 Sample 11 T11-1 376 41.98 12.6 4-7 6.75 Sample 11 T11-2 387 42.83 12.5 4-7 6.75 Sample 11 T11-3 1204 87.70 9 4-7 9.75 Sample 11 T11-4 1214 88.22 9.2 4-7 8.75 Sample 11 T11-5 1574 94.76 8.5 4-7 9.75 Sample 11 T11-6 1587 94.82 8.5 4-7 7.75 Sample 11 T11-7 1597 94.86 8.5 4-7 7.75 Sample 12 T12-1 258 31.41 13.1 4-7 6.75 Sample 12 T12-2 386 42.75 12.2 4-7 5.75 Sample 12 T12-3 534 52.80 11.5 4-7 6.75 Sample 12 T12-4 553 53.79 11.3 4-7 6.75 Sample 12 T12-5 1264 90.82 9.1 4-7 9.75 Sample 12 T12-6 1592 94.84 8.8 4-7 9.75 Sample 12 T12-7 1605 94.90 8.8 4-7 9.75

6.29

Table 6.5. Test programme in Samples 2 - 12: Model pipe/shaft tests

Sample

Test

name Type of test

Time

(min)

Sample

U%

Test

velocity

(m/s)

Test

distance,

u/Dpipe

Embedment

ratio,

Lpipe/Dpipe

Sample 2 2-S1 Shaft test for 5-PS1 486 50.53 4.2 139.3 *

Sample 3 3-PS1 Model pipe test 249 30.96 1.63 43.7 2.5


Sample 3 3-PS3 Model pipe test 1775 95.06 0.09 10 2.5






Sample 7 7-S1 Shaft test for 6-PS1 75 12.35 4.04 135 *




















Sample 10 10-PS3 Model pipe test 424 45 0.25 13.8 2.5








Sample 11 11-PS2 Model pipe test 1643 95 0.007 6.4 2.5




Sample 12 12-S1 Shaft test for 11_PS1 291 34.87 3.18 101.6 *

Sample 12 12-S2 Shaft test for 10_PS3 424 45 0.25 13.8 *


Sample 12 12-S4 Shaft test for 11_PS2 1643 95 0.007 6.4 *




6.30

Table 6.6. Drag and vertical forces data from Zakeri et al. (2008) – Flume test

Non

-New

ton

ian

Reyn

old

s n

um

ber,

Re

no

n-N

ewto

nia

n

Den

sity

, ρ

(kg/m

3)

Hori

zon

tal

velo

city

, v

(m/s

)

Hori

zon

tal

forc

e, H

(N)

Osc

illa

tin

g

vert

ical

forc

e r

an

ge

(N)

*1M

ean

vert

ical

lift

forc

e, V

L

(N)

Hers

chel-

Bu

lkle

y

equ

ati

on

*2

Pip

e

dia

mete

r,

Dp

ipe (m

)

*3

Operati

ve

str

en

gth

,

su

-op (

Pa)

*4

Pip

e

len

gth

,

Lp

ipe (m

)

*5q

H/s

u-o

p*

6q

V-L

/su

-op

30

.71

68

5.7

0.8

35

8.2

2.1

- 3

.42

.75

0.0

28

63

8.3

0.2

37

.41

2.6

11

61

68

11

.14

69

.83

.8 -

5.9

4.8

50

.02

22

19

0.2

90

57

.4

49

.51

68

5.7

1.0

86

11

.64

- 6

.75

.35

0.0

28

64

0.2

0.2

50

.52

3.3

62

.21

68

5.7

1.2

31

23

.2 -

7.7

5.4

50

.02

86

41

0.2

51

.22

3.2

64

.91

68

5.7

1.2

59

12

.56

.2 -

8.7

7.4

50

.02

86

41

.20

.25

3.1

31

.6

22

.81

68

7.7

1.0

26

14

3.8

- 4

.74

.25

0.0

28

67

7.9

0.2

31

.49

.5

56

.11

68

5.7

1.1

63

9.8

5.6

- 7

.46

.50

.02

86

40

.60

.24

2.2

28

22

.51

68

7.7

1.0

27

.76

.6 -

7.4

70

.02

86

78

0.2

17

.21

5.7

20

.71

68

7.7

0.9

95

6.7

3.3

- 4

.23

.75

0.0

22

28

0.7

0.2

14

.51

0.5

* D

ata

no

t o

rig

inally

in

Zakeri

et

al. (

2008)

1 A

vera

ge o

f th

e o

scilla

tin

g v

ert

ical fo

rce r

an

ge

2 B

ack-c

alc

ula

ted

fro

m v

, R

en

on

-New

ton

ian

an

d t

he c

orr

esp

on

din

g H

ers

ch

el-

Bu

lkle

y e

qu

ati

on

3 A

ssu

min

g s

u-o

p =

τ

4 A

ssu

min

g L

pip

e is

th

e s

am

e a

s t

he w

idth

of

the t

est

flu

me

5 q

H =

H/L

pip

eDp

ipe

6 q

V-L

= V

L/L

pip

eDp

ipe

6.31

Table 6.7. Drag force data from Zakeri et al. (2011) – Centrifuge test

4.1

71

58

0.7

60

.16

0.0

06

35

24

.41

61

.02

6.2

95

1.3

78

.17

0.0

06

4.6

91

58

0.7

60

.21

0.0

06

35

32

.91

82

.28

7.2

95

2.1

47

.15

0.0

09

5

4.4

61

58

0.7

60

.11

0.0

06

35

16

.54

41

.34

6.4

74

5.5

97

.05

0.0

02

7

4.1

71

58

0.7

60

.11

0.0

09

51

11

.04

27

.65

.81

37

.19

6.4

0.0

03

4.1

71

58

0.7

60

.20

.00

95

12

1.3

55

3.3

66

.21

40

.25

6.4

90

.01

05

4.3

15

80

.76

1.3

0.0

09

51

13

6.7

34

1.7

57

.71

64

.09

8.3

20

.34

7

5.9

51

58

0.7

60

.78

0.0

09

51

81

.91

20

4.7

81

0.1

37

6.5

37

.55

0.0

94

7

* D

ata

no

t o

rig

inally

in

Zakeri

et

al. (

2011)

1 C

alc

ula

ted

usin

g E

qu

ati

on

6.1

3 (

Gs

= 2

.6, av

era

ge v

oid

s r

ati

o, e

v =

1.7

6)

2 c

alc

ula

ted

based

on

a T

-bar

dia

mete

r o

f 7.5

mm

at

a p

en

etr

ati

on

rate

of

3 m

m/s

(Z

akeri

et

al. 2

011)

3 C

alc

ula

ted

usin

g E

q. 6.2

(assu

min

g m

= 0

.1)

4 C

alc

ula

ted

based

on

Eq

uati

on

6.4

(assu

min

g τ

= s

u-o

p)

H

oiz

on

tal

pressu

re,

qH

(k

Pa)

*q

H/s

u-o

p

*4N

on

-New

ton

ian

Reyn

old

s n

um

ber,

Re

no

n-N

ewto

nia

n

*S

train

rate

,

Refe

ren

ce

stre

ng

th,

s u-r

ef

(T-b

ar)

(k

Pa

)

*1D

en

sit

y, ρ

(kg

/m3)

H

ori

zo

nta

l

velo

city

, v

(m/s

)

Pip

e

Dia

mete

r,

Dp

ipe

(m)

*2

*3

Operati

ve

str

en

gth

su

-op

(k

Pa)

/re

fγγ

=v

/

(s-1

)

6.32

Figure 6.1. UWA drum centrifuge

Figure 6.2. National Instruments® PCI-7344 motion controller card (extracted from the National

Instruments® website: http://sine.ni.com/nips/cds/view/p/lang/en/nid/13244)

Tool table

Sample containment channel

6.33

Figure 6.3. New UWA drum centrifuge tool-table circumferential movement control interface

6.34

Figure 6.4. T-bar penetrometer

Figure 6.5. Instrumented model pipe

6.35

Figure 6.6. Test setup in the drum centrifuge channel

Figure 6.7. Scale used to monitor sample filling and settlement – photo taken from an on-board

camera shown in Figure 6.6

200 mm scale for measuring

sample settlement (Figure 6.7)

Upper channel

0 mm10 20

30 40 50

60 70 80 90

100 110 120

130 140 150

160 170

180 190

200 mm

300 mm

Drum centrifuge channel

Lower channel

Base PPT

Cyclic T-bar

test

10 × 40 mm T-bar

(Figure 6.4)

20 × 120 mm model

pipe (Figure 6.5)

2.5Dpipe

Initial height of kaolin

slurry = 184 mm

Water level maintained

at 200 mm

On-board camera to

monitor scale

z

184 mm mark

Lower

channel

6.36

Figure 6.8. General experimental programme flowchart

Investigate consolidation

behaviour of kaolin slurry

(Sample 1)

∆ht, ρ,

su-in

Model pipe test in fully

consolidated soil (Sample 1)

H (U=100%),

uf, ub

Investigate pipe end effects

Lpipe/Dpipe = 6, 8, 10 (Sample

EET)

End effects significant

End effects negligible

Overcome end effects problem on

model pipe

Model pipe test in consolidating

soil (Samples 2 – 12)

H (U<100%),

uf, ub

NH, CD, uf, ub

∆ht, ρ,

su-in, H

∆ht, ρ,

su-in

LEGEND

Lpipe – pipe length

Dpipe – pipe diameter

∆ht – sample settlement

uf – front PPT total pore

pressure

ub – back PPT total pore

pressure

H – horizontal force

ρ – soil density

su-in – soil shear strength

NH – bearing capacity factor

CD – drag coefficient

Not pursued in experiments

6.37

Figure 6.9. End effects test (EET) test sequence

0

1

2

3

4

5

0 1 2 3 4 5 6

Tes

t v

elo

city

(m

/s)

Cycle number

Used for comparison

Backward sweep

Forward sweep

1) EET – 1,2,3

2) EET – 4,5,6

3) EET – 7,8,9

4) EET – S1, S2, S3

6.38

Figure 6.10. Measured sample height during consolidation

Figure 6.11. Degree of consolidation, U% of centrifuge test samples (Note: U% is calculated based

on each sample settlement (Equation 6.8)

80

100

120

140

160

180

200

1 10 100 1000 10000 100000

Sa

mp

le h

eig

ht

(mm

)

Time (mins)

Sample 1 Sample EET

Sample 2 Sample 3

Sample 4 Sample 5

Sample 6 Sample 7

Sample 8 Sample 9

Sample 10 Sample 11

Sample 12

0

20

40

60

80

100

120

1 10 100 1000 10000 100000

Deg

ree

of

co

nso

lid

ati

on

, U%

Time (mins)

Sample 1 Sample EET

Sample 2 Sample 3

Sample 4 Sample 5

Sample 6 Sample 7

Sample 8 Sample 9

Sample 10 Sample 11

Sample 12

6.39

Figure 6.12. Evolution of undrained shear strength at z = 2.5Dpipe

Figure 6.13. Change in shear strength with U%

0

0.5

1

1.5

2

100 1000 10000 100000

Un

dra

ined

sh

ea

r s

tren

gth

(k

Pa)

Time (min)

T-bar

Model pipe test

Sahdi (2012)

0

0.5

1

1.5

2

0 20 40 60 80 100

Un

dra

ined

sh

ear

stre

ng

th (k

Pa

)

Degree of consolidation, U (%)

T-bar

Model pipe

Eq. 6.10

6.40

(a)

(b)

(d)

Figure 6.14. End effects tests in Sample EET (U = 10%) at velocities of: (a) 4 m/s; (b) 2.75 m/s;

(c) 1.88 m/s

0

5

10

15

0 500 1000 1500 2000 2500

Ho

rizo

nta

l p

ress

ure

on

pip

e, q

H (

kP

a)

Horizontal displacement, u (mm)

EET-1 (20 x 120 mm pipe)



0

2

4

6

8

0 500 1000 1500 2000

Hori

zon

tal

pre

ssu

re o

n

pip

e, q

H (

kP

a)

Horizontal Displacement, u (mm)




0

2

4

6

0 200 400 600 800 1000

Hori

zon

tal

pre

ssu

re o

n

pip

e, q

H (

kP

a)

Horizontal Displacement, u (mm)




6.41

(a)

(b)

Figure 6.15. Test results for 5-PS1(velocity v = 4.2 m/s, degree of consolidation U = 50.5%): (a)

velocity – displacement profile; (b) horizontal force – displacement profile

(a)

(b)

Figure 6.16. Test results for 11-PS4 (velocity v = 0.005 m/s, degree of consolidation U = 95%): (a)

velocity – displacement profile; (b) horizontal force – displacement profile

0

1

2

3

4

5

0 1000 2000 3000

Tes

t vel

oci

ty, v (

m/s

)


Steady

velocity

0

10

20

30

40

0 1000 2000 3000

Hori

zon

tal

forc

e on

pip

e, H

(N

)


Steady

horizontal force

0

0.002

0.004

0.006

0.008

0.01

0 50 100 150

Tes

t vel

oci

ty, v (

m/s

)


Steady velocity

0

5

10

15

20

0 50 100 150

Hori

zon

tal

forc

e on

pip

e, H

(N

)


Steady

horizontal force

6.42

Figure 6.17. Horizontal pressure on model pipe

Figure 6.18. Variation of normalised resistance in the viscous region (1-PS1 to 1-PS4)

0.1

1

10

100

0.001 0.01 0.1 1 10

Hori

zon

tal

pre

ssu

re o

n m

od

el p

ipe,

qH

(kP

a)

Horizontal velocity, v (m/s)

1-PS1 to 1-PS4 3-PS1 to 3-PS3 4-PS1 to 4-PS3

5-PS1 6-PS1 8-PS1 to 8-PS5

10-PS1 to 10-PS6 11-PS1 to 11-PS4 12-PS1

0.8

1

1.2

1.4

1.6

0 10 20 30 40 50 60

Norm

ali

sed

pre

ssu

re,

qH

/qH

-ref

Normalised strain rate,

Test 1 (U = 100%)

Eq. 6.14

ref

6.43

Figure 6.19. Variation of pipe resistance with the rate enhanced shear strength

Figure 6.20. Variation of drag coefficient with non-Newtonian Reynolds number

0.1

1

10

100

0.1 1 10

Ho

rizo

nta

l pre

ssu

re o

n m

od

el p

ipe,

qH

(kP

a)

su-op (kPa)


5-PS1 6-PS1 8-PS1 to 8-PS5

10-PS1 to 10-PS6 11-PS1 to 11-PS4 12-PS1

NH = 7.35

1

10

100

1000

10000

100000

1000000

10000000

0.00001 0.0001 0.001 0.01 0.1 1 10 100 1000

Dra

g c

oef

fici

ent,

CD

Renon-Newtonian = ρv2/su-op

1-PS1 to 1-PS4

3-PS1 to 3-PS3

4-PS1 to 4-PS3

5-PS1

6-PS1

8-PS1 to 8-PS5

10-PS1 to 10-PS6

11-PS1 to 11-PS4

12-PS1

6.44

Figure 6.21. Variation of normalised lateral pressure on pipe with non-Newtonian Reynolds

number

1

10

100

1000

0.00001 0.0001 0.001 0.01 0.1 1 10 100 1000

Ho

rizo

nta

l bea

rin

g c

ap

aci

ty f

act

or,

NH


NH = 7.43

CD = 1.43

NH = 7.35

CD = 1.06

NH = 6.30

Eq. 6.16 (fitted to Zakeri et al. 2008; 2011) 1-PS1 to 1-PS4

3-PS1 to 3-PS3 4-PS1 to 4-PS3

5-PS1 6-PS1

8-PS1 to 8-PS5 10-PS1 to 10-PS6

11-PS1 to 11-PS4 12-PS1

Zakeri et al. (2011) Zakeri et al. (2008)

Oliveira et al. (2010) Eq. 6.16 (fitted to current data)

6.45

Figure 6.22. Variation of normalised total pore pressure difference with non-Newtonian Reynolds

number

Figure 6.23. Boundary layer separation concept for Newtonian flow (ReNewtonian > 40)

1

10

100

1000

0.00001 0.0001 0.001 0.01 0.1 1 10 100 1000 10000

No

rma

lise

d to

tal p

ore

pre

ssu

re d

iffe

ren

ce,

(uf -u

b) /s

u-o

p



5-PS1 6-PS1 8-PS1 to 8-PS5

10-PS1 to 10-PS6 11-PS1 to 11-PS4 12-PS1

Front

PPT

(uf)

Back

PPT

(ub)

Soil

flow

v m/s1 Low pressure

2

3

Separation points

Boundary layer

Vortex

6.46

Figure 6.24. Mean vertical lift pressure

0

10

20

30

40

50

60

70

0.01 0.1 1 10 100 1000

No

rma

lise

d v

erti

cal

pre

ssu

re, q

v-L

/su

-op


Eq. 6.18 (CL = 0.96, NL ~ 0)

Negligible vertical force (Oliveira et al. 2010; Zakeri et al. 2011)

Zakeri et al. (2008)

7.1

CHAPTER 7. CONCLUDING REMARKS

INTRODUCTION 7.1

This thesis has focused on the factors contributing to the change in undrained shear

strength of fine-grained soils. These are important geotechnical design considerations

when dealing with: (i) submarine slide-offshore pipeline interaction (ii) cyclic offshore

pipeline-soil interaction due to repeated operational and environmental loadings

followed by calm periods. In order to better understand the role of the undrained

strength on the two aforementioned pipeline-soil interactions, two research objectives

were declared in Chapter 1 (Section 1.2), which are:

1. Study the effects of imposing a wide range of strain rates when shearing a clay

material. This objective is directly relevant when considering submarine slide-

offshore pipeline interaction.

2. Provide a better understanding on the remoulding characteristics of clay when

subjected to cyclic loading and the ensuing strength recovery due to a calm

period. These two competing effects are important when considering various

cyclic offshore pipeline-soil interactions.

For an overview of the above research objectives and the associated chapters addressing

them, readers are invited to revert to Figure 1.6.

This chapter provides a summary of the key outcomes of this research, and suggestions

for future research.

PRINCIPAL OUTCOMES: RATE EFFECTS 7.2

This section will outline the main outcomes of the chapters associated with Objective 1

(see Figure 1.6). It can be divided into considerations of viscous effects alone and a

coupled viscous-inertial treatment of the undrained shear strength of clays.

7.2

Viscous effects 7.2.1

The effect of strain rates on the undrained shear strength of clay was first addressed in

Chapter 3 using the novel vertically oriented penetrometer (VOP). The motivation for

the innovation of this new penetrometer stems from the need to better characterise the

near-surface shear strength for pipeline design, and to provide continuous shallow shear

strength characterisation along a pipeline route, which can extend to more than 100 km.

The VOP used a sensing element that is analogous to a shallowly-embedded rigid pile.

Accordingly, the VOP resistance can be back-analysed using a total stress bearing

capacity concept of a laterally loaded pile (e.g. Randolph and Houlsby 1984; Murff and

Hamilton 1993; Martin and Randolph 2006) to derive the near-surface shear strength.

Part of the overall study reported in Chapter 3, focused on the range of strain rates

relevant to the early stages of a submarine slide-pipeline interaction – i.e. the viscous

regime. A parametric study of the VOP rate effects in both normally consolidated and

highly overconsolidated samples outlined in Chapter 3 showed that the viscous rate

dependency parameter for the intact and remoulded resistances is similar, but, higher for

the peak resistance, partially attributable to the development of suction during a two-

way wedge failure mechanism at small VOP lateral displacements.

Combined viscous-inertial effects 7.2.2

As an extension of the findings in Chapter 3, Chapter 6 described a hybrid soil-fluid

mechanics approach to quantify the pressure exerted by a submarine slide when it flows

over and past an offshore pipeline. This provides a link between the discontinuous

regimes that have been previously proposed: (i) the soil regime, when the slide strength

provides the dominant contribution to the resistance – i.e. the viscous regime (Chapter

3) and (ii) the fluid mechanics regime, when the slide density and velocity dominate -

i.e. the inertial regime. Extensive centrifuge tests, where the model pipe was dragged

horizontally at velocities ranging from 0.004 – 4.2 m/s, were conducted in consolidating

soil samples (with strengths ranging from 0.08 – 1.7 kPa). This passive drag test is

analogous to the passing of a submarine slide across a stationary pipeline (active

loading) and allowed the influence of soil strength, density and model pipe horizontal

velocity to be quantified in a controlled and systematic manner.

7.3

The cyclic T-bar test results discussed in Chapter 5 were utilised to quantify the

changing strengths and densities of the consolidating samples. The normalised recorded

pipe resistance, qH/su-op (where qH – horizontal pipe resistance and su-op – operative soil

strength, augmented for viscous effects) was shown to vary linearly with the non-

Newtonian Reynolds number, Renon-Newtonian, leading to a combined soil-fluid mechanics

formulation. This superposition method is similar to that proposed by Randolph and

White (2012) and can be used across the viscous-inertial regime without having to

invoke a separate soil/fluid mechanics model. Reinterpretation of existing data in the

literature from flume (Zakeri et al. 2008) and centrifuge (Zakeri et al. 2011) tests further

supports the usage of this hybrid approach.

Interpretation of the data gathered from the two pore pressure transducers on the model

pipe suggested the existence of vortex shedding beyond a critical threshold of non-

Newtonian Reynolds number. This phenomenon is similar to the onset of vortex

shedding in Newtonian flow pass a cylinder, thus highlighting the need for a method to

quantify the vertical forces that accompany vortex shedding. Based on data gathered

from previous studies using flume tests (Zakeri et al. 2008), it was further demonstrated

that the hybrid approach developed to estimate the qH can also be used to quantify the

vertical uplift pressure induced by vortex shedding.

PRINCIPAL OUTCOMES: EFFECTS OF CYCLIC 7.3

LOADING AND STRENGTH RECOVERY

The outcomes of the research portion tailored to address Objective 2 (Section 1.2.2) are

summarised in this section. These outcomes are divided in relation to the studies of:

1. The role of soil microfabric on undrained strength degradation due to cyclic

T-bar tests – Chapter 2.

2. A study of the deterioration of undrained strength from the ‘true’ intact to

remoulded states from VOP cyclic tests – Chapter 3.

3. A study of the remoulding and reconsolidation (strength recovery) of clays

via VOP tests – Chapter 4.

7.4

4. Investigate the correlations of strength degradation and index properties of

clay – Chapter 5.

Effect of soil microstructure on strength degradation 7.3.1

The study of the effect of soil microfabric on the soil shear strength has been limited so

far to laboratory element tests (e.g. Nagaraj 1964; Bai and Smart 1997; Sachan and

Penumadu 2007b; Pillai et al. 2011). Chapter 2 presented the first study on the effect of

kaolin microfabric on cyclic T-bar response. Different arrangements of kaolin platelets

were created by introducing various types of dyeing agents. T-bar tests on dyed and

undyed samples, normally consolidated in the drum centrifuge, demonstrated that the

shear strengths during the initial T-bar penetrations increased in varying magnitudes in

all, except one dyed cases, relative to the undyed samples. However, upon intense soil

remoulding via subsequent T-bar cyclic tests, the remoulded strengths for all samples

were identical. Index tests showed that plasticity indices of the dyed samples were

similar to that of undyed kaolin. Scanning Electron Microscopy (SEM) indicated that all

the samples (both dyed and undyed samples) each had a flocculated, random particle

orientation. The qualitative observation from the SEM was complemented by a

quantitative study using the x-ray diffraction (XRD) method. The Fabric Index, derived

from XRD studies, and which characterises the orientation of the kaolin platelets,

indicated that the increase in intact shear strength was associated with a higher degree of

edge-to-face particle contacts.

The findings in Chapter 2 is relevant to offshore pipe-soil interface interactions where

the axial movement of pipelines may cause the clay particles to become aligned parallel

to a very thin shearing zone (to the extend varying with pipe radius (White and Cathie

2010)) upon reaching the critical or residual state akin to that observed in a direct shear

interface or ring shear tests (Atkinson 2007). This chapter shows that the remoulded

strength as measured from continuous cyclic T-bar tests is similar despite the initially

different orientation of clay microstructure. It therefore establishes the T-bar remoulded

strength as a fundamental strength property that is not affected by some of the subtle

soil strength effects that can occur in field conditions. It therefore strengthens the

applicability of the observations from these experiments to practical conditions,

7.5

offshore. This cyclic T-bar test is analogous to the cyclic movement of pipelines during

storms or temperature induced buckling. Neglecting hardening of the foundation soil

due to reconsolidation (discussed in length in Chapter 4), the remoulded strength of the

foundation soil should be comparable irrespective of the initial clay microstructure.

Decay of strength from the undisturbed to remoulded states 7.3.2

Besides studying the viscous effects using the VOP (see Section 7.2.1), Chapter 3 also

focused on studying the strength degradation characteristics of clay from cyclic VOP

tests. As a potential inclusion in the current arsenal of site investigation tools for

shallow strength characterisation, this aspect is particularly important.

In Chapter 3, analyses of the VOP peak and steady resistances data confirmed that the

soil strength could be back-calculated via the laterally loaded pile framework,

considering a wedge mechanism (either one-way or two-ways) at a shallow depth and a

flow-round failure at greater depth. Careful analyses of the steady resistance showed

that an increase in the su/γ'DVOP (where su – soil undrained strength; γ' – soil effective

unit weight; DVOP – VOP diameter) ratio increases the depth where the 1-way wedge

mechanism evolves into the flow-round plane strain failure mechanism around the VOP.

The transition of the peak to steady resistance in soil with high su/γ'DVOP ratios is

governed by a loss of suction (associated with a 2-way wedge mechanism) at the VOP

rear. It was shown that if the su/γ'DVOP ratio is low, the transition from peak to steady

VOP resistance is governed by strain softening effect because the flow-round

mechanism dominates throughout the VOP length. This is akin to an initially wished-in-

place and subsequently displaced numerical simulation of a T-bar penetration in a strain

softening soil (Zhou and Randolph 2007; Zhou and Randolph 2009b), where the peak

resistance (neglecting rate effects), is the ‘true’ intact shear strength.

Neglecting the minor disturbance created during vertical embedment of the VOP in the

soil sample, the VOP is essentially also wished-in-place before lateral displacement,

which is not the case for conventional full-flow penetrometers. On this basis, the VOP

provides a simpler means in which the ‘true’ intact strength of a surficial soil can be

estimated. Cyclic VOP tests results (progressing from the intact to remoulded soil

7.6

strengths) shown in Chapter 3 further revealed that the VOP can provide comparable

estimates of the soil sensitivity to the T-bar penetrometer.

Degradation and recovery of undrained shear strength 7.3.3

The cyclic loading effects theme was extended in Chapter 4, by coupling the competing

effects of remoulding and reconsolidation. These effects were simulated in the beam

centrifuge, where VOP tests were performed at varying embedment depths, different

cyclic velocities and test distances. At high velocities, the degradation of the VOP

resistance was similar to that of the T-bar cyclic tests, regardless of VOP test

embedment and distance, reflecting the dominating effect of soil remoulding. For the

slower VOP cyclic tests, the VOP resistance dropped during the first few cycles –

reflecting remoulding – but showed a subsequent gain in VOP resistance - reflecting

pore pressure dissipation and reconsolidation.

The rate of increase in the VOP resistance intensified with decreasing test embedment

and velocity and increasing test distance, owing to the increased dominance of soil

strength recovery (through reconsolidation) over soil remoulding. An analytical

framework, which is an extension of the remoulding and reconsolidation model

postulated by White and Hodder (2010), was outlined in an attempt to replicate this

competing behaviour. The framework is extended to include a simple one-dimensional

dissipation model (towards the soil surface) between two passes of the VOP at a given

soil horizon. Using only minor optimisation of one of the framework parameters, the

framework predicted the evolution of the measured VOP resistance with cycle number

with reasonable agreement.

The documented VOP cyclic test results and the outlined analytical framework revealed

that soil strength recovery is possible, even within a cyclic shearing episode, provided

that the elapsed time between multiple shearing events is sufficiently long, with respect

to the drainage rate. This remoulding-reconsolidation sequence is directly analogous to

the cyclic movements of a pipeline within an engineered on-bottom lateral buckle that

occur due to temperature fluctuations during startup and shutdown episodes

(Rismanchian et al. 2011).

7.7

Strength degradation-index properties relationship 7.3.4

Chapter 5 has outlined a study into the origin of the parameters in the exponential

strength decay model (used in chapters 2 and 4), through experimental studies that

reveal how the ductility (N95) and true sensitivity (St-r) parameters of kaolin vary with

the soil state. The results of extensive centrifuge T-bar tests (81 tests) in samples of

consolidating kaolin slurry were analysed with the aid of large-strain numerical analyses

for self-weight consolidation. These produced reliable measurements of the T-bar initial

and remoulded strengths, and the associated changes in voids ratio and N95, important

also for the interpretation of the pipe results in Chapter 6 (see Figure 1.6). Following

further interpretation, a linear relationship between N95 and LI was established, and was

shown to compare well with data obtained from other cyclic T-bar tests performed on

five onshore natural clay sites and one centrifuge test on reconstituted offshore high

plasticity clay.

The centrifuge test results also confirmed that the intact and remoulded strengths

inferred from T-bar tests are somewhat different to those obtained from other methods

(e.g. simple shear for intact strength and vane shear for remoulded strength), leading to

an apparent sensitivity that is lower than St-r. The St-r-LI exponential correlation

proposed by Wood (1990) was validated by including additional onshore data, thus

allowing the intact strength to be estimated from the remoulded soil strength. Finally, an

empirical exponential decay model, which incorporates the newly proposed N95-LI and

the existing St-r-LI relationships, was formulated. This proposed empirical model

compared well to the T-bar cyclic tests obtained from recent tests on Burswood clay

(NGI-COFS 2006).

These correlations provide insight into the origin of sensitivity, and the ductility of this

remoulding behaviour, and provide a link between soil index properties with the St-r and

N95 parameters – which have previously been simply fitted to T-bar data.

7.8

OVERALL SUMMARY 7.4

This thesis has discussed in length the factors contributing to the change in undrained

shear strength of clay, which affects various pipeline-soil interactions. It was shown that

the initial undrained strength of clay (from T-bar tests) is influenced by the different

microstructural arrangements of clay particles, where upon severe cyclic remoulding,

have comparable remoulded strengths, nullifying the influence of initial clay

microstructure. This thesis has shown that it is possible to measure directly the ‘true’

undisturbed strength of clay using the novel VOP, which is an improvement over the T-

bar and ball penetrometers, where a strain-softened strength is often measured. By using

extensive centrifuge and existing field data, the exponential decay model often fitted to

the T-bar or ball cyclic data, was shown to correlate well with the index properties of

soil.

While cyclic remoulding decreases the strength of clay due to accumulation of positive

excess pore pressure, it can regain its strength through reconsolidation where positive

pore pressure dissipates. These two competing effects have been shown to exist even

within a cyclic loading episode, which is captured adequately using a critical state-one

dimensional consolidation type framework. Viscous and inertial effects present during a

submarine slide-pipeline scenario cause the undrained strength to increase also. It was

shown that the undrained strength of clay increase is limited by its viscosity, and an

additional soil inertial (density-velocity) term must be added to account for the increase

resistance of pipelines in the path of debris flows.

RECOMMENDATIONS FOR FUTURE RESEARCH 7.5

The following recommendations are made to advance the findings of this thesis:

1. The relationship between soil structure and strength degradation observed in T-

bar cyclic tests documented in Chapter 2 was only limited to different

flocculation degrees of kaolin platelets. The influence of soil structure on the T-

bar cyclic strength degradation response could be extended to include kaolin

with dispersed microfabric. Using triaxial tests, this face-to-face kaolin platelet

7.9

arrangement has been shown to exhibit a dilative nature despite being on the

‘wet’ side of the critical state line (Sachan and Penumadu 2007b).

2. The centrifuge study of the vertically oriented penetrometer (VOP) discussed in

Chapter 3 provides a useful validation of a simple tool that complements

existing methods of assessing soil strength in soil samples. The study was

limited to two extremes of strength ratios, su/γ'DVOP – normally and highly

overconsolidated clays. Future studies could systematically include uniform

clays with varying intermediate su/γ'DVOP ratios in order to further quantify the

depth at which full-flow failure occurs, thus potentially providing a more refined

framework to deduce suitable bearing factors for surficial shear strength

estimations. More centrifuge tests using different clay samples are also required

to confirm the capability of the VOP in estimating the ‘true’ intact soil strength.

The suitability of the VOP to infer the coefficient of consolidation using either

the dissipation or twitch test should also be investigated. It is also proposed that

the possibility of using the VOP in site investigation practice be explored –

perhaps from an ROV or AUV platform.

3. The remoulding and reconsolidation framework outlined in Chapter 4 suggests –

from extrapolation – that a steady state exists when the soil has reached the

ultimate remoulded state at the confining effective vertical stress where there is

no longer a tendency for excess pore pressure generation. This aspect should be

confirmed by conducting a cyclic VOP centrifuge test that allows intermittent

strength recovery through reconsolidation between two VOP passes at larger

cycle numbers than that documented in Chapter 4, where none of the tests

reached a steady state. If a steady state does exist, then the long term cyclic

resistance between a pipe and the seabed, undergoing any form of cyclic loading

with concurrent consolidation, should be constant and can be represented by a

single operative steady-state strength that could be estimated from this

framework. Further experimental studies could validate this prediction, which

would be a potentially very useful outcome from a design perspective.

4. The N95-LI (ductility-liquidity index) and St-r-LI (true sensitivity-liquidity index)

relationships should be further explored by including more extensive data from

both offshore and onshore test sites. If the relationship is proven for more soils,

7.10

and a micromechanical explanation for such correlation is established, this could

improve the utility of the correlations set out in Chapter 5.

5. The hybrid soil-fluid mechanics approach to estimate the impact load induced by

submarine slides on pipelines reported in Chapter 6 is perhaps most applicable

for suspended pipelines situated at pipeline crossings. As most export pipelines

are laid on the seabed, further centrifuge tests simulating this scenario should be

conducted. This could be done by repeating the test procedures outlined in

Chapter 6, but allowing a minimal clearance between the model pipe and a stiff

underlying clay layer.

6. This thesis has focused primarily on controlled centrifuge testing using

reconstituted kaolin having stress profiles corresponding to either normally or

overconsolidated clays. These stress histories simulated in the documented

centrifuge tests are relevant to those found in the Gulf of Mexico and offshore of

West Africa. Occasionally, unusual soil conditions may be encountered offshore,

which includes organic calcareous clays (found offshore Australia, for example)

and many more. Although these unusual soil conditions may not affect the

hybrid viscous-inertial equation outlined in Chapter 6 (since for debris flows, the

soil is already remoulded), future tests may be devised to simulate these unusual

soil conditions, particularly on the behaviour of cyclic degradation and the

ensuing recovery of undrained strength.

8.1

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