seismic performance of ductile and nominally ductile reinforced concrete moment resisting frames

5/25/2018 Seismic Performance of Ductile and Nominally Ductile Reinforced Concrete Moment Resisting Frames

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Seismic performance of ductile and nominallyductile reinforced concrete moment resistingframes. I. Experimental study

Andr Filiatrault, ric Lachapelle, and Patrick Lamontagne

Abstract: This paper is the first of two companion papers on the evaluation of the level of protection offered by ductile andnominally ductile reinforced concrete structures in Canada. In this paper, the seismic behaviour of two half-scale reinforcedconcrete moment resisting frames is investigated by shake table tests. In the second paper, the experimental results obtainedfrom the shake table tests are compared with the results generated from inelastic time-history dynamic analyses. Each framehad two bays and two storeys with overall dimensions of 5 m in width and 3 m in height. The first structure was designed as aductile frame according to current Canadian standards; and the second structure incorporated only nominally ductile details.Two levels of intensity were retained for the historical ground motion used in the tests. The first level was representative ofthe design earthquake considered; the amplitudes were doubled for the second intensity. The ductile structure performed wellduring both tests. The frame with nominal ductility performed well during the first test, but was on the verge of collapse after

the second test. Based on these experimental results, recommendations are presented to harmonize the seismic protection ofductile and nominally ductile reinforced concrete frames in Canada.

Key words: moment resisting frames, earthquakes, reinforced concrete, seismic, shake table.

Rsum: Cet article est le premier de deux sur lvaluation du niveau de protection sismique des ossatures noeuds rigidesen bton arm au Canada. Cet article prsente les rsultats dessais sur table vibrante de deux ossatures noeuds rigides, chelle une-demie, en bton arm. Les rsultats exprimentaux sont compars aux prdictions danalyses dynamiquesnon-linaires dans le deuxime article. Chaque ossature tait compose de deux traves et de deux tages ayant cinq mtres delargeur et trois mtres de hauteur. La premire structure consistait en une conception ductile selon les normes Canadiennesactuelles. La deuxime structure incorporait des dtails darmature en vue de lui assurer une ductilit nominale. Deux niveauxdintensit ont t retenus pour lexcitation historique la base. Le premier niveau tait reprsentatif du sisme de calculutilis, alors que les amplitudes furent doubles pour le deuxime niveau dintensit. Lossature ductile sest bien comportedurant les deux essais. Lossature ductilit nominale sest bien comporte pour le sisme de calcul, mais tait prs de la ruine

la suite du deuxime essai. En se basant sur ces rsultats exprimentaux, des recommandations sont prsentes afinduniformiser la protection sismique dossatures ductiles et ductilit nominale en bton arm au Canada.

Mots cls: bton arm, ossatures, sismique, table vibrante, tremblements de terre.

Introduction

The seismic design lateral loads and the level of seismic rein-forcement detailing to be incorporated in a reinforced concretemoment resisting framed structure in Canada depend on itsavailable ductility capacity. In ductile moment resistingframes, the design lateral loads are reduced significantly, buthigh ductility capacity is ensured by strict detailing require-

ments to avoid premature brittle failure modes. For framewith nominal ductility, the design loads are higher, but verlittle seismic reinforcement detailing is required. According tthe seismic design philosophy of the National Building Codof Canada (NBCC 1995), both approaches should offer thsame level of seismic protection against the design earthquakat the construction site.

The determination of the design earthquake at a given sitis associated with a high degree of uncertainty (Heidebrech1995). Attenuation relations for peak ground horizontal acceleration, for example, include aleatory uncertainty characterized by the standard deviation in a log-normal distributionTypical attenuation relations yield standard deviation up to0.30 in a log-normal plane (Heidebrecht 1996). This meanthat 68% of the seismic data used to construct such an attenuation relation are within a factor of 2 of the median value useto define the design earthquake at a site. Considering thesvery large uncertainties associated with the design earthquakeit is questionable that the levels of protection offered by ductiland nominally ductile structures are indeed the same.

The objective of this experimental investigation is to contribute to the evaluation of the level of protection offered by

Received April 10, 1997.Revised manuscript accepted August 25, 1997.

A. Filiatrault.EPICENTRE Research Group, colePolytechnique, Universit de Montral Campus, P.O. Box 6079,Station Centre-Ville, Montreal, QC H3C 3A7, Canada.. Lachapelle.Calculatec Inc., 4455 St-Hubert Street,Montreal, QC H3L 4M3, Canada.P. Lamontagne.Shector Barkacki Shemie & Associates, 1550deMaisonneuve Boulevard W., Montreal, QC H3G 1N2,Canada.

Written discussion of this article is welcomed and will bereceived by the Editor until August 31, 1998 (address insidefront cover).

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ductile and nominally ductile reinforced concrete structures inCanada. For this purpose, shake table tests of two half-scalereinforced concrete moment resisting frames, designed ac-cording to current Canadian standards, were performed on theearthquake simulation facility at cole Polytechnique in Mont-real and are reported in this first paper. A companion paper(Filiatrault et al. 1998) compares these experimental resultswith the predictions of inelastic time-history dynamic analy-ses.

Design and description of test structures

Design procedureThe two test structures considered in this investigation weredesigned, at their reduced scale, according to the provisions ofthe National Building Code of Canada (NBCC 1995) and ofthe Canadian concrete standard (CSA 1994). Each structurewas assumed to be part of the lateral load resisting system of abuilding, with two storeys (each 1.5 m high) and two bays(each 2.5 m wide) located in a seismic zone 4 in Canada (Za=Zv=4). The geometry of the structures was chosen to allowthe simultaneous seismic performance evaluation of interiorand exterior beam-column assemblies. The various assump-tions and parameters used in the design of the two structuresare listed in Table 1. The design base shear,V, specified by theNBCC for each structure is given by

[1] V=VeU

R

whereVeis the required base shear if the structure would re-main elastic under the design earthquake, U is a calibrationfactor equal to 0.6, and R is a force reduction factor which

depends on the ductility capacity of the lateral load resistingsystem. For the ductile structure,R is equal to 4; and for thstructure with nominal ductility,Ris assigned a value of 2.

The use ofR = 4 for the ductile structure was justified bimplementing the strict seismic detailing requirements contained in the Canadian concrete standard (CSA 1994). Thstructure with nominal ductility (R = 2), on the other handincorporated only nominal detailing, according also to the Canadian concrete standard, since its design lateral loads werhigher than the ductile structure and, according to the seismidesign philosophy of the NBCC, the ductility demand by thdesign earthquake should be limited.

Description of the test structuresThe two reinforced concrete moment resisting frames considered in the shake table investigation are shown in Fig. 1. Foboth frames, all longitudinal reinforcement was made of continuous standard 10M bars (11.1 mm diameter). Undeformesteel wires, 3 and 6 mm in diameter, were used as transversereinforcement. A concrete clear cover of 15 mm was incorporated in all members. Table 2 presents the material propertieobtained from tensile tests on the reinforcing steel and fromcompressive tests on concrete cylinders.

The dimensions and the longitudinal reinforcement of thbeams are similar for both structures, as they were designefor the same gravity loads (Lachapelle 1997; Lamontagn1997). The column sizes and longitudinal reinforcement, however, are very different for both structures. The flexurastrength of the columns for the ductile frame (R=4) is baseon the flexural capacity of the associated framing beams according to the weak beams strong columns design philosophy adopted by the Canadian concrete standard (CSA 1994)

Parameters Assumptions

Service loads Dead load on first floor=3.2 kPaDead load on roof=1.7 kPaWeight of structural members=24 kN/m3

Live load on first floor =2.4 kPa

Snow load on roof=2.3 kPaWind pressure neglectedSeismic zone:Za= Zv=4Seismic loads according to eq. [1] with importance and foundation factors I= F=1

for both structures andR =4 for the ductile structure andR=2 for the structurewith nominal ductility

Material properties used in design Concrete: compressive strength,fc =25 MPaLongitudinal reinforcing steel: yield strength,Fy=400 MPaTransverse reinforcing steel: yield strength, Fy=700 MPa

Design base shear V=19 kN (R=2)V=9 kN (R=4)

Design and analysis assumptions Tributary width=3 mGravity loads applied as concentrated loads from transverse joists at the 1/3 and 2/3

spans of the beams and at the beam-column joints

Rigid links incorporated for computing internal forces at the columns faces

Rigid links removed for computing lateral deflections40% of gross inertia used for the beams and 70% of gross inertia used for the columns

Table 1.Design assumptions and parameters.

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The columns of the frame with nominal ductility (R=2) weredesigned only to resist the factored loads. The transverse rein-forcement is also quite different for each structure. The ductile

structure incorporates full seismic details, composed of rectan-gular hoops, with 135 hooks, spaced at 30 mm on centre incritical locations of the beams, columns, and joints. The spac-ing of the hoops in the structure with nominal ductility is larger(65 mm on centre in the columns and 60 mm in the beams)except in the beams near the column faces where the spacingis reduced to 30 mm, according to the provisions of the Cana-dian concrete standard (CSA 1994).

One important aspect of the design of beam-column jointsis the development length of the longitudinal reinforcementrequired to ensure plastic hinges in the beams at the columnfaces. This aspect is particularly important for interior beam-

column joints where plastic hinges can develop in oppositdirections on each side of the columns. The longitudinal reinforcement is therefore required to develop simultaneously it

probable tensile strength on one side of the joint and its probable compressive strength on the other side of the joint. Tensure sufficient anchorage, the Canadian concrete standar(CSA 1994) limits the diameter of the longitudinal reinforcement,db, across an interior beam-column joint of length ljafollows:

[2]db

lj

20

dblj

24

forR=2

forR=4

Since the standard 10M bars used as longitudinal reinforce

Fig. 1.Test structures: (a) structure with nominal ductility (R=2) and (b) ductile structure (R=4). (All dimension are in millimetres.)

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ment in the test structures did not satisfy the requirements of[2], sleeves were installed in the interior beam-column jointsas shown in Fig. 2. Each sleeve was made of a hollow 20M barin which the longitudinal 10M bars were inserted and plug-welded on each side. This larger diameter was chosensuch thatthe same average bond stress of 10 MPa between the concreteand the steel underlying [2] would develop when the 10M barsreach their probable tensile and compressive strengths(Lachapelle 1997; Lamontagne 1997). Filler rings were in-serted around the 10M bars at both ends of each sleeve to avoidbearing between the concrete and the sleeve. Preliminary ten-sile tests of the sleevebar assembly confirmed the sufficientstrength of the welds. The anchorage of the longitudinal rein-forcement in the exterior beam-column joints was achieved byconcrete end blocks and 180 hooks as shown in Fig. 2.

Shake table test program

Experimental setupThe seismic tests were performedon theearthquake simulationfacility at cole Polytechnique in Montreal (Filiatrault et al.1996). Figure 3 shows a photograph of the structure withnominal ductility (R=2) ready for testing on the shake table.The frame was secured to the earthquake simulator by a500 mm deep foundation beam anchored to the top plate of theshake table. This foundation beam, in which thecolumns of theframes were anchored, was designed to remain within the ten-sile strength of concrete under the full flexural and shear

strength of the base columns of the structure. Note that thisfoundation beam (5 m) is longer than the shake table (3.4 m)such that the structure overhangs on both sides.

The mass of the structure was provided by four concreteblocks attached at the 1/3 and 2/3 points of the span of eachbeam to simulate concentrated gravity loads from framingjoists. The geometry of each block was an inverted U-shapesuch that its centre of mass could coincide with the centre ofgravity of the beam. In this way, overturning effects above thefloor levels could be eliminated. The total weight of eachframe, excluding thefoundation beam, was 95 kN. This weightcorresponds to 100% of the design dead load. Because of pay-load limitations, no live load could be added. To avoid out-of-

plane motion of the structures during the seismic tests, a rigilateral steel bracing system was installed on the shake tablaround each floor. This system incorporated vertical columnshorizontal beams, and horizontal bracing members. Rollebearings were installed between each concrete block and eachorizontal steel beam to ensure a frictionless lateral motion othe structure.

InstrumentationA variety of instruments were used to monitor the behaviour othe two structures during the seismic tests. Displacementransducers were used to measure the absolute horizontal displacements of the shake table and of the floor levels. The displacements of the floors, relative to the shake table, could thebe obtained by subtracting the shake table displacement fromthe floor displacements. The vertical deflections of the firsfloor beams, under the loading points of the concrete blockswere also measured using displacement transducers. Accelerometers were used to record the shake table and floor levelabsolute horizontal acceleration time-histories. A total of 1strain gauges were installed on the longitudinal reinforcemenof the beams and columns, near the beam-column joints, anat the base of the first floor columns, where severe inelastideformations were expected to occur.

Five video cameras were used to record the tests. One camera was located outside the shake table to record the generabehaviour of the structures. Four other cameras were installeon the shake table to record the relative motion of the interioand exterior beam-column joints and base columns. The cracking pattern was marked and photographed after each test.

Selection of earthquake ground motionThe ensemble of historical earthquake accelerograms considered by Filiatrault et al. (1994) for the city of Vancouver (Z

= Zv = 4) was used as input for preliminary linear spectraanalyses (Clough and Penzien 1993) of both structures. Baseon these analyses, the N04W component of the accelerogramrecorded in Olympia, Washington, during the April 13, 1949Western Washington earthquake was chosen as the base motion input for the seismic tests. Figure 4 presents the acceleration time-history of this seismic event. The peak horizontaacceleration (PHA) is 0.16gand the peak horizontal velocity(PHV) is 0.21 m/s. The strong motion duration is around 30 with a total duration of 89 s. The accelerogram was scaled to PHA value of 0.21gfor the first test (intensity 1) and to 0.42for the second test (intensity 2).

Experimental results

Preliminary system identification testsThe dynamic characteristics of the model frames were estimated from impact tests and from free vibration tests. For thimpact tests, each frame was excited manually by repetitivhorizontal hammerings at the top floor. A dedicated ambienvibration analysis software (Experimental Dynamic Investigations 1993) was used to determine the natural periods of thestructure from power spectral density plots of the absolutfloor horizontal displacement records. In the free-vibrationtests, the structure was excited manually at its first naturaperiod. When a steady-state response was obtained, the inpu

Material Properties

Longitudinal reinforcing Youngs modulus,E =224 600 MPasteel Yield strength,Fy=438 MPa

Yield strain, ey=0.00195Tensile strength,Fu=601 MPa

Ultimate strain, eu=0.199Transverse reinforcing steel Yield strength,Fy=750 MPa

Tensile strength,Fu=900 MPa

Concrete for structure withnominal ductility (R=2)

Youngs modulus,E =25 200 MPaCompressive strength,fc =31 MPaPoissons ratio, = 0.17

Concrete for ductilestructure (R=4)

Youngs modulus,E =22 800 MPaCompressive strength,fc =26 MPaPoissons ratio, = 0.16

Table 2.Actual material properties of test structures.

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was suddenly stopped and the floor relative displacementswere recorded. The first modal damping ratio of the structurewas then established by the logarithmic relative displacementdecrement at each floor (Clough and Penzien 1993).

The fundamental periods were measured at 0.36 and 0.28 for theR = 2 and theR = 4 frame, respectively. The fundamental period of the structure with nominal ductility (R= 2) corresponds approximately to the period of two-storey reinforce

Fig. 2.Details of beam-column joints of test structures.

Fig. 3.Structure with nominal ductility (R=2) on shake table.


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concrete moment resisting frames as specified in the 1995 edi-tion of the NBCC (0.36 s for storey height of 4 m). The appliedloading in the tests was, therefore, representative of actual fieldconditions. The fundamental period of the ductile structure (R=4) was shorter because of the increased lateral stiffness causedby larger columns required to satisfy the weak beams strongcolumns design philosophy. First modal damping ratios of3.3% and 3.0% were measured for the R =2 and the R = 4frame, respectively. These values are typical for reinforcedconcrete framed building structures.

Shake table performanceOne critical aspect of shake table testing is the ability of theelectronic control system to reproduce accurately the desiredground motion input. Figure 5 compares the absolute accelera-

tion response spectra, at 5% damping, of the accelerogram ofFig. 4 (desired signal) with the response spectra of the accel-eration time-histories recorded on the shake table (feedbacksignal). These results were obtained from the tests on the R =2 structure (intensities 1 and 2).

As discussed later, the natural period of the test framesvaried between 0.28 and 0.76 s at various stages of testing. Forthis period range, the mean difference between the desired andthe feedback spectral values is 8.5% for the first intensity, and2.8% for the second intensity. For the same period range, themaximum absolute difference is 16.6%. Considering the se-vere inelastic response of the structures during the seismictests, the shake table performance can be considered adequate.

General behaviour and cracking patternsFigure 6 illustrates the cracking patterns for both structuresafter each test. The ductile structure (R= 4) behaved, duringboth tests, according to the capacity design philosophy pro-moted by the Canadian concrete standard. The structure hadreserve ductility at the end of the second test. Plastic hingesand flexural cracking occurred in the beams, near the columnfaces, and at the base of the first floor columns. No crack wasobserved in the columns below and above the beam-columnjoints. During the first test (intensity 1), column cracking wasconcentrated at the base of the central column. This columnwas stiffer than the exterior columns and attracted most of the

bending moments until a full plastic hinge developed. Duringthe second test (intensity 2), flexural cracking migrated to thbase of the exterior columns. Diagonal cracks occurred in ondirection only in the exterior beam-column joints, suggestingthat the negative dead load end moments were not overcomby the positive end moments induced by the seismic loads.

The structure with nominal ductility (R= 2) behaved welduring the first test. Flexural cracking occurred mainly in thbeams near the column faces and at the base of the three columns. During the second test, however, large horizontal crackoccurred at the top of all first floor columns underneath thbeam-column joints. This cracking pattern suggestedthat a fuplastic column-sway mechanism had formed in the first storeythereby compromising the lateral stability of the structure. Ithe earthquake accelerogram had a longer strong motion dura

tion or a higher intensity, the structure would have eventuallycollapsed.The flexural cracking pattern observed on each structure i

localized within half the effective depth of the members at thends of the beams and the columns. The distances on whicthe seismic confinement detailing was incorporated (seFig. 1) could have been reduced without adverse effects on thseismic behaviour of the test structures. In particular, it waunnecessary to confine the full height of the first floor columnof the ductile structure.

Strain ductility in the longitudinal reinforcementFigure 7 presents the distribution of strain ductility in the longitudinal reinforcing bars instrumented by strain gauges. For

given cross section, the maximum strain ductility,s, was obtained by

[3] s=emax

ey

where emaxis the maximum recorded tensile strain in all barinstrumented at a given cross section and eyis the measureyield strain of the longitudinal steel, as given in Table 2.

For the first intensity, the distribution of strain ductility isimilar for both structures. Only the longitudinal reinforcement in the beams, near the column faces, and at the base othecolumns suffered inelastic tensile strains. Thecolumns lon

Fig. 4.Accelerogram for the N04W component of the ground motion recorded in Olympia,Washington, during the April 13, 1949, WesternWashington earthquake.

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gitudinal bars, below the first floor beam-column joints, re-mained in the elastic range of the steel.

For the second intensity, however, the distribution of strainductility is very different for the two frames. For the ductilestructure (R=4), the cross sections exhibiting inelastic behav-iour were the same as for the first intensity. Only the maximumstrains increased in each cross section. For the structure withnominal ductility (R = 2), the longitudinal bars in the first floorcolumns, below the beam-column joints, experienced signifi-cant inelastic tensile strains. Again, it is clear that a completecolumn-sway mechanism had formed in the first floor.

Peak response parameters

Table 3 shows the peak response parameters recorded duringthe seismic tests on the two structures. The National BuildingCode of Canada (NBCC 1995) limits the inelastic inter-storeydrift at 2% of the storey height for a building of normal impor-tance. The ductile structure (R = 4) meets this requirementunder the design level earthquake (intensity 1) and slightlyexceeds it during the second test at twice the ground motionamplitudes. This reserve stiffness exhibited by the ductilestructure is due to the application of the capacity design con-cept, which causes the columns to increase in size and stiffnessto meet the weak beams strong columns design philosophy.

The structure with nominal ductility also respects the inter-storey drift requirement for the first test. During the second

test, however, the formation of a plastic column-sway mechanism in the first floor caused a very large inter-storey drif(4.67%) in the first floor. This level of deformations would bdetrimental to nonstructural and architectural elements of real building.

The floor acceleration amplification for both structures varies between 1.23 and 4.30, which is typical of lightly dampesystems under ground motion excitations. The amplificationduring the second test are less than those during the first oneThis result is expected, as severe inelastic deformations in thstructures during the second test cause the structural dampinto increase. Also, amplifications exhibited by the frame witnominal ductility are less than the amplifications in the ductil

structure. The nominally ductile frame experiences more severe inelastic deformations than the ductile frame, thereby increasing the damping and limiting the acceleration amplification

Although the ductile structure was designed for a level olateral loads much smaller than thestructurewith nominal ductility, it developed larger base shear coefficients. This phenomenon is, again, a result of capacity design. Even if smalleseismic lateral loads were used for the design of the ductilstructure, the strength of the columns far exceeds the effectof the factored loads to respect the weak beams strong columns design philosophy. As a result, the lateral strength developed by the ductile structure is larger than that by thstructure with nominal ductility.

Fig. 5.Absolute acceleration response spectra, at 5% damping, of the reference and feedback shake table acceleration time-histories from thetests on theR=2 structure.


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The vertical deflection ratios experienced by the first floorbeams are similar for both structures. During the first test, thevertical deflections of the beams are well below the service-ability limit state considered by the Canadian concrete stand-ard for building floors (L/360, whereLis the clear span of thebeam). For the second tests, however, this limit is exceeded.Current Canadian seismic provisions do not address the verti-cal deflections of beams during an earthquake. This aspectshould be considered, particularly if positive plastic hinging

occur along the beam spans causing a shake down phenomenon.

Lateral stiffness degradationThe system identification tests described earlier were repeateafter each seismic test to assess the variation of the dynamicharacteristics of the frames. Table 4 presents the variations othe fundamental period, first mode damping ratio, and laterastiffness ratio for each structure at various stages of testing.

Fig. 6.Cracking patterns of test structures after seismic tests.

Fig. 7.Distribution of strain ductility in longitudinal reinforcement.

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The results are similar for both structures. After the firsttest, the fundamental period of each structure has increased bya factor of 1.5. The corresponding decrease in lateral stiffnessis more than 50%. At the end of the second test, the fundamen-tal period of each structure had more than doubled, therebyreducing the lateral stiffness by more than 75%. These resultsare significant when considering the behaviour of real build-ings to aftershocks or future earthquakes.

The first modal damping ratio varies between 3.0% and4.8% for both structures. These values are typical of reinforcedconcrete structures at levels of excitation causing yielding inthe reinforcing steel.

Hysteretic behaviourThe global hysteretic behaviour of each structure, in terms ofbase shear top floor relative displacement loops, is presentedin Fig. 8. The experimental base shear time-histories were ob-tained by the summation of the inertia forces at each floor.Therefore, the effects of damping forces were neglected in thebase shear calculation.

During the first test (intensity 1), two different effectivelateral stiffness values can be distinguished from the slope ofthe hysteresis loops. The first slope corresponds to the stiffnessof the structures when cracking and some modest inelasticdeformations have occurred. The second slope, exhibiting thelargest inelastic excursions, corresponds to the inelastic shear

deformations of the beam-column joints coupled with largeflexural inelastic deformations. These deformations of thejoints are accompanied by the diagonal cracking described ear-lier. This phenomenon can be observed for both structures, butis particularly significant for the structure with nominal duc-tility (R = 2), for which the joint regions are not as confinedby the transverse reinforcement as the joints in the ductilestructure (R=4).

For the second test, the hysteresis loops of the ductile struc-ture are stable with no significant pinching. Pinching can beobserved, however, for the structure with nominal ductility.

Also indicated in Fig. 8, are the maximum displacementductility factors,, achieved by each structure. The yield dis-

placements used to estimate the ductility factors were obtainefrom static pushover analyses, as discussed in the companionpaper (Filiatrault et al. 1998). The ductility factor exhibited bthe ductile structure during the first test (2.21) is much lesthan the force reduction factor (R = 4) used for its designAgain, this can be explained by the fact that the frame wamuch stronger than required by the factored loads to meet thecapacity design philosophy. This increased strength caused reserve ductility. Even under the intensity 2 earthquake, thductile structure exhibited a ductility factor less than 4.

During the first test, the ductility factor for the structurwith nominal ductility was practically equal to the force reduction factor used for its design (R= 2), and was twice as largfor the second test. This result is, again, consistent, since thlateral strength of the structure with nominal ductility was established from the factored loads only.

Conclusions and recommendations

The shake table tests described in this investigation has provided an opportunity to contribute toward a better understanding of the seismic behaviour of ductile and nominallyductile reinforced concrete moment resisting framed structuredesigned according to current Canadian standards. Based onthe experimental results obtained, the following conclusionand recommendations are presented to harmonize the level oseismic protection offered by these two lateral load resistingsystems.

1. The ductile structure (R=4) performed very well durinthe tests, showing that the capacity design philosophy, as applied in current Canadian standards, is effective. Even for seismic event at twice the intensity of the design earthquakethe ductile structure exhibited a reserve ductility capacity.

2. The structure with nominal ductility (R= 2) performeas expected under the ground motion intensity correspondingto the design earthquake at the site. Inelastic deformation

were concentrated mainly in the beams, near the column facesand at the base of the columns. For the second test at twice thintensity, however, a plastic column-sway mechanism occurred in the first floor bringing the structure to the verge ocollapse. Considering the uncertainty related to the desigearthquake, the level of seismic protection offered by this typof structure can be questioned. Only the incorporation of thweak beams strong columns design philosophy could ensura proper hierarchy of plastic hinging in structures with nominaductility.

3. The plastic hinges were localized at the end of the members and within 1/2 of their depth. This result suggests that thdistances on which the confinement reinforcement is require

Fundamentalperiod (s)

First modedamping ratio

Lateralstiffness rati

Testing stage R=2 R=4 R=2 R=4 R=2 R=4

Before intensity 1 0.36 0.28 0.033 0.030 1.00 1.00

After intensity 1 0.55 0.44 0.037 0.039 0.43 0.40After intensity 2 0.76 0.55 0.042 0.048 0.22 0.26

Table 4.Variations of the dynamic characteristics of the teststructures.

Peak value

Intensity 1 Intensity 2

Parameter R=2 R=4 R=2 R=4

First floor inter-storey driftratio* (%)

1.98 1.58 4.67 2.74

Second floor inter-storeydrift ratio (%)

1.28 1.44 1.75 2.23

First floor accelerationamplification

1.50 2.19 1.23 1.55

Second floor accelerationamplification

2.63 4.30 1.60 2.45

Base shear coefficient 0.39 0.55 0.47 0.64First floor beam vertical

deflection ratio691 694 272 284

*Inter-storey drift over the storey height times 100.Floor acceleration over peak table acceleration.Base shear over weight of the structure (excluding foundation beam).Span of first floor beams (2500 mm) over peak vertical deflection.

Table 3.Peak response parameters recorded.


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could be reduced. Confining the full height of the first floorcolumns could be unnecessary.

4. The beams experienced significant vertical deflectionsduring the tests. Current Canadian seismic provisions shouldaddress the vertical deflections of beams during earthquakes.Positive plastic hinging can occur along the beam spans andtrigger a shake down phenomenon.

5. The lateral stiffness of each structure had reduced bymore than 75% at the end of the seismic tests. Procedures to

quickly estimate the residual stiffness of buildings after thmain shock of an earthquake are urgently needed to predictheir responses to aftershocks or future earthquakes.

Acknowledgements

The authors acknowledge the assistance of the Natural Sciences and Engineering Research Council of Canada (NSERCand the Fonds pour la formation de chercheurs et laide l

Fig. 8.Global hysteretic behaviour of test structures.

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recherche (FCAR) of Quebec which provided research grantsand a scholarship in support of this project. The assistance ofDr. Marc Savard of the Quebec Ministry of Transportation,which supplied a data acquisition system, is also gratefullyacknowledged. The authors wish to express their appreciationto Anne Blanger, Marie-Claude Janelle, NickLaganire, Luc-Andr Taillon, Sylvain Bdard, and Philippe Morin, all under-

graduate research assistants, for their contributions to theexperimental part of the project. Finally, the technical staff ofthe Structures Laboratory at cole Polytechnique is sincerelyacknowledged for its invaluable assistance.

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Experimental Dynamic Investigations. 1993. U2 & V2 manual. Van-couver, B.C.

Filiatrault, A., DAronco, D., and Tinawi, R. 1994. Seismic sheardemand of ductile cantilever walls: a Canadian code perspective.

Canadian Journal of Civil Engineering,21(3): 363376.Filiatrault, A., Tremblay, R., Thoen, B.K., and Rood, J. 1996. A sec-

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Lamontagne, P. 1997. tude du comportement sismique dune ossature ductile (R=4). M.A.Sc. thesis, Department of Civil Engineering, cole Polytechnique, Universit de Montral CampusMontreal, Que.

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