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Load Testing and Ratings of Salvaged Precast Prestressed Concrete Channel Bridge Girders

Mike Lau, P.Eng., Ph.D., Dillon Consulting Ltd. Asnee Pochanart, Ph.D., P.Eng., AECOM Canada Limited Ruth Eden, P.Eng., M.Sc., Manitoba Infrastructure and Transportation Paper prepared for presentation at the Bridges – Adjusting to New Realities (A) Session of the 2010 Annual Conference of the Transportation Association of Canada Halifax, Nova Scotia

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ABSTRACT Six representative precast prestressed concrete channel bridge girders, salvaged from the previous CN Rail overpass on the Portage La Prairie By-Pass, section of the Trans-Canada Highway, were tested to failure. Three girders were tested under single point loading at mid-span (flexural tests) and the other three girders were tested under single point loading at quarter point (shear-flexural tests). The conditions of girders were assessed prior to the tests. In addition, core samples were taken from the girders to determine their compressive strengths, chloride ion contents and air void contents. All six girders exhibited higher load carrying capacities than initially predicted using sectional analyses. After examining all the test measurements including strains and deflections, we concluded that the observed higher load carrying capacities were a result of load spreading and the ability of sections adjacent to the loaded sections to develop plastic deformation prior to the collapse of the loaded girders. However, we do not expect the same higher loading capacity, observed under the monotonic single point loading, to be realized under the real bridge loadings which are multiple point and repetitive. All the girders were load rated in accordance with the AASHTO Allowable Stress Rating (ASR), Load Factor Rating (LFR) and the Load Resistance Factor Rating (LRFR) methods. All six girders failed the ASR and the LRFR service limit ratings under the standard AASHTO vehicles. All six girders failed the ratings at inventory level under HL-93 loading, HSS 30 and CL-625 design trucks, several RTAC vehicles and CS-1 to CS-3 rating vehicles at the strength limit states. All girders with poor condition also failed at the operating rating level. Based on our review of the methods and comparisons with the observed test results, we recommend the LRFR method to be used for future girder load ratings. INTRODUCTION As a part of the reconstruction of the CN Overpass on the Portage La Prairie By-Pass project, AECOM was retained by Manitoba Infrastructure and Transportation (MIT) to conduct a girder load testing and rating study (1). The objective of this testing program was to determine the load carrying capacities of the deteriorated girders after 40 years of service. Six representative girders, salvaged from the previous overpass, were load tested to failure in the W.R. McQuade Structure Laboratory at the University of Manitoba (2). We analyzed all the data collected from the testing program and then compared them to the theoretical results. We also performed load ratings on the girders in accordance with the AASHTO ASR, LFR and LRFR methods (3, 4, 5). GIRDER GEOMETRIC PROPERTIES AND CONDITIONS Cast in 1965, the precast prestressed concrete channel girders were 864 mm deep, 1219 mm wide and 19810 mm long, as shown in Fig. 1. The girder specified 28-day concrete compressive strength was 37.9 MPa. Each girder was prestressed with twenty 13-mm diameter straight strands and fourteen 13-mm diameter singly draped strands. All strands were seven-wire strands with a tensile strength of 1724 MPa. The initial force in each strand was 112 kN. The girder

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shear reinforcement consisted of inverted U-shaped 13-mm diameter bars, spaced at 381 mm on center throughout the length of the girder. The flange was reinforced with five longitudinal 13-mm diameter bars and 13-mm diameter U-shaped top bars at 381 mm on centre and 13-mm diameter bottom bars at 191 mm on center transversely.

Fig. 1: Plan and section views Of the six girders selected for load testing, two girders (G1 and G4) were in good condition, two girders (G2 and G6) were in fair condition and two girders (G3 and G5) were in poor condition. The relative condition categorization of good, fair and poor are as described below:

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• Good - no horizontal cracking at the regions of pre-stressing • Fair - some horizontal cracking in pre-stressing regions but no significant concrete cover

loss • Poor - web sectional loss had occurred and there was noteworthy strand deterioration Two cores samples, one from the flange and the other one from the web, were extracted from each girder to determine the concrete compressive strength. The obtained compressive strengths varied over a large range, i.e. from 37 MPa to 84 MPa, and had an average strength of 63 MPa. Following the procedure outlined in Annex A14.1, equivalent material strengths from tests of samples, of the CHBDC S6-06 (6), the necessary statistical parameters were calculated and shown in Table 1.

Table 1. Concrete strength statistical parameters Mean Strength,

'cf , (MPa)

Standard Deviation

(MPa)

Coefficient of Variation, V

N (# of Samples)

Modification Factor, kc

62.9 12.7 0.202 12 1.06

Using the CHBDC equation shown below, the corrected compressive strength was 51.3 MPa.

]0.0015]n

(kcV)1.28[[1f9.0cf' 0.52

'c +−=

To calculate the strength in accordance with the AASHTO LRFR, we subtracted 1.65 times the standard deviation from the mean strength, yielding the corrected strength of 41.9 MPa. Two cores samples, one from the flange and the other one from the web, were extracted from each girder and used to determine the chloride ion content in concrete. The results of the chloride ion content test did not reveal a significant amount of chloride present in the flange or the web of the six girders tested. Two cores samples, one from the flange and the other one from the web, were obtained from each girder to determine the air void ratio and spacing factor in the concrete. The average air void content of 2.8 percent was below the minimum required air void content of 3.0 percent. Many individual tests yielded results well below this minimum. The test results yielded an average spacing factor of 790 µm, well above the acceptable limit of 230 µm. Thus, the girder concrete was susceptible to freeze and thaw damage. TEST SET-UP For each test, the girder was supported at each end with a bearing pad, 300 mm wide. The loading mechanism used had a capacity of 890 kN with 450 mm of stroke. A single point

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loading with a loading area of 300 mm x 600 mm (equivalent to that of a tire contact area) was used for all girder tests. The load magnitude was recorded with the use of a load cell which was fixed to the hydraulic jacking device and connected to the data acquisition system. Fig. 2 illustrates the loading frame in full detail.

Fig. 2: Loading frame Flexural Testing – Instrumentation Nine linear variable displacement transducers (LVDTs) were installed to measure displacement at various points along the girder. These LVDTs were placed at the supports and at every sixth length (3150 mm) along the girder. An additional LVDT was placed at the point of load to

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measure flange deflections relative to the webs. Three electronic strain gauges (one on top and two on the bottom of each web) were placed at every 3150 mm along the girder. The top strain gauges were placed at the transverse center except for those at the mid-span where two strain gauges were used on each side of the loading plate, 150 mm from the girder edge. On each web, a 200 mm Pi gauge (surface mounted displacement transducer) was applied to the exterior side, 50 mm from the bottom. Each small displacement measured by the Pi gauges can then be converted to a strain value. When the gauge encompasses cracked and non-cracked regions, the calculated strain represents the averaged steel strain among the encompassed regions. All nine LVDTs, 16 strain gauges, and two Pi gauges were connected to the data acquisition system during the flexural tests and measurements were recorded at a rate of one reading per second. Flexural Testing - Procedures The girders were subjected to a static monotonic load increment of 50 kN. At each load increment, the load was held to allow for inspection of cracks. Once the load had reached 400 kN, the LVDTs were reset to increase the range of deflection measurement. The load was then increased in steps of 25 kN until failure. After failure had been reached the load was removed and a final inspection was carried out including damage assessment and photography. Shear Testing - Instrumentation Three sets of electronic strain gauge rosettes were installed on the exterior surface of each web directly under the load. The rosettes were located at 115, 345 and 575 mm from the bottom of the web. In addition, a set of Pi gauges (forming a rosette) was installed at the mid height location on each web to confirm the readings given by the electronic strain gauge, as shown in Fig. 3. The Pi gauges should yield reading up to failure whereas most strain gauges no longer function through a crack. The goal of this instrumentation was to record principle strains and determine if at the quart-span the plane cross-section remained plane as the load increased. At the loading location, two strain gauges (one on each side of the load) were also placed on the flange. Five LVDTs were used to measure displacement along the length of the girder. They were situated at the supports, at the mid-span and on both sides of the load. This setting allowed for measurement of bearing pad compaction, span displacement and any possible torsion. The 20 electronic strain gauges, the six Pi gauges, and the five LVDTs were wired to the data acquisition system and measurements were recorded at one reading per second. Shear Testing - Procedure Before deciding to conduct all the shear tests at quart-span, G4 had originally been loaded at distance of 1000 mm from the support face. The girder withstood a load of 750 kN without any observed cracking. Similar to the flexural testing method, the girders were subjected to static monotonic loading in steps of 50 kN. At each step the load was held to allow for inspection of cracks. After 400 kN, no further inspections were made as it was deemed unsafe. The load was then increased in increments of 25 kN until failure was reached. After failure had been reached the load was removed and a final inspection was carried out including damage assessment and photography.

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Fig. 3: Strain gauges and Pi gauges rosette on girder web at loading point TEST RESULTS AND COMPARISON WITH THEORETICAL PREDICTIONS Flexure - Theoretical Predictions We considered the following two girders conditions: • Fully intact girder with no loss in concrete cover and no loss in prestressing steel area. This

condition represented G1 and G2. • Girder with 50 mm loss in bottom concrete cover with two ineffective prestressing strands.

This condition represented G3. For each girder condition, we calculated the mid-span flexural capacity using the following two methods: • AASHTO-LRFD code equations: The equations were based on an empirical estimate of stress

in the prestressing steel as a percentage of its ultimate strength and based on the use of a rectangular concrete stress block. The prestressing steel areas were lumped at their group centroid.

• Moment-Curvature using RESPONSE-2000 Program (7): This program, developed by Professor Collins and Dr. Bentz at the University of Toronto, uses a more realistic representation of the prestressed concrete section. The concrete and prestressing steel constitutive relationships are represented by power curves and Ramberg-Osgood curves

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respectively. In calculating the bending moment, the program also considered the stress in each individual strand separately.

Table 2 summarizes all the predictions made for the flexural testing. It should be noted that the predicted test loads were based on the following assumptions: • Each girder failed as soon as the mid-span bending moment reached its flexural capacity. • Both webs of each channel girder failed simultaneously. • No other failure mechanism took place between the instance of mid-span flexural capacity

being reached and the collapsing of the test girder. With the above assumptions, the magnitude of each predicted test load was estimated based on the relationship shown below.

Table 2. Flexural testing - theoretical predictions

Girders

Flexural Capacity per AASHTO-

LRFD (kN.m)

Flexural Capacity using RESPONSE

2000 (kN.m)

Self-weight Bending Moment (kN.m)

Maximum Predicted Test

Load (kN)

G1 and G2 2690 2970 512 461~520 G3 2512 2757 512 423~475

Flexure - Test Results The failure loads and maximum deflections of the three girders are summarized in Table 3. Fig. 4 shows the loading versus displacement plot of G1. The loading versus displacement plots of G2 and G3 are similar.

Table 3. Flexural testing – test results

Girders Failure Load (kN)

Maximum Mid-Span Deflection (mm)

G1 650 343 G2 626 317 G3 575 403

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Fig. 4: Deflection along girder length for various load increments of G1 The strain gauges in the top flange provided compressive strain readings up to the maximum load while a number of strain gauges (used to measure tensile strains) at the bottom of the webs ceased to perform after the cracking load. Some of the bottom strain gauges continued to provide strain readings up to the maximum load. The readings from these gauges were relatively small, exhibiting the fact that they were located between cracks. Both Pi gauges provided displacement readings (which were later converted to strains) up to the maximum load and so did all the LVDTs. From each set of LVDT readings along the girder span, we calculated the apparent slopes for the six equally spaced regions along the span. Then we calculated the curvatures from the slope changes among the regions. These calculated values represented averaged curvatures from combined cracked and non-cracked sections. Another series of curvatures were calculated from the compressive strain and the tensile strain readings. These curvatures represented those of either cracked or non-cracked sections, depending on the locations of the tensile gauges with respect to cracks. We then plotted all the calculated curvatures versus their corresponding test loads. Figs. 5, 6 and 7 show these mid-span loads versus curvatures plots for G1. The increased rate of curvature changes at the two middle-third sections at higher loads indicate plastic rotations between these two sections and the mid-span section. In these figures, “G” and “PI” denotes curvatures derived from the electronic strain gauge readings and Pi gauge readings respectively and “LVDT” denotes the curvatures derived from deflection measurements. The calculated moment-curvature at mid-span using the RESPONSE-2000 program is also shown in Fig. 6. The predicted moment-curvature at mid-span is in good agreement with the test data. Similar to G1, we used the available readings of G2 and G3 from strain gauges, Pi gauges, and LVDTs to calculate the curvatures at all loading levels. Similar load versus curvature behaviours were observed.

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Fig. 5: Mid-span load versus curvature at 1/3 Point of G1

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Fig.6: Mid-span load versus curvature at mid-span of G1

Fig.7 : Mid-span load versus curvature at 2/3 Points of G1 Failure Modes G1 - Due to safety concerns, the first girder was not taken to failure. However, we believe that

the girder was loaded very close to its ultimate strength based on a comparison to the other girders within this program.

G2 - Web spalling was observed at loads of 300 kN and greater which was expected due to the cracking at the bottom layer of strand reinforcement seen during the condition cataloguing. The mode of failure was a non-ductile explosion of the compressive zone.

G3 - During the test, two strands of the bottom layer of reinforcement ruptured. A considerable amount of web loss due to concrete spalling was also observed. The mode of failure was a non-ductile explosion of the compressive zone. Fig. 8 shows the failure near the mid-span of G3.

Comparison By direct comparisons, it appears that all three girders had load carrying capacities higher than predicted by 20-25%. We do not believe that the girders had sectional flexural capacities significantly higher than predicted. The higher load carrying capacities were due to the fact that all three girders did not collapse under “single mid-span hinge” failure as assumed. It is evident from all the test results that there were several plastic sections developed simultaneously with

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varying degrees of plastic rotations. The region of these plastic sections extended beyond the middle-third portion of each girder.

Fig. 8: Failure near mid-span of G3 To verify the aforementioned conclusion, we made use of the test results. Test results from G1 were selected because they were more complete than those of the other two girders. We simplified the failure mechanism of G1 as consisting of three plastic hinges located at the left third point, at the mid-span, and at the right third point. Plastic rotations were assumed to be concentrated at these three sections. Curvatures at these three sections were taken from Figs. 5, 6 and 7. Then we determined the internal moment at these sections from their moment-curvature relationships from the RESPONSE-2000 program. By equating work done by external load to energy dissipated by plastic rotations, we back-calculated the failure load. The back-calculated failure load agreed with the failure load from the test as demonstrated by the calculations as shown. Back-Calculation of Failure Load from Test Results – Girder 1

Girder Weight: mkN10.98w g =

Intermediate Diaphragm: kN3.99Pd = Maximum Test Load at mid-span: kN650P = Third point (1/3) deflection and plastic rotation at failure: m0.252Δ1 = rad0.013θ1 = Mid-span deflection and plastic rotation at failure: m0.340Δ 2 = 0.048radθ 2 = Third point (2/3) deflection and plastic rotation at failure: m0.278Δ3 = 0.035radθ3 =

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Averaged Deflection: m0.1892Δ

4Δ2ΔΔ

2Δ

31Δ 33211

avg =⎟⎠⎞

⎜⎝⎛ +

+++=

External Work: kJ 262.38ΔPPΔΔP(L)ΔwW 3d21davggext =+++= Sectional Plastic Moment: mkN2690M P1 −= mkN2975M P2 −= P1P3 MM = From Figure 5, the curvature at 1/3 point (at failure) = 0.006 radian/m. From moment curvature relationship, sectional moment was approximately 77% of its full plastic moment value at this curvature. Internal Plastic Work: kJ 262.17θ(1.0)MθMθ0.77MW 3P32P21P1P =++=

Back-calculated test load: kN 649Δ

ΔPΔPLΔwW

2

3d1davggP =−−−

and is in good agreement

with the failure shown in Table 3. We believe that the following factors contributed to the development of the failure mechanism observed during tests: • The fact that the girders were channel beams, i.e. consisting of two webs. • The loading condition was a single point loading, i.e. allowing the load to spread its effects. • The multiple layers bottom reinforcement, i.e. allowing cracks to form at adjacent sections

instead of rapid propagation of the first crack. Shear - Theoretical Predictions We considered the two following girder conditions: • Fully intact girder with no loss in concrete cover and no loss in prestressing steel area. This

condition represented G4 and G6. • Girder with 50 mm loss in bottom concrete cover with two ineffective prestressing strands.

This condition represented G5. For each girder condition, we used Table 5.8.3.4.2-1 from the AASHTO-LRFD Bridge Design Specifications to construct a shear-flexure interaction diagram for the girder at the test section. For this theoretical prediction, the AASHTO shear reduction factor was omitted. Then we calculated the shear forces and bending moments corresponding to the combined effects of dead load and each increment of the test load. As shown in Fig. 9, the intersection between the loading path and the shear-flexure capacity envelop represents the estimated shear and bending moment at the failure load. Each shear-flexural envelop is made up from two lines. One line was constructed from the parameters listed in the AASHTO-LRFD Table 5.8.3.4.1-1. The other line was constructed using AASHTO-LRFD Equation 5.8.5.3-1. This line represents the shear-flexural capacity governed by the failure of the longitudinal reinforcement, i.e. flexural dominant. All the test girders were expected to fail under flexural shear failure.

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Fig. 9: Shear-flexure envelop and estimated failure load at ¼ point – G4 Table 4 summarizes the theoretical prediction for G4 and G5. Failure load for G6 was expected to be between the two.

Table 4. Theoretical Prediction - Shear Testing

Girder Test Load (kN)

Shear Force (kN)

Bending Moment (kN-m)

Long. Strain at Mid-

depth ( )με

Shear Crack Inclination

Angle (deg.) G4 515 463 2033 750-1000 34-37 G5 460 420 1853 1000 34-37

Shear - Test Results The maximum loads and failure modes of the three girders’ shear tests are summarized in Table 5 below.

Table 5. Shear test results

Girder Maximum Load (kN) Failure Mode

G4 840 Shear-flexural G5 706 Shear-flexural G6 816 Shear-flexural

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Interestingly, all the girders had sheared asymmetrically, as shown in Fig.10. We made use of the strain gauge readings from the top flange and the horizontal Pi gauge readings. To arrive at the total strains, an initial (estimated) compression strain of 210 με was added from those readings. We then plot the total strains versus depth at the predicted failure load level and at the actual failure load level. From the plot, we read the stains at the mid-depth of the shear depth, i.e. at 550 mm from bottom of girder, and compared those strains to the predicted longitudinal strains. Fig.11 shows the longitudinal strain versus depth of G5.

Fig. 10. Asymmetric shear-flexural of G4

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Fig. 11: Longitudinal strain versus depth of G5 Comparison By direct comparisons, it appears that all three girders had loading carrying capacities substantially higher than predicted, i.e. around 50%. We believe that the same factors stated in Section “Flexure-Test Results” significantly contributed to these high capacities. From Figure 11, it is evident that the mid-depth strains at the predicted load level were significantly lower than predicted, i.e. almost zero or negative for G5 and around 500 με for G6. The predicted (longitudinal) mid-depth strains were 750~1000 με. When both girders collapsed, these strains far exceeded 1000 με. It should be noted that the actual concrete strength, which was higher than the designed strength used in the prediction, would have been a factor to the higher observed capacity. However, we expect the contribution to be small, i.e. around 5%. This is due to the fact that higher concrete strength has a marginal effect in increasing the flexural capacity and that concrete contribution to shear strength is related to its tensile strength. LOAD RATINGS Load Ratings using Standard AASHTO Vehicles To identify the critical limit state and the critical section, we performed load ratings on the bridge girders using all three methods (ASR, LFR and LRFR). The vehicles used in the rating were the standard AASHTO vehicles stated for each method, that is, HS20-44 for ASR and LFR and HL-93 for LRFR. To get both the upper limit and the lower limit of the girder rating, we considered the two following sets of girders: • Good to Fair Condition (G1, G2, G4, and G6) • Poor Condition (G3 and G5) The material strengths used in the rating were those stated on the as-built drawings. LRFR at Strength Limit and Load Factor Rating (LFR) We conducted flexural-Shear strength ratings for LRFR and LFR in a similar manner. First, we calculated the flexural capacities at tenth points along the girders. To be consistent with the shear strength models, we used the code equations to estimate the flexural capacities. Despite being presented using different notations, the flexural capacity equations from the AASHTO Standard Specifications and from the AASHTO-LRFD are essentially the same. However, the shear strength models are different. The AASHTO Standard 17th Edition still uses the ACI shear equations while the AASHTO-LRFD uses the model based on the Modified Compression Field Theory. The two models yield comparable strengths at regions of high shear with a low bending moment and at regions of high bending moment with low shear. For the regions in between, where both shear and bending moment are moderate, the strengths obtained using the AASHTO

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Standard is conservative when compared to those from the AASHTO-LRFD. We then compared the factored load effects to the flexural-shear capacities on the shear-flexure interaction diagrams and computed the rating factors. The results are as shown Tables 6 and 7. It should be noted that the live load factor for HL-93 in the LRFR was 1.75 and the live load factor for HS20 in the LFR was 2.17.

Table 6. LRFR Strength-I (Inventory) rating - HL93 Girders 0.1L 0.2L 0.3L 0.4L 0.5L

G1, G2, G4, G6 – Good ~ Fair 1.47 1.23 1.01 0.98 1.09 G3, G5 – Poor 1.06 0.85 0.69 0.68 0.77

Table 7. LFR Inventory rating - HS20-44

Girders 0.2L 0.3L 0.4L 0.5L G1, G2, G4, G6 – Good ~ Fair 1.08 1.05 1.08 1.14

G3, G5 – Poor 0.94 0.93 0.98 1.03 Both sets of girders failed the LRFR and the poor condition girders failed the LFR. The critical section for the LRFR was at 0.4L which is typical for a prestressed concrete girder with single-point draping. The critical section under LFR was more toward the support. This is due to the conservative shear strength model (under moderate shear and bending moment) associated with the LFR. LRFR at Service Limit and Allowable Stress Rating (ASR) The allowable stresses for the LRFR Service Limit and for the Allowable Stress Rating are identical. As expected, the allowable stress rating under HS 20 and the service stress rating under HL-93 with a load factor of 0.80 yielded similar results, as shown in Tables 8 and 9. Only the good to fair condition girders were considered in the following ratings. The poor condition girders already exhibited failure under the service limit, i.e. crack, concrete loss, and exposed prestressing strands. The good to fair condition girders failed both the LRFR Service Limit rating and the allowable stress rating. The controlling stress criterion was the tension stress in the concrete.

Table 8. LRFR service-III limit state rating – HL93 Controlling Stress Criteria 0.2L 0.3L 0.4L 0.5L

Concrete Tension 1.37 0.99 0.88 0.91 Concrete Compression 2.46 1.86 1.67 1.72

Concrete Compression at Half Permanent Loads 1.42 1.08 0.97 1.02 Yielding – Prestressing Steel 2.17 1.65 1.50 1.59

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Table 9. Allowable stress rating – HS20 Controlling Stress Criteria 0.2L 0.3L 0.4L 0.5L

Concrete Tension 1.41 1.03 0.92 0.95 Concrete Compression 2.54 1.93 1.74 1.80

Concrete Compression at Half Permanent Loads 1.47 1.13 1.01 1.07 Yielding – Prestressing Steel 2.24 1.72 1.57 1.66

Load Ratings at Critical Section From the works done in the previous section, we feel that the LRFR should be the preferred load rating method. The shear strength model in the LRFR provides a better prediction than that of the LFR when compared to the test results. In this section, we rated both sets of girders at their critical section against highway loads other than the standard AASHTO vehicles. These vehicles were CSA-CL625, CSA CS Trucks, RTAC vehicles, and the modified HSS-30. The HSS-30 is simply one and a half times the HS20-44 with an additional rear axle. Figs. 12 and 13 are shear-flexure interaction diagrams, constructed under LFR and LRFD, with the factored load effects (in term of shear and bending moment) plotted on them. Because the load factor for LFR, i.e. 2.17, was not comparable to the live load factors usually associated with RTAC, CL and CS vehicles, the factored effects for these vehicles were only plotted on Figure 13. While not theoretically correct, the AASHTO-LRFD live load factor of 1.75 was applied to these vehicles. Ideally, the live load factor for each should be statistically determined based on the actual weight and distribution of the truck population.

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Fig. 12: LFR girder rating at critical section

Fig. 13: Girder LRFR rating at critical section Inventory rating factors for each vehicle can be directly read from the diagram. They can then be converted to operating rating factors by direct proportioning. These rating factors are as shown in Table 10.

Table 10. LRFR - Load rating at critical section

Vehicle Good ~ Fair Poor

Inventory Operating Inventory Operating HL-93 0.98 1.27 0.68 0.88

AASHTO Legal (Type 3, 3S2, 3-3) 1.52~1.71 1.97~2.22 1.05~1.20 1.36~1.56 RTAC 1.02~2.07 1.32~2.68 0.71~1.44 0.92~1.87 CL-625 0.92 1.19 0.65 0.84

CS1-CS3 0.93~1.18 1.20~1.53 0.71~1.44 0.92~1.87 HSS 30 0.64 0.83 0.45 0.58

If we were only concerned about the bridge safety, we could ignore the failed rating under the Service Limit and focus on the operating rating number under the Strength Limit. As shown in Table 10, all the girders have low rating number under the HSS 30, which is the standard design load in Manitoba on Trans Canada Highway. The failed operating ratings still hold true even if we accounted for the following factors:

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• The 10% underestimation of flexural capacity using code equations (instead of using Moment-Curvature relationship)

• The possibility of getting 25% higher failure load due to the development of a higher mode of failure as observed during the tests.

It should be noted that the latter factor was unlikely to develop under the real condition where the bridge loading would be multiple points and repetitive, i.e. not monotonic single point. CONCLUSIONS All six girders exhibited higher load carrying capacities than initially predicted using sectional analyses. After examining all the test measurements in terms of strains and deflections, we concluded that the observed higher load carrying capacities were a result of load spreading and the ability of sections adjacent to the loaded sections to develop plastic deformation prior to the collapse of the loaded girders. While this observed phenomenon is not unique to the monotonic single point loading, we do not expect the same higher loading capacity to be realized under the real bridge loadings which are multiple point and repetitive. We performed load ratings on the girders in accordance with the ASR, LFR and the LRFR. Based on our review of the methods and comparisons with the observed test results, our suggestion is to use the LRFR over the ASR and LFR. All six girders failed the ASR and the LRFR Service Limit rating under the standard AASHTO vehicles. For the strength limit, all six girders failed the ratings at inventory level under HL-93 loading as well as other heavier loads, i.e. CL-625, CS1-CS3, several RTAC vehicles, and HSS 30. All girders with poor condition also failed at the operating rating level. Based on our investigations and ratings, we believe that the Manitoba Transportation and Infrastructure made the right decision in decommissioning the previous overpass and replacing it. REFERENCES

1. Lau M. and Pochanart A. (2010), Load Testing and Ratings of Salvaged Precast Prestressed Concrete Channel Bridge Girders Report, Manitoba Infrastructure and Transportation

2. Mills B.J. (2009), An Autopsical Examination of 40-year Old Pretensioned Concrete Channel Girders, University of Manitoba

3. AASHTO (2002), Standard Specifications for Highway Bridges 17th Edition 4. AASHTO (2007), LRFD Bridge Design Specifications 4th Edition 5. AASHTO (2008), Manual for Bridge Evaluation, First Edition 6. CSA (2006), Canadian Highway Bridge Design Code CAN/CSA-S6-06. 7. Bentz E. (2001), RESPONSE-2000 User Manual

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