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G1: Index Properties, Soil G1: Index Properties, Soil Classification and Classification and

ConsolidationConsolidation

Laboratory Manual

CE 2112 : Soil Mechanics

Session 2012/2013

January 2013

Department of Civil and Environmental Engineering National University of Singapore

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Index Properties, Soil Classification and Consolidation [CE 2112: Engineering Geology and Soil Mechanics]

1. Principles of Soil Classification

Rocks in the crust are subject to weathering and erosion (denudation) with time. Soils, the product of denudation, form from transported material or in-situ decomposed rocks on the surface of the earth. Because soil is a natural product (as opposed to concrete and steel, which were man-made), the range of soil types and variety is enormous. Much of geotechnical engineering is founded on the basis that, notwithstanding the wide variety of soils and properties that exists, their engineering behaviour is generally governed by a few important parameters which are the object of geotechnical testing.

One of the basic sub-division of soils is that which classifies soil into sand and gravel (i.e. coarse grained material), silt (i.e. intermediate-sized grain) and clay (fine-grained material). The system which is used to classify sand and gravel into various sub-types is very different from that used to classify silt and clay. This difference stems from the different factors affecting the behaviour of the different soil types. Sand and gravel generally do not possess any chemical on their particle surfaces, and the major factor affecting their behaviour is the particle size distribution. For this reason, sand and gravel are often sub-classify on the basis of their particle size distribution. On the other hand, the behaviour of finer-grained soils such as silt and clay is more significantly affected by surface activity such as cation exchange and diffuse double layer, owing to the much larger surface area:volume ratio. This fact was recognized and used by Atterberg (an agricultural scientist) to develop simple classification tests which can define the activity of fine-grained soil. In this experiment, we will learn about some simple tests which can be used to evaluate the Atterberg limits.

2. Atterberg Limits

2.1 Principles Atterberg found that the activity of a soil can be succinctly defined by measuring three limiting water contents [Water content = mass of water/mass of soil solids]. Starting with a fine-grained soil which is almost fluid (which we generally call a slurry), as the water content is reduced, a point will be reached at which the soil ceases to flow like a liquid and starts to take on characteristics of a solid e.g. having a measurable strength. The water content at this point is known as the liquid limit (LL). Further loss of water beyond this point enhances the solid characteristics further, with the shear strength of the soil increasing and its volume decreasing as water content reduces. Throughout this regime, the soil behaves as a pliable and ductile solid mass, similar in consistency to plasticine. However, if loss of water is continued, another point will be reached at which the soil changes from a pliable, ductile solid to a brittle solid, which tends to crack and break up when remoulded. The water content at this point is known as the plastic limit (PL). Further drying reduces the volume even

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further until the shrinkage limit (SL) is reached. Beyond this limiting water content, the soil is so hard that further drying effectively produces no further volume change. Beyond this point, any loss of water results in the voids being replaced by air. The range of water content between the liquid and plastic limits is known as the plasticity index (PI), that is

PI = LL – PL (1)

While the shrinkage limit is lesser used, the liquid and plastic limits are two of the most commonly measured soil classification parameters for fine-grained soils. Certain basic trends can be easily discerned in the way the liquid and plastic limits changes with soil types and activity. These include the following: 1. Liquid limit increases with activity of soil. A soil which is highly active (or highly plastic) can

adsorb a lot of water onto its surface before there is enough free water to cause it to flow like a liquid.

2. For the same liquid limit, a soil that has larger particles tends to crumble or break up in a brittle fashion more easily as water content reduces, and therefore generally has a higher plastic limit. In other words, given the same liquid limit, a soil with larger particles will also tend to have a smaller plasticity index.

2.2 Methods of Determination In this experiment, you will learn to measure the plastic and liquid limit of a fine-grained soil and classify it according to the BS1377 Soil Classification System. The details of the method are given in the British Standards for laboratory soil testing BS1377. This standard is widely used in Singapore soil testing practice and you should make an effort to familiarize yourself with it. There is a copy of the BS1377 in the Central Library Reference Section as well as in the Geotechnical Laboratory. You should go through the details of the test step-by-step as well as the forms used thoroughly to ensure that you acquire good geotechnical testing practice. 2.2.1 Plastic Limit In geotechnical testing, the plastic limit is defined as the minimum water content at which soil can be deformed plastically. It is obtained by measuring the water content at which the soil can be rolled into a thread 3 mm thick. In the experiment, approximately 20 g of the soil paste is moulded in the hand until it dries sufficiently for slight cracks to appear. This start the soil paste off at a water content which is often higher than the plastic limit (this is quite easily discerned from the strength and ductility of the soil, with some experience). The sample is then divided into two approximately equal (10g) portions and each of these is divided into four sub-samples. One of the sub-samples is taken and rolled into a ball and then it is rolled over a glass sheet or some smooth surface using the open palm of a hand to form a thread of soil (see Figure 1). If the soil can be thinned down to a thread of less than 3mm diameter, the water content is higher than the plastic limit. In this event, the soil is re-formed into a ball; and then re-rolled on the glass sheet. (This in effect reduces moisture content). This process allows the heat from the hands to dry up the soil gradually, while the kneading keeps the moisture content uniform throughout the sample. By alternately kneading (drying) and rolling (testing) the soil, a point will be reached at which cracks will start to appear in the soil thread just as it thins down to about 3mm. The water content of the soil at this point is measured and taken to be plastic limit. At water content < PL, remoulding of specimen will cause soil to crack and crumble. Shear strength of soil at PL ≈ 200 kPa. The same procedure is carried out on the other three sub-samples, and other 10g portions.

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Figure 1. Determination of plastic limit 2.2.2 Liquid Limit The BS1377 recognizes two methods of determining the liquid limit. The first is the cone penetration method, which is also the preferred method. The second method is the Casagrande method. In this course, we will only deal with the cone penetration method, which has been proven to be much more reliable. The principle of this method consists of allowing a cone of apex angle 30° and 80gm weight to penetrate into a fixed volume of soil, which is contained within a standard aluminium cup. The liquid limit is defined as the water content at which the penetrometer penetrates 20mm into the soil when allowed to fall from a position of point contact with the soil surface. The liquid limit is usually conducted on a portion of soil passing 425 µm mesh. The soil is thoroughly mixed with distilled water into a smooth thick paste, and stored in an airtight container for 24 hours to allow for full penetration of the water.At the time of testing, the soil is remixed for 10 minutes to a uniform water content which is less than the liquid limit (again, you can quite easily tell that this is so from practice) and a portion of it is placed in the standard cup [taking care not to trap air within the soil mass]. The soil surface is then levelled off with the top of the cup. Placing the cup on the stand, lower the cone so that it just touches and marks the surface of the soil paste. The dial gauge is then set and reading noted. The cone is released to penetrate the soil paste for exactly 5 seconds and relocked in its new position. A second dial gauge reading in now taken. The difference between the first and second dial readings gives the amount of cone penetration in mm. BS1377 requires each water content to be penetration-tested twice and the penetration compared to ensure that they agree within a specified tolerance. If the agreement is close, the water content of the soil paste is measured. Otherwise the result is discarded and the test is repeated. If the penetration is less than 20mm, more water is then added and the soil is remixed and re-tested. At least four different water contents, with penetrations enveloping the 20mm mark, should be tested to give four points which are then plotted on a graph of water content against penetration. A straight line is then fitted to the four points and the water content at which the penetration is 20mm is then read off the graph.

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Figure 2. Liquid limit determination using cone penetrometer 2.2.3 Experimental Requirements Each group will be given sufficient soil for the conduct of Atterberg limits tests. Determine the Atterberg limits using the procedures outlined above and detailed in BS1377. Following the procedures outlined above and detailed in BS1377, complete the appropriate forms given in these appendices and determine the PL and LL. Note, in particular, that for the determination of LL, the cone penetration test should be performed at least four times using the same sample with gradual increase in water content. The amount of water added shall be chosen so that a range of penetration values of about 15 mm to 25 mm is covered. Using the Atterberg limits obtained from your experiment. Classify the soil according to the BS1377 soil classification system.

3. Compression and Consolidation of Soils

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3.1 Behaviour of Soil Under Compressive Loads

Figure 3. Typical oedometer set-up for 1-dimensional consolidation test.

When a soil is subjected to compressive loads, its volume will decrease, i.e. it will incur compressive strains. It is mainly for this reason that many structures settle when they are loaded. The compressibility of a soil sample refers to its reduction in volume as the compressive load is increased, and is commonly measured by a consolidation cell or oedometer (Figure 3). As shown in this figure, a test specimen in the form of a cylindrical disc, held inside a ring lying between two porous stones, is placed in an open cell of water to which the pore water in the specimen has free access. The upper stone which can move inside the ring with a small clearance is fixed below a metal loading cap through which pressure can be applied to the specimen through a mechanical loading system. The results of a compression test are normally presented in the form of a plot of void ratio e against logarithm (base 10) of the effective vertical normal stress óv’. Figure 4 shows a typical compression plot. Under low levels of óv’, the void ratio changes by only a small amount and these changes are largely recoverable, i.e. the sample will swell back to its original height if the load is removed. This part of the plot is the recompression line and the soil in this state is described as being over-consolidated. However, as óv’ increases beyond a certain value, known as the pre-compression pressure pc’, the void ratio decreases much more rapidly and much of this reduction is irrecoverable. This part of the plot is known as the virgin compression line (VCL) and, in this state, the soil is said to be normally compressed or consolidated. If óv’ is reduced after it has exceeded the pc’, the soil swells back along a different path from that which it has taken previously. The line along which the soil swells is known as the swelling or recompression line (SWL). The slope of VCL (in lg óv’ space) is known as the compression index Cc whilst the slope of the SWL is known as the swelling index Cs. The larger the values of Cc and Cs, the higher is the potential of the soil to compress and swell under loading and unloading, respectively. Hence, these two parameters are important design parameters for the estimation of ground movements.

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Figure 4. Typical compression curve.

The compression plot is obtained by loading the overburden stress on the soil specimen in steps and measuring the settlement that is incurred, via the dial gauge shown in Figure 1. In each incremental, sufficient time must be given to allow the soil skeleton to come into equilibrium under the total stress, i.e. for each pore water pressure to dissipate. This process of gradual dissipation of excess pore pressure is known as consolidation and will be discussed further in a later section. Cc and Cs are determined directly by plotting voids ratio e vs lg σv’. Supposing we have an oedometer specimen with cross-sectional area A and height H. In the course of applying an increment of overburden effective stress ∆σv’ = ∆P/A where ∆P is the load increment, the specimen’s thickness decreases by ∆H. It can be shown that the strain increment

∆ε = -∆H/H = -∆e/(1+e) (2) where e is the void ratio. Thus, if we know the initial void ratio (which can be calculated if we know the unit weight and water content of the soil), ∆e and thereby the new value of e, can be calculated once the settlement ∆H is measured.

pc’

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Figure 5. Determination of pc’ using Casagrande’s method.

In real soils, the transition from the over-consolidated state to normally consolidated state is usually not so clear since the e-log σv’ is commonly a curve. The method which is widely used to determine the pre-consolidation pressure is Casagrande’s empirical method, which is illustrated by Figure 5 and uses the following steps: 1. Locate, by inspection or otherwise, the point of maximum curvature P. 2. Project a straight line backwards from the virgin compression segment of the curve. 3. Construct the tangent to point P (i.e. PR) as well as a horizontal line PQ passing through P. 4. Construct the angular bisector PS to the angle QPR. 5. Locate the point of intersection between PS and CD; the abscissa of this point of intersection

gives the preconsolidation pressure. Another parameter is the coefficient of volume change mv, which can also be deduced directly from the measurement of the settlement since mv is really the compliance of the soil ie. the reciprocal of the constrained stiffness modulus. By definition, mv = ∆ε/∆σv’ (3) Thus mv can be deduced directly from the stress-strain data or as the tangent of the stress-strain curve at the prescribed point. Combining Eqs. 2 and 3 leads to

mv = )e1('

e

v +−

σ∆∆

(4)

Thus, a more common way of determining mv is from the slope of the e-σv’ curve. It has been observed that the compressibility of a soil can be correlated to its LL or PI. Several relations have been proposed. Terzaghi and Peck proposed the following two relations for remoulded and undisturbed soils: Remoulded soil: Cc = 0.007(LL-10%) (5) Undisturbed soil: Cc = 0.009(LL-10%) (6)

P Q

R

S

C

D

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Schofield and Wroth (1968) proposes several semi-empirical relations based on the critical state concept, including the following: Cc = 0.01346PI (7) Cc = 0.02116(PL - 9%) (8) Cc = 0.00828(LL - 9%) (9) Wood also proposes the following semi-empirical relations based on the critical state concept: Cc = 0.05Gs PI (10) in which Gs is the specific gravity of the soil solid, usually taken to be about 2.65. Based on local experience so far, most, if not all, of the above relations appear to give an over-estimation of Cc when compared to consolidation test results but Eq. 5 appears to give the best prediction. All these relations can be used to correlate your Cc values from consolidation tests to your LL and PL values. 3.2 Rate of Consolidation

In many fine-grained soils, compression of the soil under loading is not instantaneous. This is because, for the soil skeleton to compress, pore water has to be expelled. The theory of consolidation will be discussed in detail in your lectures. The objective of the following sections is to give you a basic level of understanding, mainly from a physical, rather than mathematical, viewpoint. Consider the specimen of soil in your oedometer when it is subjected to a sudden increment in stress, such as that applied by you when you increase the weights. The soil specimen consists of two phases, viz. soil skeleton and pore water. You can think of the soil skeleton as behaving somewhat like a sponge; in this case, the “sponge” is completely saturated with water. When you increase in the load on the load hanger, the soil specimen will try to deform by compressing vertically and expanding laterally. But lateral expansion is not possible since the soil specimen is constrained by a stiff metallic ring. In such a condition, the only way for the soil specimen to settle is to undergo a volumetric contraction. Since the specimen is completely saturated, water must be expelled from the soil skeleton (or “sponge”) if the latter is contract volumetrically. If the soil skeleton (or “sponge”) is highly impermeable, water will take a long time to flow out of the soil skeleton. In the short-term, excess pore pressure will be generated which supports the load. The process of gradual time-dependent settlement and pore pressure dissipation of the specimen under loading is known as consolidation. In the classical consolidation process under increase in loading, the excess pore pressure generally decreases whereas the settlement and effective stress increase with time. For low permeability soils, this can take a long time. In such cases, not only is the final compression important, but so too is the rate of consolidation. The parameter which is often used to describe the rate of consolidation is the coefficient of consolidation cv.

In theory, cv can be deduced from the relationship cv = k/(γw mv) (11) However, in practice, this is often difficult since it is often difficult to determine k in the laboratory reliably. Hence, this relationship should be used only as a last resort and also perhaps as a check. The coefficient of consolidation can also be measured using the same oedometer or consolidation test set-up in Fig. 1; indeed this is the normal way of determining cv. In order to determine the

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coefficient of consolidation, settlement readings must taken at various times after load application; this is different from the compression parameters Cc and Cs, which only requires the final settlement to be measured after each loading increment. There are several different methods for determining the coefficient of consolidation. However, probably the most commonly used method is Taylor’s settlement-root-time method and Casagrande’s log-time fitting method, which involves plotting settlement (on the y-axis) against square root time of time (on the x-axis). The procedures for these methods are given below.

3.2.1 Taylor’s √Time Curve-Fitting Method Real consolidation events are often coloured by two other features, namely: (a) Immediate settlement. When the load is first put on, a certain amount of acceleration occurs very

rapidly. This arises from small amounts of air being compressed in the sample and in the spaces between the sample and the end platens. Thus the real settlement curve may contain a portion, which is very steep (almost vertical), representing a rapid increase in settlement (see figure above). To correct, it is usual to project the straight line portion backwards until it intersects the settlement axis at the “corrected” zero. This point is then taken to be real origin of the settlement curve, i.e. the initial origin is disregarded.

(b) Secondary compression. In our consolidation theory, we have assumed that the soil skeleton is linearly elastic and that the time-dependent settlement during consolidation arises entirely from the gradual flow of pore water to the outside of the sample. In reality, it is often not so. In particular, the soil skeleton can exhibit creep behaviour (i.e. compression under a state of constant effective stress) due to the chemical activity of clay particles. This phenomenon is particularly significant in active soils such as marine clay and peat. To distinguish the creep phenomenon of the soil skeleton from the classical consolidation, the former is known as secondary compression/consolidation whilst the latter is often known as primary consolidation. In the consolidation curve, secondary consolidation is manifested in a departure from the classical consolidation curve towards the end of primary consolidation, in such a manner that the actual consolidation curve does not “tail off” (see figure below).

The presence of secondary compression therefore presents problem for the estimation of final settlement. Furthermore, since we know that secondary compression is significant at the tail end of primary consolidation, we should try to avoid making use of the tail end of the consolidation curve to estimate cv. The Taylor’s root-time procedure is as follows (Figure 6): 1. Plot settlement ρ vs √time. 2. Draw a straight line (say OG) that coincides with the initial linear portion of the consolidation

curve. 3. Draw another straight line (say OC) through the corrected origin which has an “inverse” gradient

1.15 times that of line OG.

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4. The point of intersection of line OC and the consolidation curve gives the point (√t90, ρ90). Thus t90 can be obtained by squaring the abscissa of the point of intersection, and cv obtained using the relation cv = 0.848H2/t90.

Figure 6. Taylor’s root-time curve fitting procedure.

3.2.2 Casagrande’s log10(time) Curve Fitting Method It has been observed that secondary compression, like most creep phenomenon, increases linearly with the log10 of time. Figure 7 shows a typical settlement vs lg (time). In this curve, the secondary compression appears as a straight line. The end of primary consolidation U = 1, is now identifiable (approximately) by the break in the slope between the primary phase and the secondary phase, thereby allowing us to find the final settlement ρ∞. The point (lg(t50), 0.5ρ∞) corresponding to 50% consolidation (i.e. when the settlement reaches half of its final value) can be found by locating the point on the consolidation curve with ordinate of 0.5ρ∞. Hence, we can determine t50. We know that at U = 0.5, T = 0.196 = cv t50/H

2. Therefore cv = 0.196H2/t50. Thus, knowing t50, cv can be found.

O

G C

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Figure 7. Casagrande’s log(time) curve fitting procedure. 3.3 Experimental Procedures for Consolidation Test Each group will be given a tube full of sampled soil, which can be used to prepare two specimens for the oedometer tests. The details of the procedure are given in BS1377; the following section contains merely a summary of the procedure for your quick reference. 1. Ensure that the metallic constraining ring for the specimen is clean. Grease the inside walls of the

ring to reduce the friction between soil and ring. Weigh the greased ring and measure its inner diameter.

2. Check that the porous stones have been thoroughly boiled beforehand, to ensure saturation. 3. Extrude the soil sample slowly from the sampling tube until the wax plug is completely exposed.

Use a wire cutter to cut off the wax plug. 5. Clamp the sharp cutting edge of the metallic ring against the exposed end of the sampling. 6. Extrude the inner core of the sample from the tube into the metallic ring until sufficient sample has

gone into the ring to completely fill the latter. Cut of the portion of the sample, which has been intruded into the ring, from the rest of the sample in the tube.

7. Trim the sample within the ring to ensure good contact between the ring and soil and that the top and bottom faces are flat. Use some of the leftover soil for water content measurements.

8. Weigh the sample and ring. 9. Place the bottom platen and porous stone in the oedometer bowl and then the sample-cum-ring

on top of the porous stone. 10. Place another porous stone, followed by the top platen on top of the sample. 11. Fill the oedometer bowl with de-aired water until the sample and porous stones are completely

submerged. 12. Put an appropriate weight on load hanger so as to apply a seating stress of 50kPa onto the

specimen. You need to take note of the load magnification ratio of the lever system in order to work out the applied stress. Simultaneously, start the computer data acquisition system which will automatically log in your settlement data for the next 24 hours.

13. Return the next day to download the data from the data acquisition system and increase the

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overburden stress to 100kPa by adding weights to the loading hanger. Repeat this sequence over the next four days so as to subject the sample to the following overburden stresses:

DAY OVERBURDEN STRESS (kPa) 1 50 2 100 3 200 4 400 5 100 6 Download final readings and wash up.

3.4 Data Processing (1) Determine the initial water content of the sample used in the consolidation test and complete the

attached test header form. (3) Plot the void ratio against effective pressure on a linear scale and determine the coefficient of

volume change mv at different effective pressures. (4) Plot the void ratio against the log10 of the effective pressure and determine Cc, Cs and the pre-

compression pressure pc’. (5) For each load increment, plot the settlement against the logarithm of time and thus determine the

cv values for each increment. (6) By combining cv and mv in each increment, deduce the value of k for each increment. Comment

on the trend of increase or decrease of k with effective pressure. (7) Compare your measured Cc values with the given correlations. (8) Assess the reliability of your results and recommend design values for Cc, Cs, mv, cv and k.

Annex A

BS 1377 : Part 2 : 1990

Liquid limit (cone penetrometer) and plastic limit Form2.C

Location Job ref.Borehole/Pit no.

Soil description Sample no.

Test method BS 1377 : Part 2 : 1990 :4.3/4.4* Date :

PLASTIC LIMIT Test no. 1 2 3 4 5 6 AverageContainer no.Mass of wet soil + container gMass of dry soil + container gMass of container gMass of moisture gMass of dry soil gMoisture content %

LIQUID LIMIT Test no. 1 2 3 4 5 6Initial dial gauge reading mmFinal dial gauge reading mmCone Penetration mmAverage penetration mmContainer no.Mass of wet soil + container gMass of dry soil + container gMass of container gMass of moisture gMass of dry soil gMoisture content %Average moisture content %

Sample preparation *as receivedwashed on 425 um sieveair dried at ...........Deg Coven dried at ........Deg Cnot knownProportioned retainedon 425 um sieve .......% %Liquid limit : %

Plastic limit : %

Plasticity index :

* Delete as appropriateOperator CheckedApproved

Depth

LIquid Limit

0

5

10

15

20

25

30 40 50 60 70 80

Moisture Content %

Co

ne

Pen

etra

tio

n m

m

Annex B

WATER CONTENT / BULK DENSITY

Site : Natural / AfterBore Hole :Sample No : Consolidation / Direct Shear/

Vane Shear/Triaxial Shear /Test

Specimen Test No : Date :

1 2 3

Dia of ring /soil cm

Can + wet soil gm Height of ring /soil cm

Can + dry soil gm Ring + wet soil gm

Can only gm Ring only gm

Water gm Wet soil only gm

Dry soil gm Volume of soil cm3

Water content % Bulk density gm/cc

Visual Description / Remarks :

WATER CONTENT / BULK DENSITY

Site : Natural / AfterBore Hole :Sample No : Consolidation / Direct Shear/

Vane Shear/Triaxial Shear /Test

Specimen Test No : Date :

1 2 3

Dia of ring /soil cm

Can + wet soil gm Height of ring /soil cm

Can + dry soil gm Ring + wet soil gm

Can only gm Ring only gm

Water gm Wet soil only gm

Dry soil gm Volume of soil cm3

Water content % Bulk density gm/cc

Can No.

Can No.

Annex C

CONSOLIDATION TEST SAMPLE 1BS1377 : PART 5 : 1990

Location Job ref.Borehole/Pit no.

Soil description Sample no.DepthDate :

Machine no. Specimen diameter mmCell no. Height of solids Hs mmRing no. Initial voids ratio e0

WhereHeight Ho = 20 mm Cv =He ight of Solids Hs = mm e =Initial Void ratio eo= W x Gs (for fully saturated soil) mv =

= k =

Increment No Pressure P kPa

Cummulative compression mm

Consolidated height Ht mm

Void Ratio e

Incremental Height change dH mm

Pressure Change DP kPa

Mv m2/kN

H mm

Cv m2/yr

k x 10-10 m/s

Consolidation test - calculations

Test method BS 1377 : Part 2 : 1990 :4.3/4.4*Height H0 mm

4

5

Consolidation

0

2

3

1

Void Ratio Compressibility