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TRANSCRIPT
Geotechnical Considerations for Offshore Wind Turbines
Zachary J. Westgate
Jason T. DeJong
August 1, 2005
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TABLE OF CONTENTS PREFACE ....................................................................................................................................................................6 GENERAL ...................................................................................................................................................................6
OBJECTIVES ...............................................................................................................................................................6 SCOPE & LIMITATIONS ..............................................................................................................................................6 RESOURCES ...............................................................................................................................................................7
1.0 THE GLOBAL OFFSHORE WIND INDUSTRY......................................................................................8 1.1 INTRODUCTION............................................................................................................................................8 1.2 EUROPE .......................................................................................................................................................8 1.3 UNITED STATES...........................................................................................................................................9 1.4 OTHER COUNTRIES....................................................................................................................................10 1.5 FUTURE PROSPECTS IN OFFSHORE WIND ENERGY DEVELOPMENT............................................................10
2.0 TYPES OF FOUNDATIONS .....................................................................................................................12 2.1 INTRODUCTION..........................................................................................................................................12 2.2 PILED FOUNDATIONS.................................................................................................................................12
2.2.1 Introduction ........................................................................................................................................12 2.2.2 Monopiles ...........................................................................................................................................12 2.2.3 Multiple-Leg Foundations ..................................................................................................................13 2.2.4 Lattice Towers ....................................................................................................................................13
2.3 GRAVITY BASE FOUNDATIONS..................................................................................................................14 2.3.1 General Description ...........................................................................................................................14 2.3.2 Material Considerations.....................................................................................................................15
2.4 SUCTION CAISSONS ...................................................................................................................................15 2.4.1 General Description ...........................................................................................................................15 2.4.2 Installation Principle..........................................................................................................................16 2.4.3 Site Considerations.............................................................................................................................16
2.5 FLOATING STRUCTURES ............................................................................................................................17 2.5.1 Introduction ........................................................................................................................................17 2.5.2 Tension-Leg Platforms .......................................................................................................................17 2.5.3 Low-roll Floaters................................................................................................................................17
3.0 DESIGN METHODOLOGIES ..................................................................................................................18 3.1 LIMIT STATE DESIGN ................................................................................................................................18
3.1.1 Ultimate Limit State............................................................................................................................18 3.1.2 Fatigue Limit State .............................................................................................................................18 3.1.3 Accidental Limit State.........................................................................................................................18 3.1.4 Serviceability Limit State....................................................................................................................19
3.2 LOAD RESISTANCE FACTOR DESIGN (LRFD) METHOD ............................................................................19 3.3 DIRECT SIMULATION.................................................................................................................................19 3.4 TESTING BASED DESIGN ...........................................................................................................................20 3.5 PROBABILITY BASED DESIGN....................................................................................................................20
4.0 SITE INVESTIGATIONS..........................................................................................................................21 4.1 INTRODUCTION..........................................................................................................................................21 4.2 PHASES OF THE SITE INVESTIGATION ........................................................................................................22
4.2.1 Introduction ........................................................................................................................................22 4.2.2 Geological Study.................................................................................................................................23 4.2.3 Geophysical Survey ............................................................................................................................24 4.2.4 Geotechnical Site Investigation ..........................................................................................................26
4.3 OFFSHORE GEOTECHNICAL SITE INVESTIGATION VESSELS .......................................................................27 4.3.1 Introduction ........................................................................................................................................27
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4.3.2 Low Draft Barges ...............................................................................................................................28 4.3.3 Jack-up Rigs .......................................................................................................................................28 4.3.4 Specialty Geotechnical Drilling Vessels.............................................................................................28 4.3.5 Semi-submersible Drilling Rigs..........................................................................................................29 4.3.6 Supply Vessels ....................................................................................................................................29
4.4 DRILLING ..................................................................................................................................................29 4.5 SAMPLING .................................................................................................................................................31
4.5.1 Introduction ........................................................................................................................................31 4.5.2 Wheeldrive System..............................................................................................................................32 4.5.3 Grab Sampling....................................................................................................................................32 4.5.4 Push and Piston Sampling..................................................................................................................33 4.5.5 Coring Techniques..............................................................................................................................33 4.5.6 Other Sampling Methods ....................................................................................................................35 4.5.7 Sample Recovery, Storage, and Transport .........................................................................................35
4.6 IN SITU TESTING .......................................................................................................................................36 4.6.1 Introduction ........................................................................................................................................36 4.6.2 Cone Penetration Testing ...................................................................................................................36 4.6.3 T-bar Testing ......................................................................................................................................37 4.6.4 Field Vane Testing..............................................................................................................................38 4.6.5 Other In Situ Tests ..............................................................................................................................38
4.7 CHARACTERISTIC SOIL PROPERTIES ..........................................................................................................38 4.8 LABORATORY TESTING .............................................................................................................................40
4.8.1 Offshore Testing Program ..................................................................................................................40 4.8.2 Onshore Testing Program ..................................................................................................................40 4.8.3 Soil Classification...............................................................................................................................41 4.8.4 Index Testing ......................................................................................................................................41 4.8.5 Consolidation Testing.........................................................................................................................42 4.8.6 Strength Testing..................................................................................................................................43 4.8.7 Other Laboratory Testing...................................................................................................................44
4.9 EFFECTS OF CYCLIC LOADING...................................................................................................................46 4.10 SEAFLOOR STABILITY ...............................................................................................................................46
4.10.1 Introduction ........................................................................................................................................46 4.10.2 Slope Stability .....................................................................................................................................47 4.10.3 Hydraulic Stability..............................................................................................................................47 4.10.4 Earthquake Stability ...........................................................................................................................47
4.11 SCOUR.......................................................................................................................................................48 4.11.1 Scour Mechanism ...............................................................................................................................48 4.11.2 Types of Scour ....................................................................................................................................49 4.11.3 Prevention of Scour ............................................................................................................................49 4.11.4 Designing for Scour............................................................................................................................50
5.0 OTHER ENVIRONMENTAL CONDITIONS.........................................................................................52 5.1 INTRODUCTION..........................................................................................................................................52 5.2 METEOROLOGICAL PARAMETERS..............................................................................................................52
5.2.1 Wind Loading .....................................................................................................................................52 5.2.2 Wind Modeling ...................................................................................................................................53 5.2.3 Turbulence..........................................................................................................................................53
5.3 OCEANOGRAPHIC PARAMETERS................................................................................................................54 5.3.1 Wave Loading.....................................................................................................................................54 5.3.2 Wave Modeling...................................................................................................................................55 5.3.3 Current Loading .................................................................................................................................58 5.3.4 Current Modeling ...............................................................................................................................58 5.3.5 Ice Loading.........................................................................................................................................59 5.3.6 Ice Modeling.......................................................................................................................................59
5.4 COMBINED WIND AND WAVE LOADING....................................................................................................60 5.4.1 Horizontal to Moment Load Ratio ......................................................................................................60
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5.4.2 Combination Methods.........................................................................................................................61 5.4.3 Design Considerations .......................................................................................................................61
5.5 ENVIRONMENTAL CORROSION ..................................................................................................................61 5.5.1 Introduction ........................................................................................................................................61 5.5.2 Degradation Corrosion ......................................................................................................................62 5.5.3 Marine Growth ...................................................................................................................................62
5.6 OTHER LOADING CONDITIONS ..................................................................................................................63 5.6.1 Transportation and Installation Loading ...........................................................................................63 5.6.2 Vessel Collision ..................................................................................................................................63 5.6.3 Deformation Loading .........................................................................................................................64
6.0 FOUNDATION MODELING ....................................................................................................................65 6.1 INTRODUCTION..........................................................................................................................................65 6.2 TYPES OF MODELS ....................................................................................................................................66
6.2.1 Plasticity Models ................................................................................................................................66 6.2.2 Finite Element Models........................................................................................................................67 6.2.3 Other Techniques................................................................................................................................68
6.3 DYNAMIC SENSITIVITY..............................................................................................................................69 7.0 FOUNDATION DESIGN ...........................................................................................................................71
7.1 INTRODUCTION..........................................................................................................................................71 7.2 OFFSHORE TURBINE VS. OFFSHORE PLATFORM FOUNDATIONS ................................................................72 7.3 PILED FOUNDATIONS.................................................................................................................................73
7.3.1 Introduction ........................................................................................................................................73 7.3.2 General Design Considerations .........................................................................................................73 7.3.3 Grouting Operations...........................................................................................................................74 7.3.4 ULS Design.........................................................................................................................................75 7.3.5 SLS Design..........................................................................................................................................76
7.4 GRAVITY BASE FOUNDATIONS..................................................................................................................77 7.4.1 General Principles..............................................................................................................................77 7.4.2 General Design Equations..................................................................................................................77 7.4.3 Stability of Foundation: ULS and ALS Design ...................................................................................78 7.4.4 Settlements and Displacements: SLS Considerations.........................................................................79 7.4.5 Grouting Operations...........................................................................................................................79
7.5 SUCTION CAISSONS ...................................................................................................................................80 7.5.1 General Principles..............................................................................................................................80 7.5.2 Design Studies ....................................................................................................................................80
8.0 FOUNDATION TRANSPORT AND INSTALLATION .........................................................................83 8.1 MARINE OPERATIONS................................................................................................................................83 8.2 PROJECT DURATION ..................................................................................................................................83
9.0 COSTS OF WIND ENERGY.....................................................................................................................85 9.1 RENEWABLE ENERGY MARKET INCENTIVES .............................................................................................85 9.2 GENERAL COST CONSIDERATIONS ............................................................................................................85
9.2.1 Introduction ........................................................................................................................................85 9.2.2 Cost of Energy Factors.......................................................................................................................86 9.2.3 Cost Optimization...............................................................................................................................88
9.3 COSTS FOR UNITED STATES OFFSHORE WIND FARM DEVELOPMENT........................................................89 9.3.1 Department of Energy Cost Estimates................................................................................................89 9.3.2 Other Estimates ..................................................................................................................................90
10.0 OFFSHORE WIND DEVELOPMENT IN NEW ENGLAND................................................................91 10.1 INTRODUCTION..........................................................................................................................................91 10.2 ENVIRONMENTAL CONDITIONS AND CONSTRAINTS ..................................................................................91
10.2.1 Wind Speed .........................................................................................................................................91
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10.2.2 Water Depth........................................................................................................................................91 10.2.3 Wave Loading.....................................................................................................................................92 10.2.4 Subsurface Conditions........................................................................................................................92
10.3 OTHER DESIGN CONSIDERATIONS.............................................................................................................92 10.3.1 Land Jurisdiction................................................................................................................................92 10.3.2 Electrical Grid Connections ...............................................................................................................93 10.3.3 Availability of Harbors for Construction and Maintenance...............................................................93 10.3.4 Local Public Acceptance and Conflicting Areas ................................................................................93
10.4 ADVANCEMENT OF DEEP WATER TECHNOLOGY.......................................................................................93 11.0 CONCLUSIONS..........................................................................................................................................95
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Preface
General Offshore wind farms are becoming increasingly popular in the quest for renewable sources of
energy. The planning, design, inspection, and maintenance of offshore wind farms requires
careful consideration of many variables, including local climate and site conditions, economic
incentives, proximity to energy loads, environmental considerations, and legal issues. The
various technologies involved in the design of offshore structures include oceanography,
foundation engineering, structural engineering, marine civil engineering, and naval architecture,
with specific aspects of each technology summarized in Table 1.
The project certification process leading to successful development of an offshore wind farm
includes six phases that incorporate verification of the design basis, the preliminary design, and
the final design, as well as manufacturing surveys, transport and installation, and the final in-
service state. The relationship of the wind farm development tasks to these certification phases
are summarized in Table 2. This report will focus primarily on the site condition assessment and
foundation modeling, design, and to a lesser extent, the installation aspects.
Objectives The objective of this report is fourfold: 1) to provide an updated account of the current state of
the practice of the offshore wind industry and the technological direction in which the industry is
headed, 2) to provide guidance in planning, designing, and constructing an offshore wind farm,
specifically with respect to the site investigation program and foundation options, 3) to provide
insight into the current research developments in offshore wind turbine foundations, and 4) to
provide information useful for the planning and design of offshore wind turbines in the New
England region of the United States.
Scope & Limitations This document is not meant to be a standalone resource for planning or design of offshore wind
turbines. The guidelines included provide information on the activities associated with offshore
site investigations and a design basis for the foundation support elements of the wind turbine
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support structure. This document does not include recommendations for mechanical equipment
design such as the rotor-nacelle assembly, nor does it include aspects of detailed structural
design of the support towers. This document also does not account for local government codes,
legal restrictions, environmental policy, or permitting restrictions.
Resources The referenced material contains additional resources that may specify other planning and design
considerations not accounted for in this document, including industry standards (Table 3) for
both offshore construction in general and offshore wind turbine construction, state of the industry
reports and reviews, papers on both novel research concepts and proven offshore technologies,
and past and present feasibility studies for global and local offshore wind energy development.
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1.0 The Global Offshore Wind Industry 1.1 Introduction
The offshore wind industry has developed rapidly over the past 20 years with many
industrialized countries now beginning to implement renewable energy requirements as part of
their total energy production portfolio. Many of the areas with the greatest potential for offshore
wind energy development are in coastal waters where shallow depths extend relatively far
offshore (Table 4). While Europe has been the industry leader since the initiation of offshore
wind turbine development, some of the other continents planning offshore wind energy
development include Asia and North America, with countries such as China and the United
States taking recent initiative.
1.2 Europe In Europe, the offshore wind energy industry is gaining increased momentum as many countries
are following the lead of the Nordic countries (e.g. Denmark, The Netherlands) and the United
Kingdom. By the end of year 2003, there were 11 operational wind farms in Europe (Table 5),
constituting 600 MW of energy generating potential, with individual projects of up to 160 MW
each, enough to power 145,000 homes (EWEA 2004). Most of the projects have been limited to
less than 30-m deep waters in the North and Baltic Seas, primarily in the 5 to 12-m water depth
range (Musial and Butterfield 2004).
The offshore potential in Europe is considered to be greater than its total electricity consumption
(Beurskens and Jensen 2005). Based on sites within 30-km of shore and a generating density of
6 MW/km2, the offshore wind potential was estimated to be 596 TWh for water depths less than
10-m, and 3028 TWh for water depths less than 40-m (Gaudiosi 1996). Although this could be
overestimated due to political restraints, the potential for further offshore development in deeper
waters (e.g. North Sea) may substantially increase this estimate. By the end of year 2003, the
European Wind Energy Association forecasted that by 2010, up to 10 GW offshore capacity
would be operating in Europe. In the long term, a sea area of 150,000 square kilometers with
water depth less than 35-m could be available, with the generation area capable of powering all
of Europe (EWEA 2004).
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Based on European wind atlas data for wind speed values through 10-km offshore, the United
Kingdom is the leader for this potential, covering one third of the total potential (Table 6). This
year 2004 estimate is a production level equivalent to three times the energy consumption in UK
in 1998 (Sahin 2004). The UK government’s Renewables Obligation established in 2002
mandated that 10% of the UK’s electricity must be generated through renewable sources by year
2010. In 2003, this was updated to require 20% from renewable sources by year 2020.
Approximately 12,000 MW are required to meet the 10% requirement, which is equivalent to
1200-km2 of offshore generation area based on a typical 2 MW turbine (Byrne and Houlsby
2003).
1.3 United States An exhaustive assessment of offshore wind energy potential for the United States was conducted
(Kilar and Stiller 1980), resulting in a summary of the wind speed and water depth parameters
for all potential wind farm water bodies in the US (Table 7). Significant portions of the US,
namely the Bering Sea, Great Lakes, and parts of northern Alaska, are unsuitable for
development due to deep water, floating ice, earthquake activity, or poor wind conditions. The
most favorable regions found were the Northeast and Northwest Coasts and parts of southern
Alaska. The study was based on a comprehensive evaluation for a 55-unit wind farm of 9 MW
turbines on fixed and floating platforms in water depths from 3-m to 160-m. A tentative
evaluation for water depth up to 50-m and wind speeds of 7 to 8-m/s resulted in an estimated 54
GW of total generating capacity at a wind power sea surface density of 6 MW/km2, with a total
energy production of 102 TWh/year primarily in Northwest, Northeast, and Gulf of Mexico
Coasts (Gaudiosi 1996).
The Gulf of Mexico bordering the Louisiana coast has seen recent wind development interest in
the concept of mounting wind turbines on existing oil platforms. The region has recorded wind
speeds of 11-m/s at a height of 50-m above sea level. Wind Energy Systems Technologies LLC
(WEST) has plans to develop 50 MW of power from several of the 5,200 existing platforms.
WEST plans to purchase the platforms and liability for low cost, efficiently utilizing a structure
which would cost up to $12M to disassemble and remove (Powers 2005).
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Estimates of recently updated generating potential in the United States for shallow (i.e. less than
30-m water depth) and deep waters by region are summarized in Table 8. Distances of 0 to 5
nautical miles (nm) were completely excluded for environmental concerns, with distances from 5
to 20-nm having 67% exclusion, and 20 to 50-nm having a 33% exclusion of the potential
developed area. In total, without considering the Gulf of Mexico and the Great Lakes, areas
from 5 to 50-nm offshore in the U.S. contain 907 GW of energy potential, which is greater than
current US electrical generation capacity. Only 10% of this 907 GW potential is in shallow
waters, emphasizing the importance of the advancement of alternative technologies for deep
water development (Musial and Butterfield 2004).
A more detailed examination of offshore wind farm development in the New England region is
provided later in the monograph.
1.4 Other Countries In China, the extended shallow water depth of the Yellow and Eastern Chinese seas combined
with predominant wind speeds between 7 to 8-m/s results in a total energy production of 254
TWh/year from 129-GW of installed offshore power (Gaudiosi 1996). Some of the issues with
offshore wind turbines in this region of the world are their vulnerability to typhoons, resulting in
extremely large wind and wave loading against the structures.
As shown in Table 4, the countries bordering the Mediterranean and Black Seas have the
potential to utilize over 12,000-km2 of seafloor area, resulting in an energy generation potential
of 72 GW (Gaudiosi 1996).
1.5 Future Prospects in Offshore Wind Energy Development Currently, plans for offshore wind farms across Europe and North America are in various stages
of development. At present, the planning schedule is very dynamic and some projects have been
put on hold or discarded altogether. Table 9 presents a list of the planned projects thus far. It is
not considered to be comprehensive, and is based on year 2002 forecasts (Hendersen et. al.
2002).
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In general, the trend of shallow water offshore wind energy development will likely show an
increase in the number of turbines per wind farm, more efficient and cost-effective foundation
installations, lighter component utilization for the wind turbine structure, and increased design
lifetimes from 20 to 50 years (Gaudiosi 1996).
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2.0 Types of Foundations 2.1 Introduction
The four main classes of offshore foundations consist of piled foundations, gravity base
foundations, skirt and bucket foundations, and floating structures with moored foundations. The
piled and gravity base foundations can be further classified into three structural configurations,
namely, monopiles, which are designed as piled foundations and exhibit simplicity in fabrication
and installation, tripod or quadruped configurations, which can be both piled or gravity based,
and lattice configurations, which offer the most economical structural solution in terms of steel
weight-to-capacity ratio. There are advantages and disadvantages fore each foundation class and
structural subclass primarily based on site conditions and turbine size, as summarized in the
following descriptions.
2.2 Piled Foundations
2.2.1 Introduction
Piled foundations are the most common form of offshore foundation, transferring both tensile
and compressive loads from the foundation into the seabed. They have been installed since the
1940’s in water depths up to 150-m. They are simple to construct using large steel tubing and
offer the most economical manufacturing option among the different types of foundations. The
installation method involves lifting or floating the structure into position using equipment such
as floating crane vessels, drilling jack-up units, and specially constructed installation vessels
before driving the piles into the seabed. Installation depths are typically dependent on the
environmental and soil conditions, and can range from 5-m to over 120-m below the seafloor for
some offshore structures.
2.2.2 Monopiles
Currently, the most popular design for offshore turbines is the monopile, a simple design that is
made of large diameter (e.g. 4-m with tubing thickness of 5-cm) cylindrical steel tubing with a
transitional sub-structure that connects the pile to the turbine tower, effectively extending the
turbine tower below the water surface and into the seabed. The pre-fabricated transitional
component is cast with the monopile, which can account for rotational and verticality errors that
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occur during pile driving. The monopile is driven or hammered into the seafloor, and can be
unsupported or supported (Figure 1). Advantages include minimal seabed preparation
requirements, resistance to seabed movement, scour, ice flow damage, and inexpensive
production costs. The disadvantages include sub-structure flexibility at greater depths, expensive
and time-consuming installation that is dependent on turbine size and seafloor geology, and
decreased stiffness relative to other foundation types, which can result in soil degradation or
“potholing” around the pile (Irvine et. al 2003). Monopiles are also difficult to remove and as a
result may pose problems for decommissioning activities. The limiting design condition for the
monopile is the overall deflection and vibration during loading. Standard monopiles (i.e. no
lateral support) are suitable for water depths up to 25-m. Supported monopiles (i.e. monopiles
braced by lateral struts) are suitable for depths from 20 to 40-m, and are better suited to non-
homogeneous soils. Monopiles should generally be avoided in deep soft soils due to the required
length of installation.
2.2.3 Multiple-Leg Foundations
Multiple-leg foundations consist of tripod or quadruped structures made of cylindrical steel
tubing with either angular or vertically-driven leg piles (Figure 2a), suitable for water depths
from 25 to 50-m. They can also be built with a gravity base (Figure 2b), where piles driven
below the base result in shared load between gravity and the pile, suitable for water depths up to
25-m. Advantages include their versatility, resistance to wave and current loading, and
inexpensive fabrication cost. The disadvantages include expensive construction and installation,
difficulty in multiple-pile removal, and their lack of resistance to ice flow damage (e.g. if slender
sub-structure members are used).
2.2.4 Lattice Towers
Lattice towers consist of a multi-leg foundation connected to a steel braced sub-structure (Figure
3). The piles are driven inside pile sleeves to depth for structural stability, suitable for water
depths of 20 to 40-m. They are characterized by a stiff, dynamic response which makes them
ideal for deeper waters in extreme environmental conditions, with the exception of ice load
vulnerability due to the slender nature of the structural members. Another advantage is the ability
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to be fully assembled before float-out installation. They generally require less scour protection
than other piled foundation types.
2.3 Gravity Base Foundations
2.3.1 General Description
Gravity base foundations consist of a slender steel or concrete sub-structure mounted onto a
single large square or circular foundation of reinforced concrete or a ballast-filled steel shell
(Figure 4). They can comprise multiple foundations similar to the tripod or quadruped concept
described in the previous section. Gravity base foundations can be skirted, which have the
advantages of confining any soft soil layers and transferring the gravity load to the bearing soils,
improving the hydraulic conditions, reducing the scour potential, and facilitating conditions for
base grouting.
Gravity base foundations are generally well suited to homogeneous soils due to settlement and
bearing capacity distribution. However they can be used in virtually all soil conditions in water
depths between 0 to 25-m. They require a flat base and scour protection requirements are
dependent on local site conditions. The gravity foundation is designed to avoid tensile loads
between the bottom of the support structure and the seafloor by providing sufficient self-weight
dead loads to maintain stability. The overturning moment is resisted by a “push-pull” action
where equal and opposite vertical loads (i.e. soil resistance on the downwind side and self-weight
on the upwind side) act at the foundation level.
An advantage to gravity base foundations includes simple transportation, as they can be installed
partially in a fabrication yard and finished in the final position at sea, depending on the
fabrication yard capacity, the available draft (i.e. water depth) during transport, and the
availability of ballast materials. They can also be competitive when heavy lift vessels or other
special installation vessels cannot be mobilized to the site. Gravity base foundations are also
easily removed upon decommissioning.
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2.3.2 Material Considerations
Gravity base foundations are economically competitive when environmental loads are modest
and when ballast for the foundation (e.g. pumped-in sand, concrete, rock, or iron ore) can be
provided at low cost. Addition of ballast material can significantly reduce the size of the
foundation required to achieve the conservative zero tension condition across the foundation base
(Figure 5) (Houlsby et. al. 2001). When considering the foundation material, although
lightweight, steel foundations can be filled with a very heavy mineral named olivine to stabilize
the foundation (DNV-OS-J101 2004). In general, steel foundations are easier to transport due to
lower weight and are easier to erect due to lightweight crane requirements. In either material
case, gravity base foundations can be expensive due to the large weight involved, and will be
subject to increased wave and current loads when the foundation is above the seafloor elevation.
2.4 Suction Caissons
2.4.1 General Description
Suction caissons are similar to gravity base foundations in shape and size but differ in the
method of installation and primary mode of stability. They consist of a sub-structure column
connected to an inverted steel bucket through flange-reinforced shear panels (Figure 6a. The
shear panels distribute load from the column center to the edge of the bucket, which is comprised
of vertical steel skirts extending down into the seabed from the horizontal base that rests on
seafloor. The skirt length is approximately equal to the bucket width, where the volume of soil
inside the caisson acts as a permanent gravity base foundation. Typical dimensions of suction
caissons range from 2 to 4-m in diameter for water depths less than 5-m, and up to 12 to 15-m in
deeper waters (Houlsby and Byrne 2000). Suction caissons can also be designed in a tripod or
quadruped configuration, where the buckets replace piles or gravity base foundations in a
conventional multi-leg structure, offering the advantages of applicability to deeper water, smaller
required caissons, and easier ability to level the structure. The primary limiting design condition
for the monopod suction caisson is the overturning moment, while for a multi-leg suction caisson
configuration, the resistance to tensile loads is paramount. Currently, there are no design
guidelines for suction caissons subject to large moment load to vertical load ratios, and should be
designed on a site-specific basis (Byrne and Houlsby 2002).
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2.4.2 Installation Principle
The installation method of suction caissons is through a pressure differential, where water is
pumped out through top of bucket once the rim of bucket seals with the seafloor (Figure 6b).
This produces a net downward pressure, or suction, forcing the bucket into the seabed. In clays,
the pressure is sufficient to bring the suction caisson to depth, but in sands, water inflow reduces
the effective stresses in the sand, allowing the bucket to penetrate the seafloor. Once installed to
sufficient depth, the pumps are removed and the valves are sealed, with the sand quickly
regaining its bearing capacity. Suction caissons can easily be removed by reattaching the pumps
and pumping water back into the bucket cavity, forcing it out of the seabed.
The application principle of the suction caisson capacity is based on the relatively short wave
periods to which the caissons are subjected, as time for the cavity within the bucket to fill during
wave uplift is too short for significant inflow. The suctions that can develop are inversely
proportional to the soil’s permeability, with the drainage path and wave periods to which the
buckets can be subjected dependent on the depth of the bucket penetration into the seabed.
2.4.3 Site Considerations
Suction caissons are most suitable in homogenous soils due to differential settlement and bearing
capacity issues, but work well in sands and soft clays in a variety of water depths and tidal
conditions. The advantages of this type of foundation include the ability of floating the structure
to the site and the avoidance of heavy lifting equipment and pile driving requirements, making
installation quick and inexpensive, as well as easy removal upon decommissioning. There is a
large cost savings through the short installation time and minimal amount materials necessary for
ballast. It is also not as weather dependent as a piled foundation installation is. A key
requirement to suction caissons is the verification of installation ability and uplift capacity,
where the seafloor soils must be penetrable and not prone to scour. A disadvantage to using
suction caissons is the limited proven installation data for different types of soils, requiring
detailed installation analyses prior to design. Suction caissons are also susceptible to scour in
shallow waters, and piping below the bucket tip may occur in sandy soils.
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2.5 Floating Structures
2.5.1 Introduction
There are two primary types of floating structures suitable for offshore wind turbines; the
tension-leg platform and the low-roll floater. Both configurations can be used for single or
multiple turbines, however the latter is far more cost-effective due to the installation cost of
moorings and/or anchors. The feasibility of floating structures for wind farm design is still in its
infancy, with many novel propositions currently under review. These two most common types
are suitable for large water depths (greater than 50-m) and will be discussed briefly (DNV-OS-
J101 2004).
2.5.2 Tension-Leg Platforms
Tension-leg platforms, which are submerged using tensioned vertical anchor legs with or without
ballast tanks (Figure 7), can be floated to a site in fully-commissioned condition and simply
connected to the moorings or anchors. The base structure dampens the motion induced by wind
and wave loading. Advantages of tension-leg platforms include easy disconnection for transport
for repair or maintenance and installation depths in over 1000-m of water.
2.5.3 Low-roll Floaters
Low-roll floaters are stabilized by mooring chains and anchors which dampen the motions of the
platform (Figure 8). There is a stabilizer installed at bottom of floater to reduce roll. The
installation is simple, similar to tension-leg platforms. The anchors may be fluke anchors, drag-in
plate anchors, suction anchors, or pile anchors.
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3.0 Design Methodologies 3.1 Limit State Design
The limit states design methodology is the most typically used method for offshore wind turbine
support structure analysis. A limit state is defined as the point when a structural element will no
longer satisfy its design requirements. There are four types of limit states; the ultimate limit state
(ULS), the fatigue limit state (FLS), the accidental limit state (ALS), and the serviceability limit
state (SLS). For offshore foundation design, the ULS uses non-degraded soil parameters, while
the other limit states use cyclically degraded soil properties for design calculations (Feld and
Waegter 2002).
3.1.1 Ultimate Limit State
The ULS corresponds to the maximum load carrying capacity, examining strength and overall
stability aspects through elastic and plastic analyses by applying design loads and design
material properties.
3.1.2 Fatigue Limit State
The FLS is failure resulting from cyclic loading due to installation, operational, and non-
operational loads. Installation fatigue is based on a pile drivability analysis and should include
the total fatigue damage accumulation.
3.1.3 Accidental Limit State
The ALS corresponds to damage sustained through an accidental event or operational failure. It
examines load conditions resulting typically from ship impact using very high load characteristic
values applied to the structure for short durations, resulting in undrained soil conditions that may
increase the potential plastic movement of the soil surrounding the foundation. Cyclic soil
strength parameters are not considered due to the inability for soil degradation to occur during
the loading duration.
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3.1.4 Serviceability Limit State
The SLS is a set tolerance criteria applicable to normal use or durability. It uses characteristic
values to analyze deformations that result in settlements or structure rotations in which aesthetic
or operational demands are compromised.
3.2 Load Resistance Factor Design (LRFD) Method The LRDF method applies load factors to characteristic values of loads acting on the structure or
load effects within the structure, and resistance factors to the resistance of the structure or
strength of the structural materials. These factored values are compared to a specified design
criterion to meet a specified target safety level. The characteristic values of the load are
dependent on limit states accordingly; for ULS, the value is defined as the load effect with an
annual probability of exceedance equal to or less than 0.01 (100 year recurrence period); for
FLS, the characteristic load effect history is defined as the expected load effect history; for ALS,
the value equals the most probable annual maximum load effect; and for SLS, the value is
specified depending on certain requirements (DNV-OS-J101 2004). The basis for selection of
the characteristic loads for operational design conditions is summarized in Table 10.
The characteristic resistance is defined as the 5% quantile in the resistance distribution. Load
factors for the ULS and ALS are summarized in Table 11. For permanent loads (G) and variable
functional loads (Q), the load factor should be taken as ψ = 1.0, unless the load is such that a
reduced value leads to an increased load effect in the structure. In such a case, a value less than ψ
= 1.0 is required. The load factors for FLS and SLS are equal to 1.0 (DNV-OS-J101 2004). The
structure should also be able to resist fatigue loading and meet serviceability requirements during
temporary or operational design conditions.
3.3 Direct Simulation The direct simulation method is similar to the LRFD method, except that it is based on direct
simulation of the characteristic combined load effect from simultaneously applied load processes
rather than a linear combination of individual load effects. It is generally more attractive when a
structure is subject to two or more simultaneously acting loads, such as in the case of combined
wind and wave loading. It can be inadequate when a load effect of an applied load process is
20
dependent on structural properties that are sensitive to other load processes. Since correct
assumptions regarding structural damping during various loading regimes (e.g. wind flow, power
production, wave load alignment) of a turbine may be difficult to determine, separate
determination of the wave load effect alone may not be feasible. The characteristic combined
load effect is obtained directly from the distribution of the annual maximum combined load
effect from structural analysis based on the simultaneous application of two or more load
processes (e.g. for ULS design, 99% quantile or 100 year recurrence period). The characteristic
resistance is simply calculated as for the LRFD method (DNV-OS-J101 2004).
3.4 Testing Based Design The testing-based design method consists of establishing load and resistance effects by testing or
observation of performance of the full-scale structures. It is supported by analytical design
methods and includes observation of the performance of the structure.
3.5 Probability Based Design The probability-based design method defines the structural reliability as the probability that the
structure will not exceed a specified failure criterion within a specified time period. The
reliability analysis is based on Level 3 reliability methods, which are applicable to the calibration
of Level 1 methods (e.g. deterministic analysis using one characteristic value to describe each
uncertain variable), special case design problems, and novel designs for which limited
experience exists (DNV-OS-J101 2004). The Level 3 methods use probability of failure as a
measure, requiring knowledge of the distribution of all load and resistance variables. The target
reliabilities should correspond to the consequences of failure, based on established cases known
to have adequate safety. Otherwise, the target reliabilities can be based on transferable target
reliabilities similar to existing design solutions, internationally recognized codes and standards,
or other bases of transfer explained in “Classification note 30.6: Structural Reliability Analysis
of Marine Structures” (DNV-OS-J101 2004).
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4.0 Site Investigations 4.1 Introduction
The environmental conditions at a proposed wind farm location need to be determined in order to
provide an accurate characterization of the environmental loads and subsurface soils to carry out
the design process. The site conditions include data on the local geological, oceanographic,
meteorological, human, and environmental characteristics of a wind farm site. Table 12
summarizes the environmental parameters that affect offshore wind farm design, providing an
outline of the various equipment and techniques used to determine these parameters throughout
the different stages of an offshore wind farm site investigation. Although many of these
parameters have a direct impact on the foundation design, detailed knowledge of the subsurface
conditions is paramount for the safe and economical design of offshore foundations.
The subsurface conditions are used to predict the overall response of the foundation for input
into structural design models. The purpose of a subsurface investigation is to define various soil
and rock data, their associated physical, mechanical, and engineering properties, and the
potential geohazards to the structure. The extent of the site investigation is based on the depth
and area of the seafloor that will affect or be affected by the foundation.
The zone of interest to current offshore wind farm development lies within the continental shelf
region of most of the world. The average width of the continental shelf is approximately 70-km,
the widest areas occurring in the Arctic Ocean and the north and west sides of the Pacific Ocean,
and the narrowest occurring in areas near young mountain ranges. The depths of these areas
range from 10-m to 500-m with an average slope of less than 1 degree, before transitioning to the
continental ridge and continental slope (Poulos 1988). The deepest areas occur near glaciated
zones, and the shallowest near zones with extensive coral growth. The soil types occurring along
the continental shelf and ridge are primarily fluvial marine deposits of sand and silt, transitioning
to silt-clay mixtures and clay further offshore, as illustrated in Figure 9.
Seabed soils are often normally consolidated and exhibit extremely compressive characteristics.
However, they can vary considerably along coastal areas throughout the world. For example, in
the UK, loose mobile sand banks are often found at shallow depths, underlain by soft clay in
22
some areas and glacial till in others. The northwest coast of Australia and parts of the Gulf of
Mexico are characterized by deposits of soft calcareous sand-silt mixtures with localized areas of
cementation that can be embedded through tens of meters of soil depth (Byrne and Houlsby
2003). Therefore, a foundation type that works in one area of a wind farm may not be appopriate
at an adjacent location.
4.2 Phases of the Site Investigation
4.2.1 Introduction
Several major phases are necessary for an offshore site investigation (OSI) that may comprise up
to a year or more of time investment depending on the available data for the proposed location
and the extent of the project. Initially, a planning process must be completed, beginning with a
definition of the scope of the work to be done and ultimately mobilizing the contractor to
perform the work (Table 13). The initial regional desk study is completed during this process to
identify an area of interest for a wind farm and to create a plan for the site investigation. The
regional desk study incorporates all available data to define the probable site conditions that
summarize the current state of knowledge of the site. It will also identify potential constraints
and assist with further planning of the site investigation. The cost of this initial phase typically
comprises approximately 2% of the site investigation costs (Poulos 1988). It should include
investigation of geological databases, bathymetric data, geophysical and geotechnical data,
metocean data, seismicity, performance of existing offshore wind farms, human activities in the
area (e.g. pipelines, shipping routes, et cetera), and environmental concerns (e.g. bird migratory
routes). The planning process can take up to 6 months for large scale developments, and
generally takes a minimum of 3 months for small scale developments (Randolph and Kenkhuis
2001).
The geotechnical risk involved in the development of a proposed offshore wind farm requires a
concerted effort from developers and contractors alike to reduce the risk for all parties involved
in the project. The three key elements in managing geotechnical risk are the initial regional desk
study, the risk assessment, and the site investigation. The risk assessment is based on the
regional desk study and the proposed development to define the range of geotechnical risks
23
involved, their potential consequences, and the options for reducing the impact of those
consequences. This process should remain ongoing throughout the development process.
The site investigation provides the site information necessary to carry out the proposed
development. It should be optimized based on the results of the desk study, the risk assessment,
and the geological and geophysical surveys. Integration of these phases is necessary to ensure
that the soil conditions are determined to the fullest extent required to provide a sound
geotechnical foundation design. Quality control and quality assurance measures should be
implemented throughout the investigation to ensure that the data obtained are as accurate and as
relevant to the proposed development as possible.
The site investigation can then be divided into the geological study, the geophysical survey, and
the geotechnical site investigation, which account for approximately 79% of the total cost. The
remainder of the site investigation cost is divided between the laboratory testing and the
engineering analysis and reporting, which account for 8% and 11% of the cost, respectively
(Poulos 1988). The costs of a site investigation can be expensive, typically averaging $1M per
week, primarily due to the mobilization of the vessel used (Randolph and Kenkhuis 2001).
Using lump sums can help to control these high costs, but typically day rates are imposed by the
contractor due to the high degree of risk assumed. These risks are primarily due to the variable
nature of marine operations, specifically the need to halt investigation progress during inclement
weather. Table 14 shows an offshore site investigation schedule representative of a typical
offshore project. The general timeline will be tailored to the specific project, with phases such as
the collection of oceanographic data reducing in duration if the environmental site conditions are
already known.
4.2.2 Geological Study
The geological study phase is based on the geologic history of the area, and when combined with
the type, size, and importance of the wind turbine structure and the homogeneity of the soil
deposit, it serves as a basis for the extent and method of the geotechnical site investigation.
Geological maps and profiles are often available to provide general site information. Typically,
24
top soil deposits can help to identify zones of interest and aid in locating anchoring locations for
subsequent investigation activities.
4.2.3 Geophysical Survey
The geophysical survey establishes seafloor bathymetry, topography, stratigraphy, and identifies
hazardous areas on the seafloor. The upper stratigraphy should be assessed to determine
anchoring locations and potential problems with positioning jack-up rigs, as they may be used
within close proximity to adjacent existing structures, affecting the surrounding soil. The
topography should be assessed on a scale sufficient to identify outcrops, obstructions, and
seafloor non-uniformities which can affect the relatively small footprint of typical wind turbine
foundations. This is important in minimizing differential settlement of gravity base foundations
over the lifetime of the structure. Sub-base grouting is typically used to account for these non-
uniformities.
The techniques used consist of bathymetry mapping with conventional single or multi-beam echo
soundings or swathe bathymetry, seafloor mapping with side scan sonar or magnetometer
readings, and seismic profiling methods using boomers, sparkers, pingers, chirp profilers, or high
resolution digital surveys using airguns. A survey line plan is constructed for a proposed sight to
create a grid over which the various characteristics identified in the geophysical survey can be
plotted. Any existing data (e.g. boreholes) should be plotted on the survey line plan so that the
geophysicists conducting the survey can calibrate survey data against borehole data to check for
continuity between markers. Grid spacings are typically 75 to 100-m apart with cross lines
spaced at 300 to 500-m, both of which can be adjusted for areas where known hazards exist
(Randolph and Kenkhuis 2001).
4.2.3.1 Bathymetry Mapping
The conventional echo sounder is a heave-compensated device which is calibrated to the
seawater sound velocity. It produces spot measurements of depth along survey tracklines at
distances of 25-m to 50-m apart that compile to form a contour map of seafloor. A Swathe
device is a heave, pitch, and roll-compensated beam that records spot measurements as it sweeps
from side to side over the area of the proposed site. The width of the coverage is approximately
25
four times the water depth. Because of the three dimensional capabilities of the system, it is
ideal for areas of highly variable topography. The minimum required resolution for single and
multi-beam echo sounders is approximately 1% of the maximum water depth (OSIC 2004).
4.2.3.2 Seafloor Mapping
The side scan sonar system maps topographic features on the seafloor and identifies objects on or
above the seafloor. It can help identify the type and distribution of sediment and surface forms
such as sand waves. The system uses narrow beams of sound waves that are transmitted out to
the sides and across the bottom of a hydro-dynamically stable towfish from transducers and
reflected back to a receiver on the towfish. The system is dependent on frequencies of the
transmitting waves, with a trade-off between resolution and distance traveled. Higher
frequencies (500-kHz to 1-MHz) are high resolution, short distance waves, and low frequencies
(50-kHz to 100-kHz) are the opposite. The intensity of the returning wave is dependent on both
the topography and the material properties. The harder the object or material (e.g. gravel) on the
seafloor, the darker the resulting image, as they reflect energy better than soft materials (e.g.
muddy sands). Shadows show up as white areas due to the absence of reflected sound. When
sand waves and ripples are large enough to produce shadows, there is an indication that currents
are particularly high in that area and foundations would be subject to considerable scour. The
resulting map of the variation in color, when combined with bottom samples or cores, produces a
detailed map of seafloor features and interpreted sediment types. The side scan sonar system has
a minimum required resolution of detecting a 10-cm3 object or a 10-cm wide linear object
(OSIC 2004).
The magnetometer device maps the locations of cables, pipelines, shipwrecks, and other ferro-
metallic objects, such as munitions dumps. The minimum required resolution is approximately
one nanoTesla (i.e. one-billionth of the magnitude of a magnetic field vector necessary to
produce a force of one Newton on a charge of one coulomb moving perpendicular to the
direction of the magnetic field vector with a velocity of one meter per second) (OSIC 2004).
26
4.2.3.3 Seismic Profiling
The seismic profiling methods use a sound source and receiver combination that produce a
reflection record of the cross-section of an area of interest. The different methods available range
in their ability to produce clean profiles (Table 15). Similar to side scan sonar, there is a tradeoff
between the resolution and the depth of penetration, the latter of which is dependent on the
seafloor composition and the underlying layers. Typically, the minimum required vertical
resolution should be better than 0.5-m down to a depth of 5-m (OSIC 2004). The choice of the
type of device is ultimately dependent on the specific application. However, the most precise is
the high resolution digital survey (HRDS). Using an airgun as the sound source, the HRDS
consists of multi-channel receivers aligned in streamers that vary in length, spacing, and
grouping to target the resolution and depth of penetration of interest, which can reach up to 60-m
(Jenner et al. 2002). The advantage of the HRDS is that both the sound source and the streamer
are towed at shallower depths than for other equipment, resulting in increased ability to account
for reflections from the water surface. However, the method is more affected by sea-state
variability.
4.2.4 Geotechnical Site Investigation
The geotechnical site investigation (GSI) is conducted following the geophysical survey to use
the information obtained to target soil strata changes or specific seafloor features. The GSI
includes in-situ testing for stratification identification and soil coring and sampling for material
identification, characterization, and subsequent laboratory testing. The geotechnical site
investigation characterizes the material properties of the soil in terms of five components:
macroscopic (e.g. historical and regional characteristics), mechanical (e.g. strength and in situ
stress conditions), microscopic (e.g. particle size, roundness, and shape), chemical (e.g.
molecular structure), and water content properties. The GSI begins with in situ testing to obtain
the stratigraphy of the area of interest, typically done with a cone penetrometer test. Once
assessed, the stratigraphy will guide the development of the remaining in situ testing, sampling,
and laboratory testing program. After completion, the geophysical survey and geotechnical site
investigation data combine to form a geotechnical model of the area to develop the required
design parameters.
27
The depth and aerial extent of the geotechnical site investigation are based on numerous factors,
including the local geology, availability of previous investigations, site accessibility, soil
variability, foundation structural configuration, the expected environmental loads, and limitations
related to time and budget resources. The depth is primarily dependent of the presence of weak
underlying layers which may influence foundation behavior. General rules of thumb have been
established for typical foundation types to limit the required extent of the investigation. For pile
foundations, borings and in situ testing should generally extend to at least one pile diameter
below the level of the pile tip. For gravity base foundations, borings and in situ testing should
extend to a depth equal to at least the largest horizontal dimension of the foundation structure.
For single wind turbine structures, one boring to sufficient depth for recovery of a sample is
recommended (DNV-OS-J101 2004). For wind farms, the soil investigation program should
include combined techniques (e.g. one boring in each corner and center of farm combined with
one CPT at each foundation location). The soil investigation program should be monitored by a
geotechnical engineer to allow for program adjustments in case of soil non-homogeneities, the
presence of problematic soils, or modifications to the second phase of testing if using a multi-
phased investigation program. The variety of equipment and techniques used in offshore site
investigations is required to obtain quality in situ test results and quality samples for the array of
soil conditions (e.g. various combinations of sand, silt, clay, rock, and shell deposits) in the
offshore environment.
4.3 Offshore Geotechnical Site Investigation Vessels
4.3.1 Introduction
Site investigations can be performed from a variety of vessel types, the selection of which is
controlled by sea conditions (e.g. water depth, sea state, anchoring conditions), the planned
program of investigation, and the availability and cost of the equipment (Watson 2000). Vessel
deployment and use can be expensive and usually constitutes the largest component of the site
investigation cost. The options are grouped into four basic categories: in waters less than 20-m
deep, a support vessel or low-draft barge using commercial divers is typically used; in waters
greater than 20-m, specialized drilling vessels are used, or jack-up rigs if the water depth is less
than 80-m; and for all water depths deployment of remote equipment from support vessels is
typically included (Randolph and Kenkhuis 2001). Depending on the type of vessel used,
28
anchoring may be necessary and is limited to water depths less than 200-m; otherwise
dynamically positioned vessels are used.
4.3.2 Low Draft Barges
Low-draft barges are ideally suited to perform structure-to-shore pipeline and cabling
investigations. Anchored using a multi-point mooring system, the minimum number of anchors
is generally four unless high ocean currents cause instability.
4.3.3 Jack-up Rigs
Jack-up rigs provide a stable working platform, however, expensive daily rates (e.g. up to $150K
per day) and significant support requirements (e.g. manpower, transportation, and support
vessels) can reduce their cost-effectiveness (Randolph and Kenkhuis 2001). They are typically
used for oilfield activities and therefore tend to be unsuitable for geotechnical work unless land-
based drilling equipment is provided. They also have space limitation issues and are not used for
multi-boring investigations, as repositioning the rig is time-consuming.
4.3.4 Specialty Geotechnical Drilling Vessels
Specialized geotechnical drilling vessels are ideal for multi-boring investigations. They typically
have an on-deck crane with ample space, a moonpool for lowering equipment, and use a heave
compensation system that can produce high quality geotechnical results. Their disadvantages
include cost (e.g. up to $100K per day), limited endurance, and limited fresh water supply, which
reduces the choice of drilling fluids to seawater-compatible types (Randolph and Kenkhuis
2001). The availability of these types of vessels is fairly limited. In the year 2002, there were
less than 10 geotechnical drilling vessels in operation throughout the world (Jenner et. al. 2002).
Most of these vessels are self-anchored with dynamic positioning systems (DPS) that allow deep
water investigations to be conducted. The DPS also allow for large-scale investigations while
minimizing the risk of losing expensive equipment. A wheeldrive drilling system is typically
used that allows for a high production rate and affords the capability of continuous profiling of
the subsurface stratigraphy.
29
4.3.5 Semi-submersible Drilling Rigs
Semi-submersible drilling rigs are similar to jack-up rigs but more expensive with costs up to
$275K per day (Randolph and Kenkhuis 2001). The anchoring process can take up to two days
between installation and removal. However, multi-boring capabilities are possible with the long
anchor lines that allow the rig to move over the footprint of a large foundation.
4.3.6 Supply Vessels
Supply vessels are used for deploying remotely operated equipment using an A-frame or on-deck
crane. They have a multi-point mooring system limited to less than 125-m of water depth without
anchor handling assistance. They typically have a poor ability to maintain a stationary position
and therefore should only used to transfer low-risk equipment. Costs are typically up to $25K
per day (Randolph and Kenkhuis 2001).
4.4 Drilling The drilling process should be conducted with the objectives of minimizing sample disturbance
and providing borehole stability. The drilling process can contribute to the site condition insight
as some of the drilling parameters give a rough idea of the soil conditions via
interpolation/extrapolation of soil data, explanation of sample disturbance, and detection of soil
stratigraphy. In certain soil types, such as soft rock or hard clays, rotary core drilling may be the
best option. The sample quality is much less than for push or position systems, but recovery
tends to be adequate. The most common types of rotary core drilling are Christensen coring and
high speed diamond coring.
Where more detailed investigation is required, geophysical borehole logging can be used through
a wireline system. Some of the tools in use include: calipers, which measure the borehole
diameter; gamma radiation, which measures the radioactive isotope concentration and identifies
soil type and grain size distribution; neutron devices, which provide information on porosity;
sonic devices, which measure sound velocity to identify soil type and character; and density
devices, which provide information on the soil density.
30
Where shallow gas may be encountered, special precautions should be taken during drilling
procedures, which include but are not limited to, no hot work or smoking on deck, keeping a
supply of heavy mud available, having gas detectors on deck, and utilizing wind and current
meters to provide information on gas dispersion should it escape from the borehole. A pilot hole
without any downhole tool use should be drilled with a non-return valve in the drillstring. Using
this technique, a borehole can be drilled next to the pilot hole with full use of downhole tools. A
safety valve on top of the drillstring should be included when using wireline downhole tools, and
an offset between the vessel and the borehole should be applied based on wind and current
conditions.
The drilling and coring process differs across the various types of vessels used. With barge
vessels, subsea methods (i.e. divers on seafloor conducting the test) have several limitations
including operating only during daylight hours, divers having little geotechnical drilling
experience, short bottom times, shallow borehole depths (e.g. < 40-m), and drilling fluids are
limited to seawater due to visibility requirements for the divers. Video and audio
communication allows for the geotechnical supervisor aboard the barge to direct activity.
However, complete oversight is uncommon as it also their responsibility to log and pack
samples.
From jack-up units, onshore drilling equipment can be mobilized on the rig, allowing
experienced drillers to conduct the drilling operations. This technique usually produces the
highest quality results provided that all necessary investigation equipment is available. A choice
of drilling fluids is possible through ample fresh water supply, and operations can continue 24
hours per day.
From specialty geotechnical drilling vessels or semi-submersible rigs, four main components
comprise the operation: a seafloor reaction frame and utility guide frame, a heave compensation
system, a bottom hole assembly, and umbilical downhole probes. The seafloor guide frame is
lowered into position under tension, using ballast to provide reaction on the seafloor. During
drilling operations, a hydraulic clamp mounted on the guide frame provides the reaction to the
drill string. The drill string is typically 5-in in diameter to allow for the downhole tools to pass
31
through it. On specialty drilling vessels, a mud swivel on the top of the drilling unit allows
insertion of the tool. However, on semi-submersible rigs, the drill string must be broken to allow
insertion. The bottom hole assembly consists of an open drag bit, drill collars, and latching
sections to provide weight to the setup and connect the downhole tool. The weight is necessary
to provide reaction to the motion compensator to prevent disturbance to the testing and sampling
activity.
4.5 Sampling
4.5.1 Introduction
There are three main types of testing, sampling, and coring methods: the wheeldrive system,
push and piston sampling, and coring techniques. Within these three categories are the types of
equipment used, with consideration in order of priority given to 1) thin-walled piston sampler, 2)
thin-walled push sampler, 3) thick-walled push sampler, 4) hammer sampler, and 5) rotary core
sampler. The techniques and equipment type should be chosen with consideration of the soil
type, the type of testing to which the samples will be subject to, and of minimizing sample
disturbance during both penetration of the device and extraction to the surface.
Sampling can be done using one of two primary methods: the seafloor mode, where sampling is
performed from the seafloor using a reaction frame with drilling; and the drilling mode, where
sampling is performed through a drillstring at the bottom of the borehole. The seafloor mode is
a umbilical technique that uses a electrical wires to send data to the surface in real time and
hydraulic hoses to perform the test. The technique allows for alternative sampling and testing to
be conducted at one borehole.
The length of the sample obtained is dependent on several factors that include the geometry and
characteristics of the sampler, the soil type, and the type of applied force (e.g. push, vibration,
hammering). The cores removed from the sampling device should be photographed upon
extraction and recorded in terms of color, texture, odor, and moisture characteristics. This
should be performed immediately following removal as the appearance of the materials can
change quickly when exposed to air. If rock coring is expected, is it possible to use a “piggy-
back” hard-tie (i.e. motionless) system of mounting a separate drill rig on top of a platform
32
which is mounted onto a separate riser that is secured to the guidebase on the seafloor. A
separate drill string is then used to core the rock samples through the riser system.
The speed of the sampling process and penetration rate is dependent on the phase of the site
investigation. For the geotechnical phase, the penetration rate of sampling equipment will be
tailored to achieving the highest sample quality for subsequent laboratory testing, while during
the geological phase, the sampling penetration may be faster in order to cover as much area as
possible for determining site stratigraphy. Any shock or vibration should be avoided unless
samples are being retrieved for classification purposes only. Recovery of the sample on deck
should also be conducted in a manner that minimizes sample disturbance.
4.5.2 Wheeldrive System
The wheel drive system utilizes a heavy frame to provide thrust reaction, driving the rod into the
seafloor using hydraulic motors mounted in series on top of the frame. It is primarily used for
cone penetration testing, T-bar testing, and vane shear testing. However, it can be modified to
perform rock coring and piston sampling. It is currently the most advanced equipment for deep
water site investigations.
4.5.3 Grab Sampling
Grab sampling is used to obtain large volumes of soil for classification purposes. Its applicability
is suited for pipeline investigations where scour and erosion can be assessed for long lengths of
the seafloor in a relatively short period of time. The grab sampler is essentially an articulated
bucket ranging in size from a few liters to a cubic meter that closes upon contact with the
seafloor. The method is simple and inexpensive, and can be equipped with hydraulics to obtain a
higher recovery. It can also be used when other devices (e.g. vibro corers or gravity corers)
cannot penetrate the seafloor sufficiently due to harder surface materials. The disadvantages
with grab sampling are the high level of sample disturbance, the risk of washout during retrieval,
and shallow depth of penetration. Non-representative samples may be obtained in cobbled soils
and gravels and therefore require multiple grabs to achieve a large enough volume to adequately
represent the soil type.
33
4.5.4 Push and Piston Sampling
Piston samplers are a modification to drop corers via addition of a cantilevered weight attached
to the top of a core barrel. The weight acts to penetrate the core barrel into the seabed at a pre-
determined height above the seafloor. The piston inside the plastic-lined barrel ensures adequate
suction pressure to retain the sample in the barrel during extraction to the surface. For both
piston and push sampling, thin-walled tubes (i.e. < 2-mm with diameter ≈ 75-mm) should be
used whenever possible to minimize soil disturbance. In highly sensitive soils such as soft clays,
shallow cutting angles (i.e. 5 degrees) on the shoe of the sampler should be used (Norsok 2004).
4.5.5 Coring Techniques
4.5.5.1 Remote Coring
The remote coring system is an autonomous drilling unit that is hydraulically operated from a
deck control system. The sampling is conducted via video monitoring in real time. The drilling
fluid consists of seawater and the maximum drillable depth is approximately 5-m. The
maximum water depth where this system can be used without weight compensation for the
hydraulic umbilical line is approximately 130-m (Randolph and Kenkhuis 2001). It is ideally
suited for pipeline investigations where numerous borings must be taken. The subsurface
stratigraphy can be monitored during testing by mounting additional instrumentation onto the
reaction frame resting on the seafloor.
4.5.5.2 Vibro Corer
The vibro corer is used when sandy soils are prevalent and seafloor penetration resistance may
be too high for the self-weight insertion of the piston corer. An electrical vibrating motor allows
the plastic-lined steel tube corer to penetrate the seafloor via rotating eccentric counter weights,
with barrel lengths typically reaching up to 8-m (OSIC 2004). To retain the sand sample upon
extraction to the surface, a core catcher is fitted to the cutting shoe at the toe of barrel. Since the
sample is considered highly disturbed due to vibration and densification in the barrel, their use is
typically for classification purposes only. The water depths suitable for the vibro corer are
generally less than 800-m (OSCI 2004). The advantages with vibro coring include its simple,
lightweight, and inexpensive deployment. The disadvantages include lack of penetration ability
in dense sands or hard clays, lack of stratification identification accuracy due to compaction,
34
plowing, sample loss issues, high level of disturbance applied to the soil, and difficulty with the
handling of the larger vibrocore equipment.
4.5.5.3 Gravity Corer
The gravity corer is typically used in place of the vibro corer in soft soils, as sample disturbance
tends to be lower without vibration, especially when piston-type gravity corers are used. The
principle of the device is to utilize the force of gravity during the last 5 to 10-m of deployment to
allow the corer to penetrate the seafloor under its own weight. Quality samples can be obtained
for depths up to 20-m below the seafloor with some of the larger piston-type gravity corers. The
method is inexpensive, quick, and easy to conduct. The disadvantages with the gravity corer
include lack of penetration is dense sands and stiff clays, and danger or difficulty in handling the
trip-release mechanism on the piston-type corer.
4.5.5.4 Box Corer
The box corer is a lightweight system used to obtain an undisturbed block of clay that can later
be trimmed into multiple specimens for laboratory testing. The box corer is suitable primarily
for soft clays and can retrieve samples from 10 to 50 liters in size (OSIC 2004). The primary
disadvantage is the limited depth of penetration for the device.
4.5.5.5 Rock Corer
The rock corer is a seafloor mounted device that can obtain continuous samples of rock from 3 to
9-m in length (OSIC 2004). It consists of a rotary drive mechanism with a diamond or carbide
tipped drill bit with an inner core barrel. Remote systems exist that allow for autonomous
handling of the equipment in deeper waters. Although the remote systems allow for deep water
use, the reduction in the human factor of equipment operation tends to decrease the core quality.
The main disadvantage to the rock corer is the lack of ability to obtain high quality cores. Lack
of hole stability below granular soils, and size and weight of equipment are other drawbacks to
its use.
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4.5.6 Other Sampling Methods
Besides soil sampling, pore water, gas, or ambient pressure sampling can be conducted to obtain
other data important to site investigations and foundation behavior. For deep water sampling,
drillstring-based equipment such as wireline percussion samplers, latch-in push samplers,
hydraulic piston corers, pressurized core barrels, and samplers operated in a stabilized drillstring
can also be used to obtain soil samples (Poulos 1988). Further details for soil investigations and
sampling are explained in the Norsok Standard G-001 Marine Soil Investigations (Norsok 2004)
and the DNV Classification notes No. 30.4 (DNV-OS-J101 2004).
4.5.7 Sample Recovery, Storage, and Transport
Once the sample has been retrieved from the seafloor and brought to the deck of the vessel, care
must be taken during handling and storage activities to minimize sample disturbance and
maximize sample recovery. Samples can be stored in the tubes with which they were obtained in
for push and piston samples. By placing a cap on the bottom of the tube immediately upon
retrieval, soil and water escape will be prevented. The sample is then either extruded for
offshore laboratory testing or sealed for transport to the onshore laboratory dependent on the
strength and stiffness of the soil and the possibility of access to free water or mud. Extruded
samples intended for onshore testing should be sealed with plastic and aluminum foil, followed
by waxing to seal in the moisture content. It is important to note the spatial orientation of the
sample in its in situ state to provide a correct reinstatement of the in situ stress state during
laboratory testing.
All samples being stored should be located in an area with a constant temperature (e.g. 7°C in
onshore laboratory) and humidity, away from locations where excessive vibrations or heat
transfer may occur, such as engine rooms. Samples should never be exposed to temperatures
below freezing, as ice will form in place of the water causing swelling and characteristic soil
properties to change. Transportation of samples should be conducted to minimize shock and
impact loads to the samples. The samples should also be stored and transported in the same
orientation from which they were retrieved. Horizontal storage and transport can be used for
firm or stiff soils.
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4.6 In Situ Testing
4.6.1 Introduction
In situ testing techniques include cone penetration testing (CPT), field vane testing, and various
geophysical survey techniques such as downhole seismic testing. In situ testing is conducted
using one of two methods similar to the sampling modes, namely the seafloor mode and the
drilling mode. In both methods, the in situ tool is inserted into the seabed soils to either a pre-
determined depth or until refusal is met. With the seafloor mode, it is important to ensure that
the rig does not affect the results of the in situ test. This can be reduced by considering whether
the rig’s footprint is circular or square shaped with an open space where the in situ tool is placed,
whether skirts could transfer the rig weight to stiffer surrounding soils, and whether the weight of
the rig is sufficient to provide no more than what is required to maintain position during testing.
Video cameras can be mounted to the rig to monitor testing and verify that these conditions are
optimized. Remotely operated vehicles can also be used to perform shallow testing. The drilling
mode is conducted to minimize soil disturbance below the drill bit. The in situ tool should be
extended to a depth of 1-m below the drill bit if soil strength and density allow.
4.6.2 Cone Penetration Testing
CPT testing typically uses a instrumented probe that allows for measurement of cone tip
resistance, qc, sleeve friction, fs, and for the piezocone device, excess pore water pressures, u,
that generate as the cone is pushed through the soil. The CPT is primarily a profiling tool,
however, the measurements obtained from the two load cells can provide estimates of undrained
shear strength with consideration for the degree of uncertainty, which can be rather high. The
parameters that can be derived from the test include soil type, relative density, and stress history
using empirical correlations. The advantage of the CPT over coring methods include a greater
penetration depth, continuous measurement of the stratigraphy, real time data acquisition,
reduction in the amount of laboratory testing that is typically required for samples, reliability of
relative density determination for granular soils, and speed of testing. The main drawback of the
test is the lack of retrieval of a soil sample, which is necessary to verify that the empirical
correlations used are valid for the particular type of soil. The weight of the equipment required
for reaction force is typically high, ranging from 5 to 10-tonnes, and the expertise required to
conduct and interpret the test properly are other disadvantages of the CPT.
37
The testing procedure should be conducted at a penetration rate of 2-cm/s as continuous as
possible dependent on the limitations of the equipment. Pore pressure dissipation tests are
conducted during pauses in the penetration phase to assess the porosity of the soil. They should
be carried out to greater than 50% consolidation, depending on the importance of the results and
the time necessary to achieve the target level of consolidation. The accuracy of the
measurements is based on soil type, distance between measurements, and the required degree of
precision necessary (Table 16). Class 1 is for precise characterization in soft or loose soils. Class
2 is more appropriate for stiff clays and sands, while Class 3 is for stiff or dense soils.
A seismic capability can be added to the CPT to measure the shear wave velocity (to calculate
the shear modulus of the soil), in which sensors are located behind the friction sleeve of the
penetrometer. The seismic source is installed on the seafloor frame and the data acquisition is
connected through a umbilical line. The depth of shear wave penetration should reach down to
80-m below the seafloor under favorable conditions.
The electrical conductivity of the soil can also be measured during the CPT using two electrodes
placed on the penetrometer behind the friction sleeve.
4.6.3 T-bar Testing
The T-bar tool is a recently developed device similar to the cone penetrometer that provides a
more accurate means of estimating undrained shear strength in soft clays. Although the sleeve
friction is not obtained, the resistance measured is subject to less uncertainty, and has the ability
to measure the remolded strength (wheeldrive version only) of the soil as well. The surface of
the T-bar is slightly textured through sandblasting, while the cone penetrometer tip surface is
machined smooth. The T-bar is mounted on the cone penetrometer base to utilize the cone tip
load cell to measure the T-bar resistance. There is no current standard operating procedure for
conducting the T-bar test. However, the Norwegian Geotechnical Institute and the Center for
Offshore Foundation Systems in Australia recommend that the T-bar resistance be measured
during both penetration and extraction at a rate similar to the CPT. They also recommend that
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for tests deeper than 5-m, the rod inclination be measured, and that for cyclic testing the vertical
displacement limits are to be set at 0.5-m for at least 6 cycles (Norsok 2004).
4.6.4 Field Vane Testing
The field vane shear test is conducted with a cross-shaped probe that is pushed into virgin soil at
least 0.5-m ahead of the sampling device and rotated to measure the peak and residual shear
strengths of the soil. The speed and number of rotations, direction of rotation, and the size of the
vane (Table 17) can be adjusted to suit the soil type being tested and the type of strength desired
(i.e. drained or undrained). The measured undrained shear strength is considered to be more
accurate than the CPT measurements in soft clay soils down to penetration depths of 3-m below
the seafloor. The main drawback is the slow operation speed and its limited applicability to
other soil types.
4.6.5 Other In Situ Tests
Other in situ tests that may be used in certain circumstances include pressuremeter testing
(stress-strain characteristics), hydraulic fracture testing, dilatometer testing, permeability testing,
in situ density testing, pile-model testing, seismo-acoustic techniques, plate loading tests, and
pipeline trenching evaluation. The primary advantage to using some the model tests is the ability
to directly model the predicted foundation behavior. Disadvantages include the size and cost of
the larger-scale testing and their narrow scope of target parameters.
4.7 Characteristic Soil Properties In situ testing and sampling are performed to gather the necessary geotechnical data for the soil
deposits of interest. The exact information needed depends primarily on the type of foundation,
but numerous parameters are needed that are independent of the type structure to be installed.
The data should include, but are not limited to (Norsok 2004, OSIC 2004):
1) Summary of soil conditions: soil classification, description, and stratigraphy:
total unit weight, solids unit weight, water content, void ratio, porosity, relative
densities, liquid and plastic limits, and grain size distributions
39
2) Basic soil parameters: effective in situ overburden stress, σ’vo; in situ pore
pressure, uo; preconsolidation stress, σ’p; overconsolidation ratio, OCR;
coefficient of lateral earth pressure at rest, Ko; relative density of sand layers, Dr;
3) Deformation properties: undrained shear modulus, G; drained Young’s
modulus, E; Poisson’s ratio, ν; constrained modulus, M; horizontal and vertical
coefficients of consolidation, ch and cv; coefficient of permeability, k; creep
parameters; cyclic loading parameters for settlement calculations; small strain
shear modulus, Gmax; damping ratio, ξ
4) Shear strength parameters: friction angles for granular material, index,
undisturbed, and remolded undrained shear strengths, su, and sensitivity for fine-
grained material; pore pressure parameters; parameters to describe excess pore
pressure development and shear strength degradation due to cyclic loading; for
drained clay analyses, cohesion and friction angles are needed
5) Parameters for specific applications: contour diagrams for cyclic effects; base
contact stress parameters; skirt penetration resistance parameters; for piled and
gravity base structures, mud mat stability and settlement parameters; for jack-up
platforms, parameters for stability settlement and punching failure; for geohazard
analysis, slope stability shear strength parameters; for earthquake analysis,
dynamic soil parameters
Along with the basic parameters, the appropriate types of in situ and laboratory tests to obtain
specific parameters vary between soil type (Table 18). Additional soil parameters and their
respective appropriate analysis for other offshore applications that may overlap with wind farm
investigations in shown in Table 19.
When characterizing soil parameters, consideration of the potential variability of the soil
properties based on the volume of the tested sample should be addressed. A larger volume
results in an averaging effect over the variability from point to point within the soil volume,
while a smaller volume needs to account for the full variability of the local soil property. The
characteristic value should be a cautious estimate of the property where the probability of a
worse value (i.e. having a negative effect on the capacity of the soil property) is not greater then
5%, depending on the design limit state in question (DNV-OS-J101 2004).
40
4.8 Laboratory Testing The laboratory testing phase of the offshore site investigation provides the majority of the
geotechnical soil data. It is divided into two phases: the offshore testing program and the
onshore testing program.
4.8.1 Offshore Testing Program
The offshore testing program consists of routine classification tests following extrusion from
sampling devices and undrained shear strength index tests for cohesive soils, and is necessary to
update the in situ testing and sampling program based on the observed conditions. Wet and dry
densities of the soil and the water content are also determined, as well as a determination of the
sample disturbance. Select sections of the sample are then sealed with wax and stored in a
humidity and temperature controlled environment for further onshore laboratory testing.
4.8.2 Onshore Testing Program
The extent of the onshore laboratory testing phase is tied to the in situ test results and soil
classification. Therefore, it is dependent on the degree of homogeneity of the soil deposit,
possible stratification, the presence of critical and weak soil layers, and the design strategy. It
should account for effects from thin problematic layers or pockets whose location relative to a
foundation is unfavorable. Much of this information is based on the geophysical survey and the
in situ testing results. The lab program should also account for actual stress conditions in the
soil, including effects of cyclic loading caused by environmental loads. The duration of the
laboratory testing phase can vary depending on the aforementioned considerations, however
typical durations range from 3 weeks to 3 months (Randolph and Kenkhuis 2001). After this
phase is complete, the results should be integrated with the in situ testing results to provide the
design team with a comprehensive geotechnical design basis to begin analyses of foundation
options. After a foundation design is finalized and construction begins, verification of the design
basis should be conducted through monitoring foundation installation activities. This may
include, but is not limited to, 1) blowcounts during pile driving to verify penetration resistance,
2) suction pressure measurements during caisson penetration to verify uplift capacity, 3) primary
and secondary settlement measurements during and after gravity base foundation installation,
41
and 4) installing geotechnical instrumentation for long term verification of performance (e.g.
storm event deformation, cyclic load deformation, differential settlement).
4.8.3 Soil Classification
The soil description and classification should be based on an established classification system
(e.g. Unified Soil Classification System) and should include the following characteristics: soil
type, secondary or minor soil components important to the soil properties, clay strength or
relative density and grading of sand, structure and texture of soil, color and odor of soil,
photographs of each discernible layer, and any special characteristics such as the presence of
cementation or calcareous ooze.
The water content should be obtained for any samples that would subsequently be tested further.
It should be obtained both prior to and following the advanced testing, and should comprise a
series of continuous measurements to construct a water content profile of the subsurface.
The liquid limit, wL, and plastic limit, wP, should be obtained based on standardized procedures
for Atterberg limits. The method should indicate if the soil was dried prior to testing and
whether coarse material was removed from the soil.
Other parameters to include in this phase are the bulk density of soil or soil unit weight, specific
gravity of soil, maximum and minimum density (for cohesionless soils), grain size distribution,
sand grain angularity, and radiography of the soil to determine the quality of the samples.
4.8.4 Index Testing
Index testing is performed on cohesive soil samples to estimate undrained shear strength. The
types of index tests include fall cone tests, in which the depth of penetration of a freefalling small
conical instrument is measured; pocket penetrometer tests, in which the force to penetrate a steel
rod into the sample is measured; Torvane tests and miniature vane tests, in which the torque
required to rotate a bladed cylinder embedded in the sample is measured; unconfined
compression tests (UCT), in which the vertical force to fail a cylinder of soil is measured; and
42
unconsolidated-undrained (UU) triaxial tests, in which the sample is confined to the in situ stress
for 10 minutes before being subject to a vertical force.
Also, the remolded strength and sensitivity of cohesive soils should be measured during this
phase using one of the aforementioned index tests or by using a ring shear device. The
sensitivity, St, is the ratio of the undrained shear strength of the undisturbed soil to that of the
remolded soil. Artificially low sensitivities can result from excessive sample disturbance and
should be considered as such when interpreting the results.
4.8.5 Consolidation Testing
Testing for determination of the deformation properties of fine-grained soils is primarily the
oedometer test, in which the number of tests should be at least 2 per soil layer (DNV-OS-J101
2004). The results from laboratory and in situ testing must consider the inherent uncertainty and
differences between results from the various testing methods. These differences may be from
soil disturbance due to sampling, samples not reconstituted to in situ stress history, the presence
of fissures, different loading rates between the test and the limit state, simplified laboratory test
representation of complex load histories, and/or soil anisotropy effects. The installation effects
should also be considered when implementing testing results in design.
The selection of the type of consolidation test should be based on the stress history of the soil
and the desired loading program including unload-reload cycles. The two primary types of tests
are the incremental load test and the continuous load test (e.g. continuous rate of strain (CRS)
test). The incremental load test is conducted by doubling the applied load at each increment,
allowing the specimen to reach primary consolidation at each load step according to Taylor’s
method. The CRS test compresses the specimen at a constant rate of vertical strain ranging from
0.2% to 5% per hour for a single drainage 20-mm specimen. Pore pressure is measured from the
bottom of the specimen and drainage should be allowed at the top. The rate of strain should
correspond to limiting the pore pressure to less than 10% of the total applied vertical load.
Horizontal stress should also be measured during testing to obtain the coefficient of horizontal
earth stress at rest, Ko.
43
Soil permeability should be measured using constant head permeability tests under constant
stress, where the pore pressure should always be increased during testing, again limited to less
than 10% of the total applied vertical load. Testing should continue until steady state conditions
are attained.
The coefficient of consolidation, cv, can be calculated from the coefficient of permeability, k, and
the constrained modulus, M, of the stress-strain curve, according to the following:
w
vMkcγ
= Equation 4-1
where: γw = unit weight of water Sample quality evaluation can be performed to assess the confidence in the results of the
laboratory testing. Using the initial void ratio, ei (ratio of pore space volume to soil particle
volume), and the axial strain at the effective in situ overburden stress, a qualitative indication can
be estimated using charts similar to those developed by Lunne et. al. (1998) in
Table 20.
4.8.6 Strength Testing
There are three primary test types used for determining strength properties. Static direct simple
shear (DSS) and triaxial (TX) tests are used for sands and normally–consolidated clays, while
cyclic DSS and TX tests are used for soils sensitive to cyclic loading (e.g. loose sands and
overconsolidated clays). Ring shear tests are used for determining large displacement soil
behavior along an interface between the soil and structural element. For one-way cyclic loading
due to combined wind and wave loading, the cyclic strength may be higher for a normally-
consolidated (i.e. no stress history) clay than the static strength due to rate effects. Therefore,
use of the conservative static strength tests for design can reduce testing time. For peat soils,
ring shear tests with a low-humidified (dry) sample should be used. The typical number of tests
conducted should be 4 to 5 per soil layer (DNV-OS-J101 2004). Extra soil from sampling tubes
is generally used for classification purposes.
44
TX tests provide information on the stress-strain behavior of the soil. The various ways to
perform TX test first consolidate the specimen and then either shear the specimen in drained or
undrained conditions, using either an axial compression, axial extension, or constant volume
condition. The type of test chosen is dependent on the predicted soil behavior at the location of
interest (i.e. undrained compression below a footing in clay, or drained extension next to a
bulkhead with a sand backfill). For the undrained conditions, pore pressure measurements are
obtained through testing. Undisturbed clay samples are prepared using cutting tools while
maintaining the in situ water content as well as possible, or remolded clay samples are prepared
using standard remolding techniques. Silt and sand samples are reconstituted into the triaxial
mold by hand tamping or wet/dry pluviating to the target relative density. The specimen is
enclosed within a rubber membrane for both types of tests, and saturated with water (for sand
specimens) before it is subject to a consolidation pressure of various methods (i.e. isotropic,
anisotropic, Ko, SHANSEP, back pressure/saturation). After reaching primary consolidation, the
specimen is sheared at a rate ranging from 1-µm/min to 1-mm/min. The TX test can be
performed under a static load condition, or under cyclic loading where load control limits are
imposed with a frequency dependent on the problem of interest.
DSS tests are performed using a specimen that is consolidated to Ko conditions and subjected to
a horizontal shearing load under drained or undrained conditions. The test provides stress-strain
behavior and shear strength characteristics for the soil along a forced plane of failure. The
specimen can be extruded and oriented in the device such that representative soil behavior can be
obtained along any angle relative to the horizontal plane. The specimen preparation and
consolidation phases of the test are similar to the TX test, however, only one-dimensional
volumetric change in the vertical direction is allowed. Static or cyclic loading can be conducted,
with displacement rates ranging from 1-µm/min to 2-mm/min, and cyclic load frequencies
ranging from 0.01-Hz to 0.5-Hz.
4.8.7 Other Laboratory Testing
Resonant column tests provide dynamic soil behavior characteristics such as shear modulus and
soil damping. During cyclic loading, the soil can reduce or increase its shear modulus through
volumetric contraction or dilation, respectively. While this test is conducted at higher
45
frequencies than the cyclic DSS test and is used primarily for earthquake load investigations, it
can provide useful information about the soil response not obtained through DSS testing alone.
Piezoceramic bender element tests measure shear wave velocity through a soil specimen. This
parameter can be used to calculate the maximum shear modulus, Gmax, and can be conducted at
any phase of a TX, DSS, or consolidation test. The test is performed by inserting small
piezoceramic probes at each end of a soil sample through which an electric charge is applied,
resulting in a measurement of the time for a shear wave to travel over a fixed distance. The wave
measurements are obtained through use of an oscilloscope.
Thixotropy tests measure the increase in undrained shear strength of a remolded clay specimen
over time. The clay is remolded under the in situ water content and tested using the fall cone
apparatus to measure the shear strength at increasing time intervals. The test is useful for
predicting the long-term internal shear strength characteristics of clay subjected to high
disturbance during foundation installation.
Heat conductivity tests provide information on the thermal characteristics of an undisturbed or
remolded soil sample. Using electric current applied to a long needle inserted into the specimen,
a time-temperature curve is generated to analyze the thermal conductivity, λ.
Other tests that can be performed on samples to obtain pertinent data include plane strain tests,
high pressure and variable temperature TX and consolidation tests, laboratory model tests,
geological and geochemical tests to determine geological origin and sedimentary history,
mineralogical or fossil analysis, amino acid or isotope analysis, gas analysis, carbon dating,
organic content assessment, and assessment of corrosion potential. For corrosion risk
assessment, offshore techniques include sulphide analysis, resistivity measurements, pH
measurements, coupled with color and odor description and sulphur and organic content analysis.
Onshore techniques include a more thorough sulphur-based compound analysis than the offshore
technique.
46
4.9 Effects of Cyclic Loading The effects of cyclic loading on the soil stiffness should be considered for soils prone to cyclic
strength degradation (e.g. loose sands, calcareous soils, overconsolidated clays). Cyclic loading
is caused by repeated wind and wave forces, and should be investigated for single storms or
several successive storms, as relevant. Cyclic loading may lead to a gradual increase in excess
pore pressure and a decrease in effective stress in undrained conditions (fine-grained soils), or
the degradation of shear strength due to permanent volumetric strains in drained conditions
(coarse-grained soils). For low permeability soil deposits, storm events causing excess pore
pressure can be superimposed on another storm event, potentially doubling the excess pore
pressures and subsequent strength reduction. However, such events can result in
overconsolidation of the soil over time and hence a higher resistance to the excess pore pressure
generation (Poulos 1988). In SLS design, the shear modulus G should be corrected when
dynamic motions, settlements, and permanent horizontal displacements are expected (DNV-OS-
J101 2004).
4.10 Seafloor Stability
4.10.1 Introduction
There are generally three mechanisms in the continental shelf region that can produce seafloor
instability and movement: gravity forces, hydraulic forces, and earthquake forces. Gravity forces
primarily provide a mechanism for slope failure classified as either basic instability or creep
phenomena. Basic instability is characterized by rapid, large displacement movements ranging
in time from minutes to days. Creep phenomena is characterized by widely variable strain rates
dependent on stress levels and environmental conditions, and occur over time periods ranging
from hours to thousands of years. Hydraulic forces result from currents that cause erosion and
scour in near surface deposits, tides that affect soft sediment stability in the intertidal zone,
surface waves which induce pressure anomalies along the seafloor, and internal waves that result
from density stratification of sea water from temperature and salinity variations (Poulos 1988).
Earthquake forces are caused by a sudden release of strain energy from a fault, which is a
function of source mechanism and regional seismicity.
47
4.10.2 Slope Stability
Slope stability can be analyzed using limit equilibrium, continuum mechanics solutions, or finite
element analyses, all of which usually employ plane strain conditions for simplification
purposes. The slope stability should be calculated for natural slopes, slopes developed during
and after installation of the structure, future anticipated changes of the existing slopes, the effect
of continuous mudflows, and wave-induced and earthquake-induced soil movements (where
applicable). Three soil conditions which should be investigated include undrained, fully drained,
or partially drained conditions. Soils most susceptible to slope failure are soft, normally
consolidated clays or loose sands. The safety against slope failure for ultimate limit state design
uses the material factor γm equal to 1.2 for effective stress analysis and γm equal to 1.3 for total
stress analysis. For accidental limit state design, γm is equal to 1.0 (DNV-OS-J101 2004).
4.10.3 Hydraulic Stability
Soils susceptible to erosion or softening need risk assessment for the reduction of bearing
capacity due to hydraulic gradients and seepage forces, formation of piping channels within
internal soil erosion, and surface erosion in local areas beneath the foundation due to hydraulic
pressure variations from environmental loads. Measures should be taken to prevent, control, or
monitor erosion as relevant.
4.10.4 Earthquake Stability
An earthquake releases energy in the form of stress waves, which are categorized into
compression (P) waves, shear (S) waves, Rayleigh waves, and Love waves. The effects of these
waves on onshore structures are similar to their offshore effects, with additional consideration
given to the following: 1) the earthquake force will be acting in possible combinations with wave
and wind forces, 2) the presence of water changes the ground motion characteristics, 3) flow
slides occurring from earthquakes can travel further in offshore environments (e.g. up to several
kilometers (Poulos 1988)), and 4) P-waves can cause increases in wave loading on an offshore
structure in the form of a tsunami. Evaluation of earthquake stability should include limit
equilibrium of the slope stability, in which earthquake loads are represented by equivalent
vertical and horizontal loads, and liquefaction or cyclic mobility potential analysis, which
48
consists of comparison between the equivalent cyclic shear stresses incurred by the loading and
the equivalent cyclic shear stress resistance of the soil.
The assessment should be based on previous records of earthquake activity in terms of frequency
of occurrence and magnitude if available. If no seismic information is available, a detailed
investigation of geological history and regional seismic events should be conducted.
If the region is seismically active, the following parameters should be determined:
1) Location and alignment of faults
2) Epicentral and focal distances
3) Source mechanism of energy release
4) Source-to-site attenuation characteristics
5) Local soil characteristics including stiffness and damping
6) Design earthquake or maximum credible earthquake
7) Tsunami (seismic sea wave) assessment (based on water depth)
The seismic analysis can be conducted using pseudo-response spectra, specifically the response
spectral acceleration, SA, the response spectral velocity, SV, and the response spectral
displacement, SD. For lightly damped structures, SA ≈ ω2SD, and SV ≈ ωSD, where ω is the
natural circular frequency of the structure. The earthquake-induced ground motions are then
analyzed in two horizontal directions and one vertical direction. For dynamic analysis, the wind
turbine can be represented by a concentrated mass on the top of a vertical rod, typically equal to
the mass of the rotor-nacelle assembly and ¼ of the tower mass (DNV-OS-J101 2004). Site-
specific dynamic properties of soils subject to cyclic earthquake loads should also be evaluated.
The ground motion characteristics should be evaluated at the structural base for free field and
local soil conditions.
4.11 Scour
4.11.1 Scour Mechanism
Scour is the movement of seafloor sediments due to currents or waves, occurring via natural
geological processes or an interruption in the natural flow regime from structural elements. The
49
main driver of scour is the oncoming water flow in the form of a horseshoe-shaped vortex in
front of the foundation structure, carrying sediment away from the foundation. The lower
velocity flow behind the structure may allow sediment to accumulate, creating a zone of
sediment higher than the original seafloor elevation.
4.11.2 Types of Scour
There are three basic types of scour: local scour, global scour, and overall seabed movement
(Figure 10). Local scour is characterized by conical scour pits around piles or pile groups similar
to those observed in flume models (Gravesen 2004). Global, or dishpan, scour is characterized
by large shallow basins around the foundation structure, resulting from overall structure effects,
multiple structure interaction, or wave-soil-structure interaction. Overall seabed movement of
sand waves, ridges, or shoals occurs naturally in the absence of a structure and can result in
lowering or rising of the seafloor elevation. Most scour observed is usually a combination of the
above characterizations.
Similar to slope stability, the soils most susceptible to scour are loose sands and soft clays.
Scour is most prevalent in areas with strong tidal streams or wave breaking zones. The affects of
scour are specific to the foundation type around which it occurs. Piled foundations suffer from
overburden loss and lateral resistance loss, while gravity base foundations may lose soil from
beneath the foundation, resulting in bearing capacity reduction, settlement, or overstressing of
the foundation elements (Watson 2000).
4.11.3 Prevention of Scour
The extent of scour can be determined from previous records of sites with similar seafloor
characteristics, from model tests, or from calculations based on prototype tests. There are two
options for addressing scour. One option is to allow for scour in the foundation design, assuming
all materials prone to scour are removed to a calculated depth below the original seafloor. In the
absence of sufficient data, an assumed depth of scour for piled foundations can range from 1 to
2.5 times the pile diameter, which should be verified by regular monitoring (Gravesen 2004).
The other option is to place scour protection around the foundation, typically consisting of a
thick (e.g. 0.5-m), crushed rock protection bed or asphalt/concrete mattresses. Placement should
50
occur as soon after construction as possible, monitoring the seabed frequently and replacing lost
materials as needed. The most inexpensive option is typically the crushed rock bed (van der
Tempel 2002).
4.11.4 Designing for Scour
Scour risk should be considered in design unless evidence can be demonstrated that scour will
not occur for the expected range of water particle velocities. The design for scour must account
for steady current, waves, and current or waves in combination as relevant, and design wind and
wave loading with a recurrence period of 100 years, with inspection every 5 years (DNV-OS-
J101 2004). The scour protection should provide both internal and external stability (i.e.
transport of particles in underlying natural soil and excessive surface erosion, respectively).
Both global scour and local scour should be considered in design, each having implications for
construction of the p-y (i.e. lateral resistance versus depth) and t-z (i.e. shaft friction versus
depth) curves. For global scour, effects of soil removal over a large area are estimated by
constructing p-y and t-z curves on the basis of a lowered seafloor level equal to depth of general
scour. For local scour, construction of the p-y and t-z curves are based on the original seafloor
level, assuming no soil resistance within the depth of local scour on the curves (DNV-OS-J101
2004).
Scour will lead to complete loss of lateral and axial resistance down to the depth of scour below
the seafloor, as well as a reduction in the effective overburden stress, which reduces the
resistance of the lower soil layers. The depth of overburden reduction is typically 6 times the
pile diameter (van der Tempel 2002). As the length of the pile will effectively lengthen, pile
flexibility increases resulting in decreases in the natural frequency of the pile. This effect is the
driving factor in fatigue loading on the structure.
An example flowchart for determining the occurrence of scour is shown in Figure 11. On the
left-hand side of the flowchart, a maximum shear stress on the seafloor is obtained from the
wave and current particle velocities. On the right-hand side, the seafloor allowable shear stress
is determined. Accounting for amplification effects of the structure, the scour potential is
51
obtained from comparison of the stress loading to the stress resistance. Once the potential for
scour is established, the depth of the scour can be calculated from numerous empirical equations
(e.g. DNV-OS-J101 2004, van der Tempel 2002).
52
5.0 Other Environmental Conditions 5.1 Introduction
Other offshore environmental parameters impacting wind farm design, such as meteorological
and oceanographic measurements, should be characterized for proposed wind farm sites for a
period of approximately one year prior to development. There are several methods of assessing
these parameters as outlined in Table 12. The oceanographic parameters to evaluate include
wind speeds, wind directions, strength of wind gusts, and turbulence. Meteorological parameters
of importance include the wave heights, tides, currents, and water turbidity. Other site
conditions to evaluate include human factors such as fishing areas, shipping route proximity, and
other restricted areas, and environmental considerations such as marine growth potential,
migratory bird routes, or fauna reproduction areas.
5.2 Meteorological Parameters
5.2.1 Wind Loading
The wind force acting on an offshore wind turbine structure contributes approximately 25% of
the horizontal load and 75% of the overturning moment to the foundation (Byrne and Houlsby
2003). Wind loading on the structure is represented by the 10-min mean wind speed, U10, and the
standard deviation of the wind speed, σU. The short-term 10-minute stationary wind climate is
represented by the power spectral density function of the wind speed process, S(ω), which
describes how the energy of the wind speed is distributed between various frequencies. Local
wind statistics are used as the basis for long and short-term wind condition representation where
available, while empirical data from other sources must cover sufficient time period (e.g. at least
1 year). The same averaging period should be used for the statistical basis of long-term
distributions of U10 and σU (e.g. 10 minutes). Otherwise, gust factors can be used. Hub height
wind speeds should be used as the reference value; otherwise a wind speed profile above the still
water level may be constructed. The characteristic values of the wind speed should be
determined accounting for inherent uncertainties and wake effects due to other turbines upstream
in a wind farm (DNV-OS-J101 2004).
53
5.2.2 Wind Modeling
Existing wind farm data shows that offshore wind energy output is 20 to 30% larger than
traditional wind modeling methods estimate (Krohn 2002). The first step in constructing a wind
model is computing the spectral wind density, represented by the Kaimal spectrum:
( ) 3/5
10
102
61
4
+
=
UfL
UL
fSk
k
UU σ Eq. 5-1
where:
f = frequency
Lk = integral scale parameter equal to 5.67z for z < 60m and 340.2m for z ≥ 60m
where: z = height above sea water level
Full characterization of the wind speed probability distribution should be obtained for 0 to 200-m
above sea level, referenced to a height of 10-m (Sahin 2004). This requires knowledge of
upstream roughness length (zo), which is a function of the distance from shore and the height
above sea level. Surface roughness is based on the rate of growth of internal boundary layers
from surface discontinuities and assumptions of the shape of the wind profile. The wind profile
can be described by the power law within layers, where growth of an internal boundary layer
height is based on a function of zo and stability (Sahin 2004).
5.2.3 Turbulence
Turbulence is created in a wind farm due to wake effects from other upstream turbines. Wake
effects should be accounted for when turbines are spaced less than 20 rotor diameters apart
(Frandsen and Thogersen 1999). This is done through a reduced U10 value relative to the
ambient wind climate inside the wind farm perimeter, or the standard deviation σU of the mean
wind speed. Depending on the wind farm configuration, there may be multiple turbines
contributing to wake effects upon an adjacent turbine. Wake effects usually dominate fatigue
loads on turbines, and fade out more slowly and over longer distances offshore than on land.
The turbulence intensity, I, is defined as the standard deviation of wind speed divided by the 10
minute mean wind speed:
54
10U
I Uσ= Eq. 5-2
Wake effects are the main factor in decreasing the fatigue lifetime of the mechanical components
of an offshore wind turbine due to the constant fluctuations in wind speed and direction.
Offshore turbulence intensity is lower than it is onshore, and increases with the height above sea
level. Therefore, offshore turbines should optimize design between lower hub heights and the
associated smaller rotor length required. Based on the fatigue loading, the expected lifetime for
an offshore wind turbine is estimated to be between 25 and 30 years compared to an onshore
wind turbine lifetime of 15 to 20 years (Sahin 2004).
A deterministic fatigue analysis should be applied to structures with natural periods different
from the significant energy wave periods which also exhibit little dynamic amplification.
Dynamically-sensitive structures should evaluate fatigue using a spectral analysis fatigue
method, which provides a statistical assessment of the wave spectrum-based (e.g. Pierson-
Moskowitz or JONSWAP) structural response.
For conceptual design studies, methods that combine time domain and spectral analyses into
equivalent fatigue loads can be used for the acceptable level of accuracy necessary at that stage.
The time domain simulation is conducted for a calm sea state to approximate aerodynamic
fatigue loading, and a linear spectral analysis is used to approximate hydrodynamic fatigue
loading. These two components are superimposed rather than added together due to the high
non-linearity of the stress-induced fatigue damage (Watson 2000).
5.3 Oceanographic Parameters
5.3.1 Wave Loading
The wave loads acting on an offshore wind turbine structure affect both the sub-structure and the
foundation. The wave loads can be represented by two main parameters, the significant wave
height, Hs, and the spectral peak wave period, Tp. The significant wave height is a measure of
the intensity of the wave climate accounting for wave height variability. It is traditionally defined
as the mean height of the 1/3 highest wave, H1/3, but can also be defined as four times the
standard deviation of the sea elevation process (i.e. four times the area under the wave spectrum,
55
Hm0) (Watson 2000, DNV-OS-J101 2004). The wave spectrum describes the frequency content
of a sea state, typically based on a Pierson-Moskowitz spectrum for a well-developed sea state or
a JONSWAP spectrum for a limited fetch and duration sea state. The spectral peak wave period,
Tp, is related to the mean zero-crossing period, Tz, of the sea elevation process and is assumed to
be constant (approximately 10-sec) over a short-term 3 to 6-hour sea state (Byrne and Houlsby
2003). The short-term 3 to 6-hour sea state is represented by a wave spectrum, or the power
spectral density function of the sea elevation process, S(ω), which is a function of Hs and Tp and
describes how the energy of the sea elevation is distributed between frequencies.
Hydrodynamic loads should be determined by theoretical predictions and model or full scale
measurements. The degree to which each is conducted is dependent on the uncertainty of the
predictions. Wave statistics can be used to estimate these parameters given that they cover a
sufficient period of time. If using data from a nearby area, consideration of the differences in
water depth and seafloor topography should be taken into account. The Hs and Tp distributions
should be based on the same averaging period for the waves as for the load determination;
otherwise, use of appropriate adjustment factors is necessary. Simultaneous values of Hs and U10
should be obtained since waves are wind-generated and should be correlated in design. The
directionality of each should be recorded as well.
5.3.2 Wave Modeling
Wave load predictions should account for the size, shape, and type of the proposed structure.
For piled foundations, Morison’s equation can be used to calculate the wave loads, consisting of
a drag force component and an inertial force component (Watson 2000). The drag force is
proportional to the overall combined water particle velocity, V, according to the following:
dLDVCdF dd2
21 ρ= Eq. 5-3
where:
dFd = drag force component
ρ = water mass density
D = tower diameter
dL = elemental length of tower
Cd = drag coefficient
56
The inertial force is proportional to the acceleration of the water particles due to Froude-Krylov
forces and added mass forces according to the following:
( )rai aCaAdLdF += ρ Eq. 5-4
where:
dFi = inertial force component
A = cross-sectional area of tower
a = water particle acceleration
Ca = added mass coefficient
ar = relative acceleration of water particles/tower
Diffraction effects may alter the wave pattern and loading for support towers that are large
compared to the wavelength, typically significant when the tower diameter is greater than 20%
of the wavelength (Watson 2000). At this point Morison’s equation is no longer valid, and the
inertial force will dominate. For gravity base foundations, radiation analysis or diffraction
analysis should determine wave loads (DNV-OS-J101 2004).
Where steep wave crests are prevalent, the support tower and sub-structure may be subjected to
highly localized impact loads, or slap forces, which are related to the rate of change of added
mass. For monopile foundations, this can be important for structural design as the slap forces are
not distributed as well as would be in a multiple-leg foundation. If wave steepness is assumed to
be constant and the wave height is scaled to the water depth, then the drag and slap forces are
proportional to the product of the squared water depth times the diameter, and the inertial forces
are proportional to the water depth times the squared diameter (Watson 2000).
Viscous and potential flow effects should also be considered. Waves that break against the
structure are more prevalent in shallower waters where wind turbines will be located than typical
water depths for larger offshore platforms, and therefore should be considered separately from
non-breaking wave loads. There are three classifications to consider: surging, plunging, and
spilling waves. These waves can result in large amplification factors depending on the wave
frequency relative to the natural frequency of the structure.
57
A generic distribution, or scattergram, can represent the long-term probability distributions of the
significant wave height, Hs, and the spectral peak wave period, Tp, consisting of a Weibull
distribution for Hs and a log-normal distribution of Tp.
In deeper waters, the short term probability distribution function of an arbitrary wave height H is
assumed to follow a Rayleigh distribution in terms of Hs. The maximum wave height in a 3-hour
sea state can be estimated to be equal to 1.89(Hs). In shallow water, the wave height is limited
by depth, and the maximum height can be assumed to be equal to 78% of the water depth (DNV-
OS-J101 2004).
The long-term probability distribution of an arbitrary wave height H is found by integration over
all significant wave heights, which is used to calculate the distribution of the annual maximum
wave height Hmax. Wave crest height Hc can be assumed to be equal to 0.65(H) (DNV-OS-J101
2004). The wavelength, λ, is given by the following equation:
λπ
πλ dTg 2
22= Eq. 5-5
where:
T = period
g = acceleration of gravity
d = water depth
Analytical and numerical wave theories can represent the wave kinematics according to their
ranges of validity. The linear wave theory that represents waves with a sine function is valid for
d/λ ≥ 0.3. Stokes’ wave theory for high waves and the stream function theory are valid for 0.1 ≤
d/λ ≤ 0.3, and the solitary wave theory for very shallow water is valid for d/λ ≤ 0.1. The Airy
theory is valid for all ratios of water depth to wavelength (DNV-OS-J101 2004).
Recent studies indicate that traditionally-modeled wave loads (i.e. via Wheeler stretching of the
pressure and velocity profile) are not adequate for extreme design cases, specifically in the range
of 0.25 < Hs/h < 0.5, where h is equal to the water depth. A new Boussinesq model has been
proposed that can generate a time series of the wave kinematics to define the input for the
foundation, leading to calculation of the wave forces through integration of the inertial and drag
58
forces over the length structure. The results are scaled to the water depth through Froude’s
scaling law, keeping the ratio of significant wave height to water depth constant. Because of this,
both operational and fatigue extreme loads can be accounted for through accurate
characterization of combined wind and wave forces (Gravesen and Bro 2003).
5.3.3 Current Loading
The current load consists of two to four components, depending on water depth and geographical
location: wind-generated current, tide-generated current, breaking waves (for shallow water), and
ocean circulation (e.g. Gulf Stream currents). The waves and currents are assumed to be
statistically independent (Watson 2000). The wind and tide-generated currents can be
represented by current velocities, which vary with water depth (DNV-OS-J101 2004).
5.3.4 Current Modeling
The current velocity can be estimated based on water depth according to the following:
( ) ( ) ( )zvzvzv windtide += Eq. 5-6
( )71
0
+
=h
zhvzv tidetide Eq. 5-7
( )
+=
0
00 h
zhvzv windwind Eq. 5-8
where:
v(z) = total current velocity at level z
vtide(z) = tidal current velocity at level z
vwind(z) = wind-generated current velocity at level z
z = distance from still water level, positive upwards
vtide0 = tidal current at still water level
vwind0 = wind-generated current at still water level
h = water depth from still water level (positive value)
h0 = reference depth for wind-generated current
Estimations should account for the variation in current profile due to the water depth variation
from vertical wave action by stretching or compressing the profile, keeping the surface current
59
constant regardless of the wave action-induced instantaneous sea level (e.g. DNV-OS-J101 2004,
Watson 2000).
5.3.5 Ice Loading
Ice loading should be considered in areas where ice may develop or drift. Characteristics of ice
loading include the geometry and nature of the ice, the concentration and distribution of the ice,
type of ice (ridges, floes, or rafts), mechanical properties of the ice, direction and velocity of
drifting ice, ice thickness, and the probability of an iceberg encounter. The build up and fracture
of the moving ice should also be considered. Other ice loads that may need consideration
include loads due to rigid ice coverages, masses of ice frozen to the structure, pack ice and ice
wall pressures, thermal ice pressures, falling ice, and icing loads (DNV-OS-J101 2004).
Ice build-up on the turbine may be due to sea spray, rainfall, snowfall, or air humidity, which can
change the cross-sectional area of the structural elements or change the surface roughness. For
floating structures, the uneven distribution of ice or snow accumulation should also be
considered.
The structure should be designed for both static and dynamic horizontal and vertical ice loads,
assuming that horizontal ice loads act concurrent with the wind load direction. The water level
that produces the most unfavorable reaction from the structure when calculating ice loads should
be used for design (DNV-OS-J101 2004).
5.3.6 Ice Modeling
The frost index, K, is used to determine the ice loading, defined as the sum of the daily mean
temperature over all days whose mean temperature is below 0° C in one calendar year (DNV-
OS-J101 2004). It varies from year to year and can therefore be represented by a Weibull
probability distribution function and a recurrence period. The ice compressive strength, bending
strength, and thickness can be expressed as functions of the frost index or as probability
distribution functions. The following ice parameter values may be used regardless of location
(DNV-OS-J101 2004):
Density 900 kg/m3
60
Unit weight 8.84 kN/m3
Elastic Modulus 2 GPa
Poisson’s ratio 0.33
Ice-ice frictional coefficient 0.1
Ice-concrete dynamic frictional coefficient 0.2
Ice-steel dynamic frictional coefficient 0.1
The characteristic local ice pressure from moving ice can be determined as follows:
local
cCuClocal A
err2
,, 51+= Eq. 5-9
where:
ru,C = compressive ice strength (≈ 1 to 2 MPa)
ec = ice thickness
Alocal = area of applied local ice pressure
During ice breakup, static and dynamic interactions between the ice and the structure occur,
causing the breakup frequency of the ice and the natural frequency of the structure to become
tuned together. Therefore, the structure should also be designed to withstand these load effects.
5.4 Combined Wind and Wave Loading
5.4.1 Horizontal to Moment Load Ratio
Load effects can combine to result in assumed intensities of multiple parameters acting during an
environmental state (i.e. when an intensity parameters acts at an assumed constant value over a
10-minute to 1-hour period of time). Combined horizontal loads are generally equal to 1 to 5%
of the resultant moment created from wind, wave, and current loading due to the typical height of
the rotor-nacelle assembly (Byrne and Houlsby 2003). However, the ratio of moment to
horizontal load fluctuates rapidly with time, dependent on water depth, sea state, and wind
conditions. This loading scenario is atypical to that of other offshore structures due to their
larger size, where the load ratios remain relatively constant regardless of environmental
conditions. Because the wind and wave loading may not be coincident, the horizontal and
moment loads may also not be coincident. However, the most unfavorable structural response is
when wind and wave loads do act coincidently.
61
5.4.2 Combination Methods
Two methods to combine wind and wave loads are the linear combination method and the
combination by simulation. The linear method simply combines the calculated wind load effect
and the calculated wave load effect by linear superposition. It works well as a preliminary
combined load evaluation or when dynamic effects are demonstrated to be negligible (e.g. in
shallow water) (DNV-OS-J101 2004). The equivalent damping resulting from the combined
loads should be assessed in terms of turbine position. The simulation method is based on
structural analysis in the time domain for the simultaneously-applied simulated time series of the
wind and wave loads. The dynamic interaction between wind and wave loads also may be
characterized by an amplification factor applied to the wind load that is proportional to the ratio
of the dynamic wave load to the associated wind load and the relative wave load deformation to
the 1st mode of the tower/foundation deformation (Gravesen and Bro 2003).
5.4.3 Design Considerations
The load effects can be evaluated based on the ultimate and accidental limit states in terms of
multiple combinations of environmental load type and its associated recurrence period. Each
load combination should be evaluated for the operational and the non-operational states of the
turbine, using the largest load effect for design. The characteristic load effect distribution for the
fatigue limit state should be considered in terms of the expected load effect distribution over the
design lifetime, which is a distribution of stress fluctuations from environmental loads that
include all phases of turbine operation (e.g. start up, shutdown, idling, transport, etc) (DNV-OS-
J101 2004).
5.5 Environmental Corrosion
5.5.1 Introduction
Corrosion for offshore structures can be divided into two categories: corrosion due to structural
degradation, and corrosion in the form of marine growth. There are three zones under
consideration for corrosion control; the lower splash zone, the buried zone, and the upper
submerged zone. The lower splash zone is the part of the sub-structure and tower intermittently
exposed to air and sea water, the lower limit of which is defined as one-third of the 100-year
wave height (DNV-OS-J101 2004). The buried zone is the part of the support structure below
62
the seafloor (i.e. foundation), and the upper submerged zone is the part of structure in between
the lower splash and the buried zones.
5.5.2 Degradation Corrosion
Steel structures require corrosion control to ensure structural integrity throughout their design
lifetime. Degradation corrosion in the lower splash zone is estimated to be approximately 0.5
mm per year, increasing with the age of the structure. It plays a significant role in fatigue
analysis and approximately 3 to 5 mm of reduced steel thickness should be included in the
analysis (DNV-OS-J101 2004). Corrosion protection systems in the splash zone should be based
on the corrosion allowance with a surface protection layer such as glass fiber-reinforced epoxy
coating. Steel structures in the submerged zone should be protected by a cathodic coating. Near
the seafloor, piles should include a corrosion allowance of 3-mm in the fatigue life endurance
limit, reduced by a factor of 2 in the buried zone to account for the anaerobic environment that
reduces the cathodic corrosive ability. Further details for cathodic protection and protective
coatings are given in DNV-RP-B401, and DNV-OS-C101, respectively (DNV-OS-J101 2004).
5.5.3 Marine Growth
Corrosion in the upper submerged zone and lower part of splash zone consists of largely of
marine growth, which can enhance or retard degradation corrosion attack. It can retard the
corrosion process through corrosive metabolites, called Microbiologically Influenced Corrosion
(MIC). MIC can occur in flooded steel members unless measures are taken to ensure water
remains sterile via addition of a biocide. It can enhance degradation corrosion by interfering
with corrosion control systems, such as coatings or linings or cathodic protection.
Corrosion in the buried zone is predominately related to MIC. In undisturbed sediments, MIC is
significant in uppermost layer only, but sediment disturbance through drill-cuttings or other
effluents can enhance bacterial activity and MIC (Gravesen 2004). Ice scoring in arctic
environments accelerates corrosion by removing rusting protection layers, corrosion protective
coatings, or marine growth. In deeper waters, oxygen content and sea currents are the
dominating parameters where minimal corrosion usually results (Gravesen 2004).
63
Marine growth can also increase the wave loads on the structure through an increase in the
effective diameter of the support structure by 0.1 to 0.2-m, depending on the depth below mean
sea level (DNV-OS-J101 2004). Marine growth should be accounted for by assuming an
additional insulation layer with a density of 2200 kg/m3 surrounds the foundation without
stiffness or damping capability (Zaaijer 2002).
5.6 Other Loading Conditions
5.6.1 Transportation and Installation Loading
There are several environmental parameters that may affect the transportation and installation of
offshore wind turbines. These parameters include the expected wind speed, wave height and
wave crest, water level, current, and ice for the duration of operations. Further details for
characterizing these parameters are given in the DNV Rules for Marine Operations (DNV-OS-
J101 2004).
Other installation loading considerations include the damage sustained by the soil due to pile
driving activities. The total fatigue damage in a piled foundation is derived from both the soil
damage occurring during the installation phase and the cyclic soil damage occurring during
operational phase. A pile drivability analysis should be conducted to verify that the foundation
can be installed to the target depth given the local soil conditions, and to determine the effects on
soil properties relative to measured values from in situ and laboratory testing results. Drivability
analyses can also help determine the type of equipment required for installation. The analysis is
conducted using a stress wave propagation dynamic analysis along the entire length of the pile,
and should be verified against real data for accuracy. For foundation elements that cannot be
inspected for long-term fatigue damage (i.e. below seafloor), a fatigue lifetime of the structure
can be assumed to equal three times the economic lifetime of the structure (van der Tempel
2002).
5.6.2 Vessel Collision
The loading and design of the wind turbine structure and foundation based on vessel collision
should be evaluated for the ULS and the ALS design criteria. For the ULS, consideration should
be given to the impact from the stem or stern of a characteristic vessel, while for the ALS,
64
unintended vessel collision by drifting vessels such as crane barges, working vessels, or
unauthorized vessels should be considered. The drifting speed should be assessed for each case
with a minimum speed of 2 m/s (DNV-OS-J101 2004). The structure and foundation should be
designed so that impacts from any of the above cases will not reduce their capacities. This
consideration is especially important in waters with high volume of commercial shipping traffic.
The assessment should consider the volume of shipping traffic near the wind farm or turbine,
identification of shipping lanes in the area, assessment of annual shipping traffic and of the types
and sizes of ships in the shipping lanes. An environmental assessment of wind, wave, current,
ice, and visibility that may affect collision frequency should also be considered.
Calculation of the risk associated with vessel collision is defined as the product of the frequency
or probability of a collision scenario (e.g. navigational error, disabled drifting ship, steering
failure, etc.) and the consequence of a collision (structural safety, human safety, environmental
safety). The total risk should be compared among wind farm locations for siting considerations,
and verified against historical data of ship-to-ship collisions and groundings in the area. Risk
reduction measures may need to be implemented such as additional structural reinforcement of
the foundation and/or the support structure.
5.6.3 Deformation Loading
Deformation loads should be considered when differential settlements impose strains on the
support structure that may lead to increased fatigue effects or vulnerability to other
environmental loads. This is particularly important for gravity base and multiple-leg
foundations. In the case of gravity base foundations, a slight rotation from the vertical will
induce an additional moment load onto the structure. In the case of multiple-leg foundations, the
differential settlement may cause structural elements such as trusswork to assume additional
stresses.
65
6.0 Foundation Modeling 6.1 Introduction
The dynamic excitation forces acting on an offshore structure consist primarily of waves and
current on the tower, sub-structure, and foundation, wind on the rotor-nacelle assembly and
tower, and the interaction between the structure and the rotor blades during operation. In design,
the natural structural frequencies need to be avoided to minimize the chance of resonance with
the excitation frequencies. To model dynamic loading, principles of fluid mechanics and ocean
wave theories are applied using deterministic methods to simulate extreme waves in randomly
generated sea states in shallow to deep water depths (Byrne and Houlsby 2003). In conjunction
with dynamic loads within the turbine structure, load transfer into the seabed soils induce a
dynamic response known as soil-structure interaction.
Soil-structure interaction (SSI) evaluates the structural load effects on the integrated soil and
structural system, using realistic assumptions for stiffness and damping of the soil and the
structural members. To design and analyze offshore wind turbine structures, a coupled soil-
structure analysis including extreme events and fatigue loading is recommended (Feld and
Waegter 2002). Different foundation types have specific models on which to base analysis. For
piles, axial and lateral resistance of the soil is typically modeled with non-linear springs, as
shown in Figure 12. The non-linear behavior of axial and lateral pile-soil resistance can be
accounted for by ensuring the load deflection compatibility between the structure and pile-soil
system and accounting for the effects of geometrical and material non-linearities. Under the
commonly-used Winkler assumption, empirically-derived nonlinear springs and dashpots each
act independently of the springs and pile displacements at other locations (Kellezi and Hansen
2003). Although this assumption simplifies the behavior, it is established as a reliable method to
model dynamic SSI.
The lateral resistance stress-strain relationship is represented by the non-linear p-y curve, which
varies for different soil types, pore pressure conditions, and static versus cyclic loading
conditions. At higher displacements, cyclic loading produces lower resistance values than for
static loading. The shaft friction stress-strain relationship is represented by the non-linear t-z
curve, which uses dimensionless tabulated values of skin friction that also vary for different soil
66
types. The pile tip resistance relationship is represented by the non-linear q-z curve, which does
not vary with soil type (Zaaijer 2002).
For gravity base foundations, uncoupled springs, dashpots, and inertia are used to represent the
soil-structure interaction model, as shown in Figure 13. Elastic half-space models are used to
obtain the values of these elements, where radiation damping and spring and dashpot parameters
are frequency-dependent. Modeling should account for the strain-dependency of the shear
modulus and internal soil damping by using parametric studies including uncertainties in soil
properties with upper and lower boundaries on shear moduli and damping ratios of the soil. For
wind turbine structures, the dependency on frequency simplifies to allow for estimation of the
model parameters through the equations shown in Table 21, where D is the diameter of a circular
gravity base foundation, G is the shear modulus, and ρ is the soil density. For earthquake prone
areas, frequency dependent reductions of the stiffness may be necessary.
6.2 Types of Models Various modeling techniques have been used to model offshore foundations. In the following
sections, different methods will be compared showing the effect of each on dynamic sensitivity
for various foundation types typical of offshore wind turbines. Some of these models have been
proven through industry implementation, while others are still in the research and development
stage.
6.2.1 Plasticity Models
Yield surfaces can be modeled in vertical:moment:horizontal load space (V:M/2R:H), in which
the size of the surface is dependent on the vertical plastic displacement of a foundation element
(Figure 14). Recent advances in strain-hardening plasticity theories have been proposed that
provide more detailed information about displacements and the simulated loads that cause them
(Byrne and Houlsby 2003).
The foundation response can be expressed in terms of force resultants on the footing and the
corresponding displacements, consistent with the time-domain approach used for structures
which enables simultaneous modeling between soil behavior and structural analysis. Plasticity
67
models include four components consisting of the yield surface which defines allowable load
combinations, a strain-hardening expression that defines how the yield surface expands or
contracts, a flow rule that defines the plastic displacements at yield, and a model for the elastic
response within the yield surface. The rule of behavior in the model is such that if a load
combination is within the yield surface, an elastic response results; otherwise, plastic response
results as defined by the flow rule.
Results of a model study provided insight into cyclic soil-foundation interaction, such as
reduction in the stiffness of a hysteretic response as tension is applied to the foundation.
Multiple yield surfaces for the stiffness transition are each accompanied by a plastic potential to
describe the plastic flow, which can then be used in constitutive soil models. A disadvantage to
using this model is that the specific parameters, sometimes difficult to assess, must be specified
for each surface. However, the concept of continuous hyperplasticity, based on thermodynamic
principles, replaces plastic strain in conventional plasticity theory with a continuous field of an
infinite number of yield surface-specific plastic strain components. It is based on the derivation
of plasticity theory for dissipative material of two potentials, the Gibbs (Hemboltz) free energy
and the dissipation function. As indicated in Figure 15, this theory closely matches laboratory
behavior, and may prove to be a useful method of implementing plasticity models (Byrne and
Houlsby 2003).
6.2.2 Finite Element Models
Finite element models (FEM) that evaluate foundation behavior are composed of structural
elements for the foundation and soil elements for the surrounding seafloor. FEM analysis
accounts for initial conditions, nonlinear soil-structure interaction, and nonlinear soil behavior.
Boundary conditions determine the constraints for coupling of the structural and soil elements.
They are described using differential and integral operators of time and space developed through
local schemes that are independent of the frequency of excitations, making them applicable for a
time domain transient analysis. For static analysis, boundary elements are assumed to connect to
a rigid surrounding, whereas in dynamic analysis, radiation damping at the soil interface needs
consideration through multiple degree-of-freedom models. Under typical conditions, a Winkler
assumption is preferred. This assumption is unsuitable, however, for flexible gravity base
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foundations, geometrically-complex foundations, or foundations subjected to non-linear stress-
strain cyclic loading (Zaaijer 2002).
Many computer programs using FEM have been developed for the offshore industry. Examples
include ABAQUS, which uses different models such as the Mohr-Coulomb theory with soil
hardening/softening effects or a Drucker-Prager material model with a non-associated flow rule
(Kellezi and Hansen 2003), or Ramboll’s multiple FEM programs (e.g. ROSAP, RONJA,
ROSOIL) for the wind industry, which combine linear structures with nonlinear foundations.
Most of these programs automatically generate the range of environmental and structural loads,
in which any standard wave theory can be applied that comprise load situations in all limit states,
incorporating both elastic and plastic behavior of the soil in the design (Feld and Waegter 2002).
6.2.3 Other Techniques
The effective fixity length technique, which is based on the clamping effect of the soil
surrounding piles, can be modeled using a rigid restraint located at an effective depth below the
seafloor (Zaaijer 2002). Using approximate values of the effective fixity length (Table 22),
preliminary dynamic analyses can be conducted for offshore structures. Due to the lack of
bracing through a support frame as seen in typical offshore structures, monopile foundations
exhibit different mode shapes of the effective fixity model.
A stiffness matrix can be also used to represent the pile-soil stiffness at the seafloor, comprised
of forces, moments, displacements, and rotations of the pile head (Zaaijer 2002). The advantages
of a stiffness matrix include the consolidation of foundation properties helping facilitate
information exchange between the geotechnical and structural engineers for frequency
calculations. There are two methods for obtaining the stiffness matrix: a load-displacement
analysis with p-y curves, or the Randolph elastic continuum model (Zaaijer 2002), which is
based on the inverse of a matrix expression for pile head flexibility derived from dimensional
and finite element analyses of piles in an elastic continuum. The Randolph model is
parameterized in terms of both constant and linearly-increasing soil shear modulus, and therefore
works well for sandy soils.
69
6.3 Dynamic Sensitivity Due to the smaller size and loading nature of an offshore wind turbine structure, the dynamic
sensitivity may be more significant than for traditional offshore structures (Watson 2000). The
natural frequency of these structures lay within excitation frequencies that differ from traditional
offshore structures in that the natural frequencies lay well above the wave excitation frequencies
(Zaaijer 2002). The dynamic sensitivity of the structure to foundation stiffness should be
determined through natural frequency analysis to obtain the important mode shapes and
frequencies. If dynamically-sensitive, dynamic transient analysis using a detailed time history
loading should be used to evaluate the structure.
For increasing depths of foundation penetration, the natural frequency will achieve a maximum
value where the lower portion of the pile does not move, similar to the effective fixity length
principle. This value varies for foundation type between monopile, tripod, or lattice tower
structures. Based on this analysis, a minimum dynamic sensitivity of the natural frequency can
be expected for an offshore wind turbine structure, regardless of foundation penetration.
A parametric sensitivity analysis for natural frequency dependency was conducted at the Delft
University of Technology based on a 3 MW wind turbine in 21-m of water depth with uniform
soil conditions. This analysis assessed the influence of parameter uncertainty, lifetime parameter
change, and local parameter variation (e.g. within wind farms) (Zaaijer 2002). Parameter
variations considered included soil, foundation, environmental, and structural variations. The
sensitivities for the various foundation/structure configurations are summarized in Table 23. As
indicated, soil parameters dominate the uncertainty of the natural frequency for all categories.
The sensitivity of gravity base foundations to soil parameters is higher than for piled
foundations, lending to the presumption that gravity base foundations cannot be uniform within a
wind farm, but rather designed for local soil conditions at the location of the foundation within
the limits of the wind farm. For structural considerations, the rotor-nacelle assembly had the
largest effect on natural frequency, with other parameter variations resulting in frequency
changes of less than 0.2%.
The TUDelft study also examined the variation of natural frequency with foundation models (
70
Figure 16). As shown, the finite element and linear elastic models gave comparable results,
resulting partially from the soil uniformity. Randolph’s method was less suitable for piles
shorter than a critical pile length and for non-linear foundation deflections. The suitable value of
the effective fixity depth for monopiles, tripod, and lattice structures differ due to dependencies
on vibration mode shapes, pile stiffness, and soil properties. The stiffness matrix corresponded
well with the finite element model, indicating that the stiffness matrix may replace the use of
excessive computations from FEM in certain cases. For the tripod and lattice tower
configurations, lateral flexibility was paramount in comparison to axial flexibility, mainly
dependent on foundation stiffness, pile spacing, and the distribution of mass between the tower
and the rotor-nacelle assembly.
A comparison of these calculated natural frequencies were made to measured values for five
turbines at the offshore wind farm Irene Vorrink. All values were within expected deviations
form measured values, with the largest deviation being 5.3%. A comparison to the wind farm
Lely resulted in a difference of 9% observed for one turbine, and a 30% difference for another
turbine, which could not be explained without comprehensive re-assessment of model
parameters. Comparisons of natural frequency and soil properties showed no changes over a 6
year period at the Lely farm (Zaaijer 2002).
71
7.0 Foundation Design 7.1 Introduction
The loading regime of the offshore wind turbine foundation is unique to offshore structures in
that the weight of the turbine structure is low compared to the overturning moment and the
horizontal load. It is critical that the foundation can sustain extreme loading conditions via any
combination of environmental loads. When designing the foundation for an offshore wind
turbine, it is important to keep in mind that since wind turbine farms contain numerous turbines,
a single design for the entire farm is necessary to enable mass production and ensure expedient
installation, both of which are necessary for the economic feasibility of a wind farm.
The foundation should be designed in accordance with local, national, or international standards
and regulations relevant to the site of the wind turbine. While there exist well-established design
codes for offshore construction (e.g. ISO 19900 to 19903 Design of Offshore Steel and Concrete
Structures, API Guidelines for Piles, etc.) that provide a good basis for offshore design, these
codes generally do not cover the difficult design aspects such as layered soils or cyclic load
effects, which should be investigated on a case-by-case basis. In preliminary design
calculations, the number of load cases to consider can be reduced using load case lumping. This
is a common offshore design technique that decreases the overall design effort by constructing a
small number of lumped load cases to represent the full range of environmental characteristics at
a site (Watson 2000).
The choice of the foundation type should be based on site-specific information, including the
adequate characterization of the soil conditions, water depth, scour and erosion potential, turbine
capacity, foundation cost, and the environmental loading conditions. The design process must
consider both the strength and the deformation characterization of the surrounding soils. The
primary soil strength failures that can occur in an offshore environment include bearing capacity
failure, sliding failure, and pile pull-out and punch-through failure. The primary deformation
failures that can occur include large settlements or lateral displacements.
Limit states categories are applied to foundation design where cyclic loading failure is treated as
an ULS condition using partial and material load factors (DNV-OS-J101 2004). For effective
72
stress analysis, the tangent to the soil’s friction envelope is divided by material factor, γm, while
for total stress analysis, the undrained shear strength is divided by material factor.
Ultimately, the foundation design will depend primarily on the cost of installation due to the
number of turbines in a typical wind farm, the in-service performance relative to repeated
loading, large overturning moments, serviceability limit states, and considerations in the
decommissioning and removal of the structure, and possibly the foundation. Because there is
currently no economically feasible way to remove piles in offshore environment, foundations
such as the suction caisson may soon become a reality in large scale wind farm development
(Watson 2000).
7.2 Offshore Turbine vs. Offshore Platform Foundations Much of the technology in use for offshore oil and gas structures can be applied towards the
design of offshore wind turbine structures. However, there are a few major differences that must
be considered, such water depth, loading condition, and the typical number of installations.
Water depths for traditional fixed offshore structures have been in the range of 20 to 120-m,
whereas wind turbines will be located in water depths in the 10 to 25-m range for the near future.
The loading conditions for the three critical aspects, namely vertical, horizontal, and moment, all
differ for offshore turbine versus offshore platform design (Figure 17). The vertical load for a
traditional offshore platform varies between 5,000 and 30,000-tonnes, whereas the wind turbine
structure load ranges from 100 to 300-tonnes. The horizontal load for traditional platforms is
typically 10 to 20% of the vertical load, while it is much more variable for turbines, ranging from
70 to 150% of the vertical load (Watson 2000). The moment load imposed on the platform is
calculated as the multiple of the water depth and the horizontal load, but for turbines it is equal to
the multiple of the water depth plus 50-m (or more) and the horizontal load, due to the height of
the rotor-nacelle assembly above mean sea level (Watson 2000).
Typically, the number of installations of a traditional offshore platform ranges between one and
four (depending on whether each pile group or gravity base is counted), but for an offshore wind
farm, the number of installations can be greater than 200, and are usually at least 25. Because of
the average number of installations occurring for a given project, the cost per structure must be
73
proportionately lower than single offshore installations to offset the high percentage of
foundation installation in total cost (Watson 2000).
Another aspect of offshore wind turbine design compared to offshore platform design is the
importance of the structure. The offshore wind turbine is unmanned, is usually not a single
supply source, has minor environmental impact in a failure state, and costs much less than the
offshore platform. Therefore, a minimal factor of safety for the structure is required (Vugts and
Harland 1997).
7.3 Piled Foundations
7.3.1 Introduction
Offshore piled foundations evolved from onshore designs using the large database of empirical
evidence and experience to draw upon. However, the current knowledge base does not account
for particulars for offshore piles including size, soil shear strength, and type of loading for
offshore wind turbines. For these cases, the following parameters are needed for design: 1) axial
capacity of piles in tension and compression, 2) load-deflection characteristics of axially and
laterally loaded piles, 3) pile drivability characteristics, and 4) mudmat (i.e. temporary
foundation that assists in alignment and leveling of pile foundation) bearing capacity (Gravesen
2004). Mudmat design is highly dependent on accurate slope characterization and assessment of
shear strength of the soft surface soil layers. These parameters are needed to assess stability for
both temporary and operational conditions (e.g. stability of drilled holes before pile placement).
7.3.2 General Design Considerations
Piled foundations should be examined in the following contexts: 1) elastic ultimate limit state,
where only one pile per foundation is allowed to reach yield point as a maximum, 2) plastic
ultimate limit state (accounting for cyclic-load strength degradation), where piles are allowed to
yield if still absorbing design loads, 3) fatigue in terms of both actual fatigue load on structure
and damaging effects of pile driving, 4) pile driving analysis, 5) eigenfrequency analysis (i.e.
characteristic representation of the dynamic response of the pile), and 6) soil damping estimate
(Gravesen 2004).
74
Factors that need to be considered include the shear strength characteristics, deformation
properties, in situ stress conditions, methods of installation, geometry and dimensions of the pile,
and the types of load to which the pile will be subjected. For the design soil resistance for axial,
lateral, and moment loading, the material factor for effective stress analysis shall be equal to 1.2
(ULS) or 1.0 (SLS), and for total stress analysis equal to 1.3 (ULS) or 1.0 (SLS). For axial loads
in ULS design, a material factor equal to 1.3 should be applied to all characteristic values of soil
resistance (e.g. skin friction, tip resistance) (DNV-OS-J101 2004). The design pile loads can be
determined from structural analysis in which the pile is modeled with adequate equivalent elastic
stiffness or with non-linear models reflective of the true stress-strain properties of the soil. For
drilled piles, limit skin friction assumptions must be verified during installation due to drilling
mud adhesion effects. Laterally loaded piles are analyzed based on realistic stress-strain curves
for the soil and pile, which may be inelastic due to the combined effects from axial loading. The
curves can be based on plasticity theory provided that characteristic resistance is in accordance
with recognized plastic theorems to avoid non-conservative safety estimates, or based on
calculations that assume that lateral deformations will completely plastify the soil.
7.3.3 Grouting Operations
Grouting can be used for securing piled foundations into the seafloor where bedrock socketing is
necessary, or where a monopile is connected to a transitional section for tower support.
Advantages of grouting include the ability to account for pile installation tolerances, increased
soil strength capacity, and higher fatigue resistance. Some important characteristics of
competent offshore grout include its pumpability (i.e. sufficient viscosity and inner cohesion),
low shrinkage potential, low thermal development during hardening, sufficient early strength
development, and resistance to fatigue. The grouting material used should be capable of
maintaining sufficient strength throughout lifetime of structure from all forms of deterioration,
including chemical, mechanical, and/or mixing/dilution problems (DNV-OS-J101 2004). A
Danish company Densit has developed Ducorit grout, applicable for offshore use. Ducorit is
comprised of various quartz and bauxite aggregate that allows its application as an active
structural component rather than simply filler material (J. Offshore Tech. 2002).
75
7.3.4 ULS Design
7.3.4.1 Monopile Foundations
For the ULS monopile design, soil strength values should be used which are equal to the
characteristic soil strength values divided by the material factor. The design loads are calculated
by multiplying the characteristic load by the specified load factor. The characteristic load is
representative of extreme load conditions for both axial loading and combined lateral loading
and moment loading. The axial resistance is calculated by integrating the design unit skin
friction over the surface area of the pile, adding possible pile tip resistance and subtracting the
weight of the pile. A general form of the equation is shown below:
( ) pbb
L
zsu WAfdzCfP −+= ∫
0
Eq. 7-1
where:
Pu = ultimate load capacity
fs = ultimate skin friction along length of pile
Note: A static unit skin friction reduction factor of 0.9 for sands, 0.72 for clays is
recommended (DNV-OS-J101 2004).
C = circumference of pile
L = embedded length of the pile
z0 = length of pile below the seafloor over which no skin friction is assumed to develop
fb = ultimate resistance of the pile base
Ab = gross area of the pile base
Wp = weight of pile
The combined lateral and moment resistance should be ensured based on verification of lateral
pile resistance through two requirements. 1) The first is ensuring that the theoretical design total
lateral pile resistance, which can be found by vectorial integration of the design lateral resistance
over the length of the pile, is not less than design lateral load applied at pile head. This
verification can be done using FEM analysis with soil support springs in terms of p-y and t-z
curves attached at the nodal points. 2) The lateral displacement of the pile head can be
calculated based on the soil resistance and soil stiffness, and it should not exceed limits specified
based on the allowable displacement of the structure.
76
7.3.4.2 Multiple-Leg Foundations
For ULS multiple-leg foundation design, sufficient axial capacity and resistance to lateral and
moment loading should be ensured for each pile by integrated analysis of the support structure
and the foundation piles subject to design loads. P-y curves should be generated according to
API RP2A code and DNV Classification Notes 30.4. T-z curves should be generated according
to DNV/Riso: “Guidelines for Design of Wind Turbines”, dependent on the degradation potential
of unit skin friction as mentioned previously (DNV-OS-J101 2004). The same reduction factors
as for monopile foundations should be used for cyclic strength reduction. For grouped piles,
additional requirements should be imposed as piles that are closely spaced may cause the
grouped pile resistance to potentially be less than the sum of individual pile resistance, and the p-
y and t-z curves should be adjusted accordingly. Also, the load transfer between closely spaced
piles can cause soil deformations in soils supporting adjacent piles, resulting in a potentially
softer pile group response. This may be accounted for by elastic half-space solutions for
displacements in a soil volume due to applied point loads.
7.3.5 SLS Design
7.3.5.1 Monopile Foundations
For the SLS monopile design, the characteristic loads will be representative of loads causing
permanent deformations of the foundation geometry (e.g. pile head tilt), and should account for
the cumulative effects of long term cyclic loading on permanent deformations due to combined
loads. It should be ensured that deformation tolerances, including tolerances for installation, are
not exceeded in the form of permanent deformations, usually specified as a maximum allowable
rotation of the pile head from the vertical plane based on visual and operational requirements of
the wind turbine.
7.3.5.2 Multiple-Leg Foundations
For the SLS multiple-leg foundation design, the same procedures are used as for the monopile.
Deformation tolerances should not exceed maximum specified pile rotations and horizontal
displacements of the pile heads. Separate tolerances may be specified to a pile group
immediately following installation versus the permanent cumulative displacements that occur
throughout the design lifetime of the structure.
77
7.4 Gravity Base Foundations
7.4.1 General Principles
There exists a similar database and pool of experience for gravity base foundations as for piled
foundations from the offshore oil and gas industry. For design purposes, the following factors
should be accounted for: 1) total stability failure, 2) rupture in soil carrying capacity, 3) sliding
ruptures, 4) combined ruptures in soil and structure, 5) ruptures due to foundation movements, 6)
unacceptable movements and oscillations, 7) eigenfrequency analysis, 8) liquefaction risk
analysis (when set upon sandy soils), 9) zones of local strong soil or rock, and 10) design of
gravel bed (differential settlement analysis, requirement to grading curve, leveling of gravel
base, base minimum thickness) (Gravesen 2004).
For sliding rupture potential, calculation of the passive earth pressure should include the
expected damage from scour and whether filling around the foundation is to be accounted for.
This should include horizontal forces, overturning moment, and torsion moments about the
vertical axis of the structure. Special attention should be given to any softening effects for
residual top clay layers. The eigenfrequencies are modeled as springs attached to the foundation
to demonstrate soil stiffness (see DNV Classification Notes 30.4) (DNV-OS-J101 2004).
The analyses should be carried out with consideration that the wave load period is approximately
10 to 20-sec for resonance analysis, and that the soil will consolidate under the weight of the
structure, creating initial excess pore pressure that may reduce strength parameters relative to
design values (Poulos 1988). For gravity-based foundations with skirts, the hydraulic instability
must be adequately guarded against for foundations which calculate tensile forces on differential
water pressure. When tremie grouting, the stability of the structure and the surrounding soil is
verified by comparing the volume of pumped concrete to the estimated volume of voids below
the foundation (Gravesen 2004).
7.4.2 General Design Equations
The two most common equations for design of gravity base foundations are Terzaghi’s general
bearing capacity equation and Vesic’s bearing capacity equation, which accounts for many
geometrical factors (Coduto 2001). Terzaghi’s equation for bearing capacity is as follows:
78
( ) foundationqcpeak AqNcNBNV ++= γ21 Eq. 7-2
where:
Vpeak = peak vertical load that can be sustained by the foundation
B = foundation width
c = cohesion for soil below foundation
q = overburden stress at depth of foundation
Afoundation = area of the foundation
Nγ, Nc, Nq = bearing capacity factors dependent on friction angle of the soil
Due to the complicated moment and horizontal loads acting on the foundation, the factors
included in Vesic’s equation allow for a more precise design, and is currently the standard for
bearing capacity calculation for homogeneous soil conditions in the offshore industry (Byrne and
Houlsby 2003).
( ) ( )
++=
γγσ q
cNsdibgBcq zDult '5.0'' Eq. 7-3
where:
qult = ultimate bearing capacity of the soil
c’ = effective cohesion for soil below foundation
σ’zD = effective overburden stress at depth of foundation
γ’ = effective unit weight of soil
N = bearing capacity factor dependent on friction angle of soil
s,d,i,b,g = factors accounting for shape (s), depth (d), load inclination (i), base
inclination (b), and ground inclination (g)
7.4.3 Stability of Foundation: ULS and ALS Design
The risk of shear failure below the foundation base should be investigated for all gravity base
foundations, covering all potential shear surfaces, and accounting for foundation geometry.
Special consideration should be given to soft soil layer and cyclic loading effects. For skirted
gravity foundations, the base dimension is assumed to be at the skirt tip level. The design should
79
be analyzed for site-specific soil drainage conditions (i.e. fully, partially, or undrained
conditions).
Stability is evaluated based on either total stress analysis or effective stress analysis for
laboratory shear strength and pore pressure measurements via stress paths. Effective stress
analysis is based on the effective strength parameters of the soil and realistic pore pressure
estimations dependent on the initial pore pressure, the build up of pore pressures due to cyclic
load history, transient pore pressure through each load cycle, and any effects of energy
dissipation. Total stress analysis is based on total shear strength parameters determined from
tests on representative samples subject to appropriate as-built stress conditions. Stability
analyses based on conventional bearing capacity formulae are only acceptable for uniform soil
conditions and relatively small foundation bases. The material factors to be used are the same as
previously used for ULS design, with ALS material factors equal to 1.0. Cyclic load effects
should also be incorporated through load factors, and the overturning safety is evaluated in the
ULS and ALS design. Both internal soil damping and radiation damping should also be
considered in the design (DNV-OS-J101 2004).
7.4.4 Settlements and Displacements: SLS Considerations
Analyses for settlements and displacements include initial consolidation, secondary settlements,
differential settlements, long-term permanent horizontal displacements, and dynamic motions.
The differential settlement analyses should account for lateral variations in soil conditions within
the foundation area, non-symmetrical weight distributions, predominating environmental load
directions, and liquefaction in seismically active areas.
7.4.5 Grouting Operations
Base grouting is performed on gravity base foundations to avoid further seabed penetration, to
retain vertical position, to ensure uniform stress distributions below the foundation, and to reduce
the potential for piping that occurs below the base during environmental loading. It is
particularly useful on sloping or uneven seafloors (Poulos 1988). Base grouting should not cause
filling pressures to create channeling from one skirt to another or to the outside of the structural
80
periphery. Grouting material properties should be similar to the requirements summarized
according to the piled foundations grouting operations.
7.5 Suction Caissons
7.5.1 General Principles
For suction caisson foundations, little experience and the lack of onshore precedent exist to base
design procedures on. However, the design procedure is understood for relatively simple cases.
More investigation is currently needed to describe the horizontal, overturning, and cyclic loading
response of these foundations (Watson 2000). Generally, the following factors should be
considered: 1) safe and even penetration of the caisson during installation, 2) plastic ultimate
limit condition, 3) operational limit condition, 4) fatigue analysis, 5) eigenfrequency analysis, 6)
and soil damping estimate. For suction-based installation, it must be demonstrated that the soil
penetration resistance is lower than suction driving force, and that the soil inside the caisson does
not dilate more than that induced by the displacement due to caisson penetration only. Particular
consideration must be given to scour around the caisson perimeter, as this zone is critical to the
uplift capacity of the caisson (Gravesen 2004).
7.5.2 Design Studies
A simplified analysis for a multiple-leg foundation design with suction caissons was conducted
by Byrne and Houlsby (2003) (Figure 18). There two geometrical aspects to the design that
varied were the foundation separation and foundation dimension. The foundation separation
established the critical case when the foundation rotated about the two downwind foundations.
Considering an overturning moment M consisting of net horizontal load H acting through height
h above the foundations, and a restoring moment consisting of vertical load V acting through
structural center of gravity, the minimum required spacing between foundations was estimated
accordingly:
)(
)(
3232
22
tripodquadruped
VM
VHy
VM
VHy
s
=
== Eq. 7-4
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The foundation capacity established the critical case when one foundation lay downwind. For a
quadruped foundation, the tensile capacity will be larger than for an equivalent tripod foundation
because there is always more tensile resistance from the upwind foundations. Given equivalent
conditions, a comparison between quadruped and tripod configurations is shown in Figure 19.
The increase in foundation size and mass for the tripod configuration was offset by the reduced
amount of steelwork for the structure, which may also be subject to a reduced wave and current
load.
The capacity is calculated based on conventional bearing capacity theory as described
previously. The results showed no difference in caisson size for the tripod versus the quadruped
design. Figure 20a shows the relationship between vertical load, caisson diameter, and caisson
spacing based on the quadruped design with a constant aspect ratio (i.e. length divided by
diameter) of 0.5.
According to the figure, as vertical load increases (e.g. through ballast addition) the required
caisson size reduces. As horizontal loading dominates, caisson size initially reduces,
transitioning to increasing caisson size as the vertical load becomes critical. The length-to-
diameter ratios are typically restricted in certain soil types and in shallow water depths. In sands,
an L/D ratio that exceeds 1 will typically cause piping failure to occur. In stiff clays, the
maximum L/D ratio will be 2.5 (5 for soft clays) due to the occurrence of a reverse bearing
capacity failure of the soil plug at the skirt tip from the suction pressure. For shallow water
depths, the cavitation limit of the water will be the limiting factor for clay (Byrne and Houlsby
2002).
From a design calculation standpoint, the multi-leg suction caisson foundations are a more
straightforward design than for single caissons, as vertical loads on the seafloor can represent the
moment loads applied to the structure. For mass production, however, the single suction caisson
foundation has the advantage due to simplicity of installation and lower structural steelwork
requirement.
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A simplified design analysis for a single suction caisson foundation was also conducted by Byrne
and Houlsby (2003). The vertical loads on the foundation remained constant while the moment
loads varied with wave incidence. Without a standard design methodology, the results showed a
linear relationship between the vertical load and the caisson diameter based on a function of the
moment load to the horizontal load ratio and the weight of the soil plug within the caisson cavity
(Figure 20b).
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8.0 Foundation Transport and Installation 8.1 Marine Operations
Marine operations for offshore wind farm development can include the cover yard lift, load out,
sea transportation, offshore lift, and installation operations. The costs of marine operations result
in much more expensive per turbine installation costs than onshore wind turbine installation.
Close integration of design and construction is needed due to the additional challenges of
operating at sea. To appreciate the size of the equipment needed and why the costs of marine
operations are high, the typical foundation sizes for turbines and construction aspects are shown
in Table 24. The construction constraints associated with different foundation types also affect
costs as summarized in Table 25.
Offshore wind turbine foundations are usually installed by floating crane vessels or mobile jack-
up units, the choice of which is dependent on water depth, crane capability, and vessel
availability (Figure 21). When using a crane vessel, it must be capable of lifting hook heights
greater than the height of the rotor-nacelle assembly of the turbine. Some of the lift capacities
along with other equipment specifications are summarized in Table 26. In shallow waters,
conventional mobile jack-up rigs are typical, whereas for deeper waters, the floating crane
vessels are usually deployed. Because the jack-up rigs establish their stability on the seafloor,
they are susceptible to the same issues and risks that foundations are designed against.
8.2 Project Duration Existing crane vessels have not been specifically designed for wind turbine installation, but large
wind farms may benefit through time and cost incentives from purpose-built units. As an
estimate, the total duration of installation for a multi-unit wind farm will take several months.
All marine operations are subject to weather constraints and down-time, so scheduling during
calm periods when wind and wave speeds are minimal is recommended. The actual construction
time for piled versus gravity base foundations either from jack-up units or floating vessels are
similar when considering weather-related down time. However, there is a significant cost
savings for a two phase installation (i.e. foundation unit installation followed by sub-structure,
tower, and rotor-nacelle assembly installation) versus a three phase installation (i.e. foundation,
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sub-structure and tower, and rotor-nacelle assembly installed separately). To reduce marine
operation costs in the future, there are three main objectives that should be addressed: 1)
improved dissemination of offshore construction procedural and technical knowledge among
designers and developers, 2) evaluation of resistance of existing offshore pile design techniques
for low mass, fatigue-dominated applications, and 3) optimization of offshore wind structure
installation cost-effectiveness using novel construction approaches (Watson 2000). More
information on transport and installation aspects can be found in DNV Rules for Planning and
Execution of Marine Operations (DNV-OS-J101 2004).
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9.0 Costs of Wind Energy 9.1 Renewable Energy Market Incentives
There are two main types of market incentives established to increase the development of
renewable energy production, namely fixed price systems and fixed quantity systems. Both
incentive types are ways of creating a protected market and offsetting the competitive
disadvantage that results when markets neglect environmental effects of conventional energy
production. Their main purpose is to provide incentives for technological improvements and cost
reductions.
In fixed price systems, operators are paid a fixed price for every unit of output, and the extra cost
is paid by taxpayers and electricity consumers. This system is highly effective in Denmark,
Spain, and Germany. In fixed quantity systems, the national government decides the target level
of renewable electricity to be achieved over a certain time period, while markets establish the
price. This system is currently used in the UK, Belgium, Sweden, and Italy via tradable green
certificate systems. It is still in its early stages, and long-term power purchase contracts are still
challenging. An advantage to this type of incentive is that it introduces competition between the
electricity producers (EWEA 2004).
The current status of the wind industry shows that the majority of incentives in effect are
financially-based, with governmental incentives outweighing the investment incentives. Tax
incentives, including tax and duty discounts, and production incentives, such as renewable
energy portfolios, production incentives, and fixed tariffs are less prevalent in the current global
economy (Shaw et. al. 2002).
9.2 General Cost Considerations
9.2.1 Introduction
Offshore wind farm development is more expensive compared to onshore wind farm
development due to foundation costs, access difficulties, water depths, weather delays, and
corrosion protection requirements. However, the factors affecting the cost of energy are similar,
as determined by the following eight factors: 1) total investment cost, 2) project preparation
86
costs, 3) operation and maintenance costs, 4) magnitude and availability of mean wind speed at
turbine hub height, 5) technical design lifetime, 6) amortization period, and 7) real interest rate.
Although these factors contribute to a higher investment cost, offshore wind farms can offer up
to 40% more energy potential than onshore wind farms due to higher wind velocities and less
turbulence (Irvine et. al. 2003).
There are two primary economies of scale in offshore wind turbine development: wind farm size
and wind turbine size. To be economical, wind farms generally have to be greater than 100 MW
of generating capacity using turbines of at least a 2 MW capacity. Based on a study of existing
European projects in year 2001, the cost of energy for offshore wind farm development ranged
from $0.053 to $0.112 per kWh, increasing with distance from shore and decreasing with
number of turbines per project. With further advances in technology, deep water wind turbines
could reach $0.051 per kWh, while shallow water wind turbines are projected to reach $0.041
per kWh within the next seven years (Musial and Butterfield 2004).
9.2.2 Cost of Energy Factors
9.2.2.1 Total Investment Cost
The total investment cost includes the production, transportation, and erection cost of the turbine,
which can range from $1200 to $2000 per kW, with production accounting for approximately
half of this cost (Manwell 2001). Up to 40% of this cost can be the turbine itself, with the tower
and foundation contributing from 28 to 34%, grid connection from 9 to 36%, and other capital
cost accounting for 6 to 17% of the total cost. Considering the turbine size, installations costs for
offshore foundations are 200 to 250% higher than for onshore foundations for medium (e.g. 0.5
to 0.8 MW) turbines in very shallow water (e.g. < 6-m). This reduces to 150 to 200% for
installations of larger turbines (e.g. 1 to 1.5 MW) in 6 to 15-m water depths. Considering the
foundation type, piled foundations versus gravity base foundations do not influence cost
significantly for up to 15-m water depth (Gaudiosi 1996).
9.2.2.2 Project Preparation Cost
The project preparation costs include permits, land, and infrastructure. Electricity generation
falls under this category, which can cost 30% higher offshore than onshore (Gaudiosi 1996).
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Grid connection cost is proportional to the number of turbines, however the capacity of the cable
does not affect cost (Krohn 2002). Typical costs for cabling can range from $500k to $1M per
mile (Manwell 2001).
9.2.2.3 Operation and Maintenance Cost
The operation and maintenance costs (O&M) include service, consumables, repair, insurance,
administration, and site leasing costs. Offshore O&M cost is approximately 100% higher than
onshore (Gaudiosi 1996), specifically ranging from $0.01 to $0.02 per kWh (Manwell 2001).
Generally, O&M cost is proportional to the number of turbines in a wind farm, distance form
shore, and occurrence of inclement weather, decreasing with more reliable turbine design.
9.2.2.4 Wind Speed and Availability
The mean wind speed at hub height and its availability is also important to energy cost. The
annual energy output can be estimated by following expression:
( )AVE 32.3= Eq. 9-1
where:
E = annual energy output (kWh/m2 swept rotor area)
V = mean annual wind speed (m/s)
A = swept rotor area (m2)
Generally, as the mean wind speed increases, the cost per kWh decreases exponentially to a
stable value. The availability is defined as the time percentage that the wind speed falls between
the turbine’s cut-in and cut-out wind speeds, which is typically higher than 98% of the time
(Beurskens and Jensen 2005).
9.2.2.5 Technical Design Lifetime
The technical design lifetime is typically 20 years. Parts of the rotor-nacelle assembly system
and other mechanical components usually need replacement every 1 to 5 years (Beurskens and
Jensen 2005). Other parts subjected to fatigue loads may need replacement halfway through the
design lifetime. Foundations are generally built to last 50 years. Therefore, foundations can
theoretically go through two cycles of turbines, which can lower generating costs by 25 to 33%,
falling within cost comparisons to onshore sites (Krohn 2002). Decommissioning costs also
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should be considered in the overall estimate, which vary depending on the type of foundation
used and the decommissioning policy.
9.2.2.6 Amortization Period
The amortization period depends on the economic model used, the type of investor, and their
financial strategy. It is often equal to the technical design lifetime (Beurskens and Jensen 2005).
An issue to consider when defining this period is the potential for technological advancement
during the project preparation phase.
9.2.2.7 Real Interest Rate
The real interest rate influences the cost of wind energy according to the following:
mAE
aIc tot += Eq. 9-2
where:
c = cost ($/kWh)
Itot = total investment per m2 swept rotor area
A = availability
E = annual energy output per m2 swept rotor area (kWh/m2)
m = operation and maintenance cost
a = annuity factor calculated according to the following equation:
( )( ) 11
1−+
+= n
n
niia Eq. 9-3
where:
i = real interest
n = amortization period
9.2.3 Cost Optimization
A study was conducted by the Riso National Laboratory in Denmark to determine the factors
affecting cost optimization of an offshore wind turbine farm (Fuglsang and Thomsen 1998). The
reference turbine was a 1.5 MW turbine with a rotor diameter of 60-m. Two separate wind farm
configurations were examined based on turbine spacing as a multiple of rotor diameter. The cost
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analysis was based on component weight in order to avoid subcontractor and market price
fluctuations.
The analysis showed that the offshore cost to onshore cost comparison was 42% higher primarily
due to increased foundation and grid connection costs, with each cost component split into fixed
and variable cost shares (e.g. transport and component weight, respectively). Table 27 shows the
relative cost distribution for an onshore wind farm compared to an equivalent offshore wind farm
based on the reference wind turbines.
The results of the study indicate that the factors influencing the cost optimization of offshore
wind energy compared to onshore cost are an increase in swept area due to rotor diameter
increase, increase in rated power, a reduced rotor speed, and a lowered hub height. All factors
combined resulted in an optimized cost of offshore wind energy of approximately 12% below the
onshore wind energy costs, regardless of the spacing of the wind turbines.
9.3 Costs for United States Offshore Wind Farm Development
9.3.1 Department of Energy Cost Estimates
A 1979 study was conducted for United States offshore wind turbine development that resulted
in a formula for fabricated cost based on loading and material parameters (Kilar and Stiller
1980). The formula was representative of piled foundations made of concrete (0 to 15-m water
depth) or steel (15 to 250-m water depth), and gravity base foundations made of concrete (60 to
150-m water depth) or steel (15 to 150-m water depth). The resulting formula is represented as
follows:
( ) ( ) 159.0+= ze
y FhxC Eq. 9-4
where:
C = Fabricated Cost (1979 $M)
h = Water depth (ft)
Fe = Effective free-end load (106 ft-lbs)
Mt = Total tipping moment (106 ft-lbs)
x,y,z = Depth and load coefficients based on foundation material
90
The resulting costs for typical environmental conditions off the Atlantic Coast based on 100 unit
plants of 7 to 10 MW output are shown in Table 28. The cost of energy ranged between $0.064
and $0.176/kWh, with the Pacific Northwest on the lower side of the scale and New England
near $0.102/kWh, compared with land-based estimates of $0.03 to $0.05/kWh (Rogers et. al.
2003). Although the costs would have to scaled to the current economy and technological
advances since 1979 would have to be considered, the table highlights the effects of water depth
and mean wind speed on cost.
9.3.2 Other Estimates
Other studies in the U.S. that were based on generic 5-MW turbines in a 500-MW wind farm
founded on monopiles 15-miles from shore in wind speeds greater than 8-m/s (Musial and
Butterfield 2004). Shallow water cost of energy estimates were based on design improvements
and changes for offshore conditions, and a production learning curve for cost reduction per
doubling of installed capacity. Table 29 shows a breakdown of the shallow water cost
projections.
The deep water estimate was based on 600-ft of water depth, with an initial marinization cost
premium of 11% higher than the land-based value. Differences in cost were due to the floating
structure and additional electric cabling. Baseline O&M costs were higher than for shallow water
($0.018/kWh) due to added platform hardware and distance to shore, all other assumptions and
design remaining the same. Figure 22 shows COE projections graph for a novel deep water
floating single turbine option (Musial and Butterfield 2004).
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10.0 Offshore Wind Development in New England 10.1 Introduction
The history of offshore wind energy potential in the US has focused on three primary areas: the
Pacific Northwest, Great Lakes, and New England. Due to the deeper water depths of these areas
compared to the developed wind farm areas of Europe, many initial concepts were based on
floating structures. The greater prospect of wind farm development in New England compared to
the other locations is due to two primary reasons: 1) there is a high wind resource in shallow
waters close to major electrical loads, resulting in shorter electrical transmission lines, and 2)
there are difficulties developing large onshore wind farms (Rogers et. al. 2003). Political factors
that have contributed to the interest include implementation of utility deregulation legislation that
mandated System Benefit Charges and Renewable Portfolio Standards (RPS), as well as the
advocacy of wind and renewable energy professionals (Manwell et. al. 2001). RPS and the
Systems Benefit Charge require development of up to 1100 MW of new renewable energy
capacity by year 2009 (Rogers et. al. 2003).
10.2 Environmental Conditions and Constraints
10.2.1 Wind Speed
Mesoscale modeling, developed by TrueWind Solutions, has provided estimates of wind
resources out to 50 nautical miles (nm) offshore (Musial and Butterfield 2004). It is being used
to validate previous observations at 12 monitoring stations, half of which are located on islands
or along the coast, the other half located between 12 and 170-nm off of the New England coast.
The average estimated wind speeds for the sites range from 7.0 to 8.4-m/s at 60-m height. The
Measure Correlate Predict method (MCP) is being used to predict wind data at other locations
based on reference sites in the general vicinity, implemented with Visual Basic computer code.
It is combined with the log law to predict wind data at heights greater than those at which data
was obtained (Manwell et. al. 2002). Results have indicated that there are immense areas of
Class 5, 6, and 7 winds at 5 to 50-nm from shore (Table 30) (Musial and Butterfield 2004).
10.2.2 Water Depth
The Massachusetts coast consists of shallow water depths (up to 18.3-m) that extend from 2 to 4-
km from shore. In the Cape Cod Bay, water depths less than 18.3-m extend out to 15-km from
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shore. In Nantucket Sound and Buzzard’s bay, most of the water depth is shallower than 18.3-m.
Based on these numbers, shallow water projects off coast of Massachusetts could produce up to
55,000 GWh, or 116% of the state’s energy needs (Rogers et. al. 2003).
10.2.3 Wave Loading
As of yet, there is no well-defined wave data for New England. Wave, tide, and current
conditions are less defined in the U.S than in the more shallow and sheltered European seas
(Musial and Butterfield 2004). Wave current speeds of 1.3-m/s (2.5-knots) have been measured
in Nantucket Sound, while the maximum measured waves heights at the mouth of Boston harbor
were 9.1-m (Rogers et. al. 2003).
10.2.4 Subsurface Conditions
The geology of Cape Cod and its surrounding areas results from glacial deposition, creating a
mix of seafloor soil types. In the shallow waters around Cape Cod, sand is most prevalent, while
there is a transition from a sand-clay mixture to only clay in the deepest waters. There are also
scattered boulders, rocks, and outcrops throughout the seafloor. Nantucket Sound contains
numerous shifting sand shoals, and significant slopes in these areas can create problematic
installation issues (Rogers et. al. 2003).
10.3 Other Design Considerations
10.3.1 Land Jurisdiction
The coastal waters in Massachusetts are divided into three categories of jurisdiction. The state
has control for up to 4.8-km (3-miles) from shore. The Massachusetts Ocean Sanctuaries
association has established that energy generation is prohibited within these limits, which may be
subject to future amendments. Currently, there is no standardized procedure for permitting in
Massachusetts, therefore siting a wind farm in waters further than 4.8-km from shore may
significantly reduce the permitting process. The federal government has jurisdiction from 4.8 to
320-km (200-mile) from shore, headed by the Department of the Interior and the U.S. Army
Corps of Engineers. Other entities of power include the U.S. Coast Guard, Federal Aviation
Association, and the Department of Agriculture. Beyond 320-km, waters are considered to be
under international jurisdiction (Rogers et. al. 2003).
93
10.3.2 Electrical Grid Connections
Significant grid connections exist along the coast of Massachusetts. However, future power
production could be limited by grid capacity on Cape Cod or from power lines crossing the Cape
Cod Canal. The primary option for transferring the power to shore is undersea cabling,
accomplished by washing the cables using high pressure water jets below the seafloor. This is
the most economical method compared to trenching or plowing cables into the seafloor, however
it is highly dependent on seafloor geology and soil conditions. As an example of grid connection
sizes, current wind farms in Europe use 20 to 40-kV connections with a 30 to 150-kV
transformer inside each wind farm, while links to the mainland have used 120 to 150-kV
connections (Krohn 2002).
10.3.3 Availability of Harbors for Construction and Maintenance
Massachusetts has significant dock, dry dock, and shipyard facilities, located in strategic towns
such as New Bedford and Quincy. These areas are necessary for support craft, barge mounted
cranes, cable-laying vessels, staging areas, and specialty-built vessels and floating systems.
10.3.4 Local Public Acceptance and Conflicting Areas
There are certain public considerations that must be addressed when siting an offshore wind
farm. A substantial consideration is the visual impact that the turbines will have on the horizon.
Models have shown that a 64-m-tall turbine is visible above the horizon at 36.6-km away. At
night, beacons may be required to alert airlines and ships. Current FAA regulations require
anything taller than 61-m to have beacons (Rogers et. al. 2003).
Another issue with siting an offshore wind farm in the New England area is the prevalence of
shipwrecks. There are many archeological sites around coast of Massachusetts, especially off
Nantucket Sound and east of Cape Cod coast.
10.4 Advancement of Deep Water Technology The expected progression in the U.S. towards deep water exploitation of wind farms will result
from shallow water monopiles or gravity base foundations that provide application-specific
experience towards the deeper water multiple-leg foundations and floating designs (Figure 23).
94
The economical step up from shallow fixed foundations to deep floating foundations is based on
the additional floating requirements and grid connection system. These added costs can be based
on oil and gas floating platforms with reductions made for the differences between oil and gas
applications and wind turbine applications including: 1) higher safety margins for personnel,
evacuation requirements, and spill prevention for oil and gas platforms, 2) the premise that
depths up to 600-ft for floating wind turbine platforms would be adequate as opposed to 1500-ft
and greater for oil and gas platforms, and 3) that wind platforms should be submergible to
minimize wave loading as opposed to oil and gas platforms which minimize wave loading by
constructing a above-water deck area (Musial and Butterfield 2004).
A concerted research and development effort from the U.S. is necessary to address the critical
issues in floating designs, including dynamic modeling of turbines and platforms, floating
platform optimization, low-cost mooring and anchor development, floating wind turbine COE
optimization strategies, deepwater erection and decommissioning, standards governing floating
turbines, and deepwater resource assessment. Figure 24 shows a summary of the necessary steps
in achieving this goal (Musial and Butterfield 2004).
95
11.0 Conclusions The global development of offshore wind energy is heading towards larger turbines, larger wind
farms, and further distances offshore into deeper waters. The unique loading aspects of offshore
wind turbine foundations require further development of the current state of practice in design of
these structures. Although the costs of offshore wind energy can be competitive with onshore
wind energy generation, refinement of the current foundation technology and offshore
construction procedures are necessary to decrease the relatively higher foundation manufacture,
transport, and installation costs.
The technology required to achieve this advancement will require continued efforts from
government bodies, private institutions, and the offshore energy industry. Although existing
technology will be adequate to develop a substantial infrastructure for shallow water offshore
wind energy generation, advances towards use of novel foundation concepts such as suction
caissons and floating wind turbine platforms will be necessary to secure larger wind turbines in
the ocean. The potential value in the investment of this technology can be seen by the current
projections of wind energy generating capacity in the near future, as countries foresee the
increased reliance on more renewable energy sources and less resource-depleting energy sources.
Areas of the United States that are potentially high-output offshore wind energy producers
include the shallow waters along the coasts of New England and the mid-Atlantic states, as well
as parts of the West Coast, where deeper waters, and hence, a demand for technological
investment, prevail. Areas of the Gulf of Mexico may also see increased interest due to the
prevalence of existing oil and gas platforms that are nearing depletion and would otherwise face
decommissioning and removal activities.
96
REFERENCES J. Offshore Tech., (2002). Securing Wind Turbines in the Water, Journal of Offshore Technology, Vol. 10, No. 3, pp.14-19. Beurskens, J., Jensen, P.H., Wind Energy, www.ewea.org/documents/20_Beurskens_Hjuler_ RD.pdf , Accessed May 2005, 30pp. Byrne, B.W., Houlsby, G.T., (2003). Foundations for Offshore Wind Turbines, Phil. Trans. Royal Society of London, Vol. 361, pp.2909-2930. Byrne, B.W., Houlsby, G.T., (2002). Investigating Novel Foundations for Offshore Windpower Generation, Proceedings of the 21st International Conference on Offshore Mechanics and Artic Engineering, Oslo, Norway, pp.567-576. Coduto, D.P., (2001). Foundation Design: Principles and Practices, 2nd Edition, Prentice Hall, New Jersey, 883pp. Det Norske Veritas DNV-OS-J101: Design of Offshore Wind Turbine Structures, June 2004. 138pp. European Wind Energy Association, (EWEA). (2004). The Current Status of the Wind Industry, 12pp. Feld, T., Waegter, J., (2002). Integrated Support Structure Design Analysis, Journal of Offshore Technology, Vol. 10, No. 3, pp.10-13. Frandsen, S., Thogersen, M.L., (1999). Integrated Fatigue Loading for Wind Turbines in Wind Farms by Combining Ambient Turbulence and Wakes, Wind Engineering, Vol. 23, No. 6, pp327-339. Fuglsang, P., Thomsen, K., (1998). Cost Optimization of Wind Turbines for Offshore Wind Farms, Riso National Laboratory, Roskilde, Denmark, 31pp. Gaudiosi, G., (1996). Offshore Wind Energy in the World Context, WREC, pp.899-903. Gravesen, H., (2004). #10105(12/13/02) and #10189(2/24/04): Foundation Design (Text Proposal for WG03), www.ecn.nl, Accessed March 2005. Gravesen, H., Bro, C., (2003). A Simple but Complete Design Procedure for Offshore Wind Turbine Foundations, #10148(10/30/03), www.ecn.nl, Accessed March 2005, 35pp. Henderson, A.R., Morgan, C., Barthelmie, R., Smith, B., Sorenson, H.C., Boesmans, B., (2002). Offshore Wind Energy – Review of the State-of-the-Art, Proceedings of the 12th International Offshore and Polar Engineering Conference, Kitakyushu, Japan, May 26-31, pp.494-498.
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Houlsby, G.T., Byrne, B.W., (2000). Suction Caisson Foundations for Offshore Wind Turbines and Anemometer Masts, Wind Engineering, Vol. 24, No. 4, pp.249-255. Houlsby, G.T., Byrne, B.W., Martin, C.M., (2001). Novel Foundations for Offshore Wind Farms, A Research Proposal to EPSRC (August 2001), 9pp. Irvine, J.H., Allan, P.G., Clarke, B.G., Peng, J.R., (2003). Improving the Lateral Stability of Monopile Foundations, Proceedings of the BGA International Conference on Foundations innovations, observations, design and practice, Thomas Telford, London, pp.371-380. Jenner, C., Finch, M., Finlayson, K., Harker, G., (2002). Optimizing Integrated Site Investigations for Offshore Wind Farm Projects, Proceedings of the International Conference on Offshore Site Investigations: Diversity and Sustainability, November 26-28, London, United Kingdom, pp.141-149. Kellezi, L., Hansen, P.B., (2003). Static and Dynamic Analysis of an Offshore Monopile Windmill Foundation, Proceedings of the BGA International Conference on Foundations innovations, observations, design and practice, Thomas Telford, London, pp.401-410. Kilar, L.A., Stiller, P.H., (1980). An Assessment of Offshore Wind Energy, Energy Technology: Proceedings of the Energy Technology Conference, Vol. 2, pp.1471-1483. Krohn, S. (2002). Offshore Wind Energy: Full Speed Ahead, www.windpower.org/en/ articles/offshore.htm. Musial, W. and Butterfield, S., (2004). Future for Offshore Wind Energy in the United States, Proceedings of the EnergyOcean Conference, June 28-29, 2004, Palm Beach, Florida, 13pp. Manwell, J.F., (2001). An Overview of the Technology and Economics of Offshore Wind Farms, Presentation for the Cape and Islands Offshore Wind Public Outreach Initiative, Meeting No. 3, November 21, 2001. Manwell, J.F., McGowan, J.G., Rogers, A.L., (2002). Wind Energy Explained: Theory, Design, and Application, John Wiley & Sons, Ltd., England. 577pp. Manwell, J.F., Rogers, A.L., McGowan, J.G., (2001). Status of Offshore Wind Energy in the United States, Proceedings of the Second International Workshop on Transmission Networks for Offshore Wind Farms, March 29-30, 2001, Royal Institute of Technology, Stockholm, Sweden, pp10-13. Manwell, J.F., Rogers, A.L., McGowan, J.G., Elkinton, C.N., (2004). Characterization of External Conditions for the Design of Offshore Wind Energy Systems for the United States, Proceedings of the 2004 ASME Wind Energy Symposium at the 42nd AIAA Aerospace Sciences Meeting and Exhibit, Reno, Nevada, January 5-8, pp.487-497.
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Manwell, J.F., Rogers, A.L., McGowan, J.G. and Bailey, B.H., (2002). An Offshore Wind Resource Assessment Study for New England, Renewable Energy, Vol. 27, pp.175-187. Norsok Standard G-0111, (2004). Marine Soil Investigations, Revision 2, October. 66pp. Petersen, P., Riber, H.J., Ronold, K., Det Norske Veritas (DNV), (2003). Rules for Design of Offshore Wind Turbine Structures, www.dnv.com/binaries/DNVrules_offshore_2003_tcm4-29389.pdf, Accessed March 2005, 8pp. Poulos, H.G., (1988). Marine Geotechnics, Unwin Hyman Ltd., London, 473pp. Powers, M.B., (2005). Platform-Based Wind Farm May Also Pump Up Oil Industry, Engineering News-Record, Vol. 254, No. 3, pp.31. Randolph, M.F., Kenkhuis, J., (2001). Offshore Foundation Systems 406: Site Investigation Planning – Course Notes, Center for Offshore Foundation Systems, University of Western Australia. Rogers, A.L., Manwell, J.F. and McGowan, J.G., (2003). A Year 2000 Summary of Offshore Wind Development in the United States, Energy Conversion and Management, Vol. 44, pp.215-229. Sahin, A.D., (2004). Progress and Recent Trends in Wind Energy, Progress in Energy and Combustion Science, 43pp. Schneider, J., (2004). Nearshore and Offshore Site Investigations and Foundations, (Presented at the University of Massachusetts, Amherst), October. Shaw, S., Cremers, M.J., Palmers, G., (2002). Enabling Offshore Wind Developments, European Wind Energy Association, 133pp. Offshore Site Investigation Committee (OSIC): Allan, P., Cook, M., Power, P., (2004). Guidance Notes on Geotechnical Investigations for Marine Pipelines, Rev. 03, Society for Underwater Technology, 47.pp. van der Tempel, J., Zaaijer, M.B., Subroto, H., (2002+). The Effects of Scour on the Design of Offshore Wind Turbines, Delft University of Technology, The Netherlands, 9pp. Vugts, J.H., Harland, L.A., (1997). Optimization of the Design of Support Structures for Offshore Wind Energy Converters Using Advanced Offshore Wind Engineering Technology, Wind Engineering, Vol. 21, No. 5, pp.319-337. Watson, G., (2000). Structure and Foundation Design of Offshore Wind Installations, Final Report from the Offshore Wind Energy Network Workshop, March, CLRC Rutherford Appleton Laboratory. 27pp.
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Zaaijer, M.B. (ed.), (2002). Design Methods for Offshore Wind Turbines at Exposed Sites (OWTES): Sensitivity Analysis for Foundations of Offshore Wind Turbines, Wind Energy Section, TUDelft. 49pp.
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LIST OF TABLES Table Page Table 1. Technologies Involved in Offshore Structure Design. (Poulos 1988)......................... 102 Table 2. Phases of Project Certification and Associated Activities. (DNV-OS-J101 2004) ..... 102 Table 3. Design Standards for the Offshore Industry. ............................................................... 103 Table 4. Worldwide Offshore Wind Energy Potential. (Gaudiosi 1996) .................................. 104 Table 5. Operational Offshore Wind Farm Projects. ................................................................. 104 Table 6. European Offshore Energy Potential by Country. (Sahin 2004) ................................. 104 Table 7. Environmental Parameters for Coastal United States.................................................. 105 Table 8. Offshore Resource Estimates for U.S. (Musial and Butterfield 2004) ........................ 105 Table 9. Planned Offshore Wind Farms. ................................................................................... 106 Table 10. Characteristic Load Selection Basis. (DNV-OS-J101 2004)..................................... 107 Table 11. Load Factors for ULS and ALS. (DNV-OS-J101 2004) ........................................... 107 Table 12. Application of Site Condition Data for Project Phases of an Offshore Wind Farm.
(Jenner et. al. 2002)............................................................................................................. 108 Table 13. Offshore Site Investigation Planning Process. (Randolph and Kenkhuis 2001) ...... 109 Table 14. Typical Offshore Site Investigation and Analysis Schedule. (Poulos 1988)............ 109 Table 15. Seismic Profiling Methods. (Randolph and Kenkhuis 2001) ................................... 110 Table 16. Accuracy Classes for Cone Penetration Testing. (Norsok 2004) ............................. 110 Table 17. Field Vane Blade Size. (Norsok 2004) ..................................................................... 110 Table 18. Suitability of Test Methods for Soil Parameters. (OSIC 2004)................................ 111 Table 19. Other Soil Parameters for Specific Applications. (OSIC 2004) ............................... 112 Table 20. Evaluation of Sample Quality. (Norsok 2004) ......................................................... 112 Table 21. Model Parameters for Gravity-based Foundations. (Zaaijer 2002) ........................... 113 Table 22. Estimations of Effective Fixity Length. (Zaaijer 2002)............................................. 113 Table 23. Important Natural Frequency Sensitivities for Various Structural Configurations.
(Zaaijer 2002)...................................................................................................................... 114 Table 24. Typical Sizes and Construction Aspects. (Watson 2000)......................................... 115 Table 25. Foundation Construction Constraints. (Watson 2000) .............................................. 115 Table 26. Typical Installation Vessel Specifications. (Watson 2000) ....................................... 115 Table 27. Relative Cost Distribution for Onshore vs. Offshore Wind Farm Comparison.
(Fuglsang and Thomsen 1998)............................................................................................ 116 Table 28. Costs of Offshore Wind Farms for Atlantic Coast Environmental Conditions. (Kilar
and Stellar 1980) ................................................................................................................. 116 Table 29. Shallow Water COE Projections. (Musial and Butterfield 2004).............................. 117 Table 30. Wind Speeds Based on Class..................................................................................... 117
101
LIST OF FIGURES Figure Page Figure 1. a) Standard Monopile Structure, b) Supported Monopile Structure. (DNV-OS-J101
2004) ................................................................................................................................... 118 Figure 2. a) Tripod Structure, b) Gravity Pile Structure. (DNV-OS-J101 2004) ..................... 118 Figure 3. Lattice Tower. (DNV-OS-J101 2004)........................................................................ 119 Figure 4. Gravity Base Structure. (DNV-OS-J101 2004).......................................................... 119 Figure 5. Typical Relationship Between Ballast Component and Foundation Diameter. (Byrne
and Houlsby 2002).............................................................................................................. 120 Figure 6. a) Suction Bucket Structure (DNV-OS-J101 2004), and b) Installation Principle.
(Byrne and Houlsby 2003).................................................................................................. 120 Figure 7. Tension-Leg Platform. (DNV-OS-J101 2004) ........................................................... 121 Figure 8. Low-roll Floater. (DNV-OS-J101 2004).................................................................... 121 Figure 9. Ocean Sediment Distribution Throughout the Northern Hemisphere........................ 122 Figure 10. Scour Model. (Zaaijer 2002) .................................................................................... 123 Figure 11. Flowchart for Determining Scour Potential. (van der Tempel 2002)....................... 123 Figure 12. Spring Model of Pile-Soil Interaction. (Zaaijer 2002) ............................................. 124 Figure 13. Gravity-based Foundation Model. (Zaaijer 2002).................................................... 124 Figure 14. Example Yield Surface for Footings on Sand. (Byrne and Houlsby 2002) ............. 125 Figure 15. Comparison of a) Laboratory Test Data with b) Continuous Hyperplasticity Theory.
............................................................................................................................................. 125 Figure 16. Predicted 1st and 2nd Natural Frequency for Several Foundation Models. (Zaaijer
2002) ................................................................................................................................... 126 Figure 17. Loading Comparison of Offshore Platform to Offshore Turbine. (Schneider 2004)
............................................................................................................................................. 126 Figure 18. Multi-footing Suction Caisson Geometry. (Byrne and Houlsby 2003).................... 127 Figure 19. Foundation Size and Mass as a Function of Structural Configuration. (Byrne and
Houlsby 2002)..................................................................................................................... 127 Figure 20. a) Design of Quadruped Suction Caisson, and b) Design of Monopod Suction Caisson
Foundation. (Byrne & Houlsby 2003) ................................................................................ 128 Figure 21. Typical Foundation Installation Methods. (Watson 2000)....................................... 128 Figure 22. Deep Water COE Projections. (Musial and Butterfield 2004) ................................ 129 Figure 23. Expected Progression of Foundation Support Structures. (Musial and Butterfield
2004) ................................................................................................................................... 129 Figure 24. Deepwater Research and Development Strategy. (Musial and Butterfield 2004)... 130
102
Table 1. Technologies Involved in Offshore Structure Design. (Poulos 1988)
Table 2. Phases of Project Certification and Associated Activities. (DNV-OS-J101 2004)
103
Table 3. Design Standards for the Offshore Industry.
TITLE
Recommended Practice for Planning, Designing and Constructing Fixed Offshore Platforms – Working Stress Design, 21st Edition 2000Recommended Practice for Planning, Designing and Constructing Fixed Offshore Platforms – Load and Resistance Factor Design, 1993, (suppl. 1997)Recommended Practice for Planning, Designing and Constructing Structures and Pipelines for Arctic conditions, 2nd Edition 1995Recommended Practice for Design and Installation of Electrical Systems for Offshore
Design of Offshore Wind Turbine Structures, June 2004Design of Offshore Steel Structures, General (LRFD Method), April 2004Structural Design of Offshore Units (WSD Method), April 2002Fabrication and Testing of Offshore Structures, July 2004Fatigue Strength Analysis of Offshore Steel Structures, October 2001Cathodic Protection Design, January 2005Design Against Accidental Loads, November 2004Marine Operations during Removal of Offshore Structures, April 2004Foundations, February 1992Environmental Conditions and Environmental Loads, March 2000Structural Reliability Analysis of Marine Structures, July 1992Rules for classification of fixed offshore installations, 1998
General Principles on Reliability for Structures, 1998Petroleum and natural gas industries -- Offshore structures -- Part 1: General requirements, 1995-12, 1st edition. To be replaced , ISO TC 67. (ISO 19900)
Petroleum and Natural Gas Industries – Offshore Structures – Part 2: Fixed steel structures, 1995
Petroleum and Natural Gas Industries – Specific Requirements for Offshore Structures – Part 1: Metocean design and operating conditions, 2003Piled Foundations: Fixed Steel StructuresGravity Foudnations: Fixed Concrete Structures Offshore Structures: Floating systems Basis for Design of Structures – Seismic actions on structures, 2001
Rules for Classification and Construction, III Offshore Technology, 2 Offshore Installations, Edition 1999 Rules for Classification and Construction, III Offshore Technology, 2 Offshore Installations, Guidelines for the Construction/Certification of Floating Production, Storage and Off-Loading Units, Edition 1999
Mobile and fixed offshore units - Electrical installations - Part 3: Equipment, (1999-02) Mobile and fixed offshore units - Electrical installations - Part 6: Installation,(1999-02)
Cathodic protection for fixed steel offshore structures, 2000Weldable structural steels for fixed steel offshore structures, 1994
Structural Design, Rev. 3, Aug. 2000
Offshore installations: guidance on design, construction and certification (fourth edition) HMSO 1990 ISBN 011 4129614
Department of Energy (DOE)
-
American Petroleum Industry (API)
Det Norske Veritas (DNV)
International Organization for Standardizations (ISO)
Germanischer Lloyd (GL)
International Electrotechnical Commission (IEC)
German Institute for Standardization (DIN EN)
Norwegian Technology Center (NTC)
61892-6
1249510225
NORSOK N-001
3010
61892-3
19901
199021990319904
-
2394
13819-1
13819-2
RP-H102CN-30.4CN-30.5CN-30.6
OS-401RP-C203RP-B401RP-C204
-
OS-J101OS-C101OS-C201
STD. No.
RP-2A (WSD)
RP-2A (LRFD)
RP-2N
104
Table 4. Worldwide Offshore Wind Energy Potential. (Gaudiosi 1996)
Table 5. Operational Offshore Wind Farm Projects.
Table 6. European Offshore Energy Potential by Country. (Sahin 2004)
Country/RegionUsable
Offshore Area (sq km)
Wind Speed (m/s)
Power (GW)
Energy (TWh/yr)
US 9000 5.6-6.4 54 102China 21500 5.6-6.4 129 254European Union 162000 5.6-8.8 NA 1853Denmark 21000 5.6-7.0 NA 260The Netherlands 5600 5.6-6.4 NA 75UK 48000 5.6-8.8 NA 626Tunisia 1516 5.6-6.0 9 16Egyptian Mediterranean 2278 5.6-6.0 13 24Mediterranean Sea 8680 NA 52 100Black Sea 1200 NA 7 14
Country Name of Park Sea Location Year Commisioned
Installed Capacity
(MW)
No. Turbines Type of Turbine
Rotor Diameter
(m)
Production (GWh/yr)
Distance from Coast
(km)
Water Depth (m) Foundation Details
Denmark Middelgrunden NA 2001 40 20 Bonus 2.0 MW NA 81 1.7-3.5 3-5 Gravity FoundationDenmark Vindeby Baltic Sea 1991 4.95 11 Bonus 450 kW 35 11.73 1.5-3.0 3-5 Box Caissons on Sandy SoilDenmark Tuno Knob Kattegat Sea 1995 5 10 Vestas 500 kW 39 12.7 6 3-5 Box Caissons on Sandy SoilDenmark Horns Rev North Sea 2002 160 80 Vestas V80 2.0 MW NA NA 14-20 NA NADenmark Frederikshavn NA 2002 NA 4 NA NA NA NA NA NASweden Gotland-Bockstigen Baltic Sea 1997 2.75 5 Wind World 550 kW NA NA 4.5 5.5-6.5 Drilled and Grouted MonopilesSweden Utgrunden Baltic Sea 2001 10.5 7 Tacke TW1.5s NA NA 12.5 7.2-9.8 MonopilesSweden Yttre Stengrund NA 2001 10 5 NEG Micon 2.0 MW NA NA 5 6-10 MonopilesThe Netherlands Medemblik (Lely) Ijsselmeer Sea 1994 2 4 NedWind 500 kW 41 3.5 2 5-10 Steel Tower Driven in Sandy SoilThe Netherlands Dronten Ijsselmeer Sea 1996 16.8 28 Nordtank 600 kW NA NA 0.2 5 Monopiles in Fresh WaterUK Blyth Offshore North Sea 2000 3.8 2 Vestas V66 2.0 MW 43 NA 0.8 6-11 Monopiles in bedrockSources: Shaw et. Al. (2002), Beurskens and Jensen (2000), Gaudiosi (1996), Sahin (2004), Henderson et al. (2002), Manwell et al (2002)
CountryOffshore Potential (TWh/yr)
UK 986Denmark 550France 477Germany 237Ireland 183Italy 154Spain 140The Netherlands 136Turkey 130Greece 92Portugal 49Belgium 24Total 3158
105
Table 7. Environmental Parameters for Coastal United States.
Table 8. Offshore Resource Estimates for U.S. (Musial and Butterfield 2004)
Area Mean Wind Speed (m/s)
Max. Sustained Wind Speed*
(m/s)
Max. Wave Height*
(m)
Max. Surface Current (m/s)
Northeast Coast 4-8 45-65 30-50 0.2-1.1Southeast Coast 4-8 55-85 30-50 0.25-1.8Gulf Coast 4-7 45-70 30-40 0.2-0.9Southwest Coast 3.5-5.5 35-45 20-30 0.15-1.0West Coast 4-8 35-50 30-35 0.2-0.7S. Alaskan Coast 5-9.5 50-60 35-40 0.2-0.8Aleutian Islands 6-9 50-60 30-40 0.15-0.5Hawaiian Islands 4-6 40-50 20-35 0.2-0.7Puerto Rico and Virgin Islands 4-5 NA NA 0.25-0.4Notes: *Based on 100-year storm recurrenceSource: Kilar and Stiller 1980
106
Table 9. Planned Offshore Wind Farms.
Country Name of Park Sea Location Year Commisioned Installed Capacity (MW) No. Turbines Type of Turbine Production
(GWh/yr)
Distance from Coast
(km)Belgium Thorton Bank North Sea End 2005 NA 60-120 2.5-3.6 MW 1000 NABelgium Vlakte van de Raan North Sea 2003 100 50 Vestas V80 2.0 MW NA 12Belgium Wenduine NA 2003/2004 100 NA NA NA NABelgium Raan NA 2003/2004 100Denmark Roedsand Baltic Sea 2003 165.6 72 Bonus 2.3 MW 480 9-10Denmark Omo Stalgrunde NA 2004 150 96 NA NA 10Denmark Samso NA 2003 23 10 Bonus 2.3 MW NA NADenmark Laeso NA 2003 150 78 NA NA 40Denmark Adlergrund NA 2004 480Denmark Beltsee NA 2005 249France Nord-pas-de-Calais NA NA 775 NA NA 2400 5-8France Manche, Basse-Normandie NA NA 3500 NA NA 10800 5-10France Bretagne NA NA 2050 NA NA 6300 3-10France Languedoc-Roussilon NA NA 2800 NA NA 10600 3.5-10Germany Pommersche Bucht Baltic Sea NA 1000 200 5 MW NA 42Germany Arkona Becken Sudost Baltic Sea NA 945 189 4-5 MW NA NAGermany Adlergrund Baltic Sea NA 790 158 3-5 MW NA 40Germany Kriegers Flak Baltic Sea NA 315 75 3-5 MW NA 35Germany Pilot Mecklenberg-Vorpommern Baltic Sea NA 40 21 Nordex, Neptun, Brand Elektro NA 15Germany Beltsee Baltic Sea NA 415 83 3-5 MW NA 30Germany Sky 2000 Baltic Sea NA 100 50 2 MW (1/3 Vestas, rest open) NA 19Germany Dan Tysk North Sea NA 1500 300 5 MW NA 60Germany Weisse Bank North Sea NA 600 120-170 3.5-5 MW NA 60Germany Butendiek North Sea NA 240 80 3 MW NA 30Germany Offshore Helgoland North Sea NA 200 100 2 MW (Vestas) NA 13Germany Schleswig Holstein Nordsee North Sea NA 800-1000 200 4-5 MW NA 15Germany Amrumbank West North Sea NA up to 288 72 3-4 MW NA 35Germany Amrumbank/Nordsee-Ost AWZ North Sea NA 1250 250 5 MW (96x Repower NOK 5) NA 17Germany Meerwind North Sea NA 819 234 3.5 MW NA NAGermany Nordergrunde North Sea NA >200 76 2.5-5 MW NA 12-15Germany Wilhemshaven North Sea NA 4.5 1 4.5 MW (Enercon E-112) NA <10Germany Dollart North Sea NA 9 5 1.8 MW (Enercon) NA NAGermany Juist North Sea NA 1400 280 5 MW NA NAGermany Borkum III North Sea NA 60 (later 1000) 12 (later up to 285) 3.5-5 MW NA 40Germany Borkum Riffgrund North Sea NA 840 180 3.5 MW NA 34Germany Borkum IV North Sea NA 400 90-160 2.5-4.5 MW NA NAGermany Riffgat North Sea NA 135 30 4.5 MW (Enercon) NA 15Germany Borkum Riffgrund West North Sea NA 1800 458 2.5-5 MW NA 45Germany Sandbank 24 (Europe Pipe West) North Sea NA 2600 (1st phase 360) 120 3 MW NA 120Ireland Blackwater Bank 1 NA NA 260 NA NA NA NAIreland Blackwater bank 2 NA NA NA NA NA NA NAIreland Arklow Bank NA 2003-2006 (phased) 520 200 NA NA 7-12Ireland Codling and Greater Codling Bank NA NA NA NA NA NA NAIreland Bray Bank and Kish Bank NA 2003 up to 300 MW NA NA NA NAIreland Dundalk Bay NA NA NA NA NA NA NASweden Nogersund NA 1991 0.22 NA NA NA 0.25Sweden Rone Gotland NA NA 35 35 Nordic 1 MW NA NASweden Klasarden NA 2003 42 21 NEG Micon 2.0 MW NA NASweden Blekinge Oland Southern Skane NA NA 30 30 1 MW NA NASweden Blekinge Oland NA 1998 294 98 Nordic 3 MW 833 5Sweden Ystad Skane NA NA 10 NA NA NA NASweden Lillgrund Bank NA 2002 86.4 48 Enercon E66/18.70 NA NASweden Barsebank NA NA 750 NA NA NA NAThe Netherlands Q7 North Sea 2003 120 60 Vestas V80 2.0 MW 350 23The Netherlands Egmond aan Zee North Sea 2004 100 36 NEG Micon 2.75 MW NA NAUK Kentish Flats NA NA NA 30 NA NA 8UK Gunfleet Sands NA NA 29.8 NA NA NA 7UK Scroby Sands NA NA 50 25 Vestas V80 2.0 MW 98 2.3UK Cromer NA NA NA 30 NA NA 6.5UK Lynn NA NA NA 60 NA NA 5.2UK Inner Dowsling NA NA NA 60 NA NA 5.2UK Teesside NA NA NA 30 NA NA 1.5UK Solway Firth NA NA NA 60 NA NA 9UK Barrow NA NA NA 30 NA NA 10UK Shell Flat NA NA NA 90 NA NA 7UK Southport NA NA NA 30 NA NA 10UK Burbo NA NA NA 30 NA NA 5.2UK North Hoyle NA NA NA 60 NA NA 6UK Rhyl Flats NA NA NA 60 NA NA 8UK Scarweather Sands NA NA NA 30 NA NA 9.5UK NA NA 1987 20 50 400 kW 40 1-10Poland Bialogora NA 2004 61 NA NA NA NASpain Cabo de Trafalgar NA 2005 200 100 2 MW NA NACanada Nai Kun NA 2004/2008 700 NA NA NA NAUSA Nantucket Sound, MA Nantucket Sound NA 425 NA NA NA NASources: Shaw et. Al. (2002), Beurskens and Jensen (2000), Gaudiosi (1996), Sahin (2004), Henderson et al. (2002), Manwell et al (2002)
107
Table 10. Characteristic Load Selection Basis. (DNV-OS-J101 2004)
Table 11. Load Factors for ULS and ALS. (DNV-OS-J101 2004)
108
Table 12. Application of Site Condition Data for Project Phases of an Offshore Wind Farm. (Jenner et. al. 2002)
109
Table 13. Offshore Site Investigation Planning Process. (Randolph and Kenkhuis 2001)
Table 14. Typical Offshore Site Investigation and Analysis Schedule. (Poulos 1988)
Phase ActivityPre-planning Period 4-6 Weeks 3-5 Weeks 4-6 Weeks 6-8 Weeks
Work scope/budget definition XBudget approval X
Pre-qualification Period XPre-qualify contractors XTenders list approval X
Tender PeriodInvitation to tender XAssess tenders XApproval to award XAward tender X
Construction PeriodContractor mobilization XWork initiation X
3 - 6 Month Time Period
110
Table 15. Seismic Profiling Methods. (Randolph and Kenkhuis 2001)
Table 16. Accuracy Classes for Cone Penetration Testing. (Norsok 2004)
Table 17. Field Vane Blade Size. (Norsok 2004)
111
Table 18. Suitability of Test Methods for Soil Parameters. (OSIC 2004)
112
Table 19. Other Soil Parameters for Specific Applications. (OSIC 2004)
Table 20. Evaluation of Sample Quality. (Norsok 2004)
113
Table 21. Model Parameters for Gravity-based Foundations. (Zaaijer 2002)
Table 22. Estimations of Effective Fixity Length. (Zaaijer 2002)
114
Table 23. Important Natural Frequency Sensitivities for Various Structural Configurations. (Zaaijer 2002)
Structure Parameter Uncertainty Location LifetimeTubular Tower on Monopile % % %
Soil 4 4 6Foundation 0.06 - -Environment - 0.1 0.1Structure 0.02 4 4
Tubular Tower on Gravity Base Foundation % % %Soil 19 35 4Foundation 0.01 - -Environment - 0.03 0.03Structure 0.2 4 4
Tripod with Piles % % %Soil 0.9 0.9 0.7Foundation <0.01 - -Environment - 0.01 0.01Structure 0.1 3.2 3.2
Lattice Tower with Piles % % %Soil 3 3 3Foundation 0.01 - -Environment - 0.04 0.04Structure 0.04 4 4
Lattice Tower on Gravity Base Foundation % % %Soil 23 38 7.8Foundation 0.01 - -Environment - 0.02 0.02Structure 0.04 4 4
115
Table 24. Typical Sizes and Construction Aspects. (Watson 2000)
Table 25. Foundation Construction Constraints. (Watson 2000)
Table 26. Typical Installation Vessel Specifications. (Watson 2000)
116
Table 27. Relative Cost Distribution for Onshore vs. Offshore Wind Farm Comparison. (Fuglsang and Thomsen 1998)
Table 28. Costs of Offshore Wind Farms for Atlantic Coast Environmental Conditions. (Kilar and Stellar 1980)
Water DepthMean Wind
Speed @ 10-m a.s.l. (m/s)
5 7.5 10 5 7.5 10
Plant Cost (1979 $M) 7.3 5.7 5.7 7.5 7.7 5.7
Shallow Water (0 to 15-m) Deep Water (60 to 500-m)
117
Table 29. Shallow Water COE Projections. (Musial and Butterfield 2004)
Table 30. Wind Speeds Based on Class.
118
a) b)
Figure 1. a) Standard Monopile Structure, b) Supported Monopile Structure. (DNV-OS-J101 2004)
a) b)
Figure 2. a) Tripod Structure, b) Gravity Pile Structure. (DNV-OS-J101 2004)
119
Figure 3. Lattice Tower. (DNV-OS-J101 2004)
Figure 4. Gravity Base Structure. (DNV-OS-J101 2004)
120
Figure 5. Typical Relationship Between Ballast Component and Foundation Diameter. (Byrne and Houlsby 2002)
a) b)
Figure 6. a) Suction Bucket Structure (DNV-OS-J101 2004), and b) Installation Principle. (Byrne and Houlsby 2003)
121
Figure 7. Tension-Leg Platform. (DNV-OS-J101 2004)
Figure 8. Low-roll Floater. (DNV-OS-J101 2004)
122
Figure 9. Ocean Sediment Distribution Throughout the Northern Hemisphere.
(Poulos 1988)
USA
123
Figure 10. Scour Model. (Zaaijer 2002)
Figure 11. Flowchart for Determining Scour Potential. (van der Tempel 2002)
124
Figure 12. Spring Model of Pile-Soil Interaction. (Zaaijer 2002)
Figure 13. Gravity-based Foundation Model. (Zaaijer 2002)
125
Figure 14. Example Yield Surface for Footings on Sand. (Byrne and Houlsby 2002)
a) b)
Figure 15. Comparison of a) Laboratory Test Data with b) Continuous Hyperplasticity Theory.
(Byrne & Houlsby 2003)
126
Figure 16. Predicted 1st and 2nd Natural Frequency for Several Foundation Models.
(Zaaijer 2002)
Figure 17. Loading Comparison of Offshore Platform to Offshore Turbine. (Schneider 2004)
127
Figure 18. Multi-footing Suction Caisson Geometry. (Byrne and Houlsby 2003)
Figure 19. Foundation Size and Mass as a Function of Structural Configuration. (Byrne
and Houlsby 2002)
128
a) b)
Figure 20. a) Design of Quadruped Suction Caisson, and b) Design of Monopod Suction Caisson Foundation. (Byrne & Houlsby 2003)
Figure 21. Typical Foundation Installation Methods. (Watson 2000)
129
Figure 22. Deep Water COE Projections. (Musial and Butterfield 2004)
Figure 23. Expected Progression of Foundation Support Structures. (Musial and Butterfield 2004)
Deep Water Wind Turbine Development
Current Technology
Onshore Wind
Turbine
Shallow Water
0 M – 30 M
Transitional Depths
30 M – 50 M
Deep Water
50 M – 200 M
130
Figure 24. Deepwater Research and Development Strategy. (Musial and Butterfield 2004)