earthquake resistant design codes in japan 2000

JAPAN SOCIETY OF CIVIL ENGINEERS

EARTHQUAKE RESISTANT DESIGNCODES IN JAPAN

January,2000

JAPAN SOCIETY OF CIVIL ENGINEERS

EARTHQUAKE RESISTANT DESIGN CODES

OF CIVIL ENGINEERING STRUCTURES IN JAPAN

1. KEY CONCEPTS FOR EARTHQUAKE RESISTANT DESIGN OF CIVIL

ENGINEERING STRUCTURES AFTER THE 1995 HYOGOKEN-NANBU

EARTHQUAKE

2. 1996 SEISMIC DESIGN SPECIFICATIONS OF HIGHWAY BRIDGES

JAPAN ROAD ASSOCIATION

3. SEISMIC DESIGN FOR RAILWAY STRUCTURES

RAILWAY TECHNICAL RESEARCH INSTITUTE, JAPAN

4. EARTHQUAKE RESISTANT DESIGN OF PORT FACILITIES

BUREAU OF THE PORTS AND HARBORS, IvuNISTRY OF TRANSPORT

5. BASIC PRINCIPLES OF SEISMIC DESIGN AND CONSTRUCTION FOR WATER

SUPPLY FACILITIES

JAPAN WATER WORKS ASSOCIATION

6. RECOMMENDED PRACTICES FOR EARTHQUAKE RESISTANT DESIGN OF

GAS PIPELINES

JAPAN GAS ASSOCIATION

THE JAPAN SOCIETY OF CIVIL ENGINEERS

THE PUBLICATION COMMITTEE

OF

EARTHQUAKE RESISTANT DESIGN CODES

OF CIVIL ENGINEERING STRUCTURES IN JAPAN

Chairman:

Members:

Masanori Hamada (r#iseda Unievrsity)

Key Concepts for Earthquake Resistant Design

Shigeki Unjo (Public Works Research Institute, Ministry ofConstruction)

Highway Bridges

Akihiko Nishimura (Railway Technical Research Institute, Japan)

Railway Structures

Tatsuo Uwabe (Port and Harbor Research Institute, Ministry otTrensport)

Port Facilities

Seiji Une (Japan Water Works Association)

Water Supply Facilities

Hiroyuki Yamakawa (Japan Ges Associetion)

Gas Pipelines

1. KEY CONCEPTS FOR EARTHQUAKE RESISTANT DESIGN OF CIVIL

ENGINEERING STRUCTURES AFTER THE 1995 HYOGOKEN-NANBU

EARTHQUAKE

1.1 Lessons from The 1995 Hyogoken-nanbu (Kobe) Earthquake 1- 1

1.2 Key Concepts for Earthquake Resistant Design 1- 4

1.3 Technical Subjects for Revision of Earthquake Design Code 1- 6

1.4 Diagnosis and Reinforcement of Existing Structures 1- 7

1.5 Future Innovations of Design Codes and Research Subjects 1- 8

EARTHQUAKE RESISTANT DESIGN CODES IN JAPAN January, 2000

1. KEY CONCEPTS FOR EARTHQUAKE RESISTANT DESIGN OF CIVIL ENGINEERING

STRUCTURES AFTERTHE 1995 HYOGOKEN-NANBU EARTHQUAKE

1.1 Lessons from The 1995 Hyogoken-nanbu

(Kobe) Earthquake

At 5:46AM of January 17, 1995, a highly

urbanized area of western Japan was jolted by an

earthquake with a magnitude of M=7.2. This

earthquake affected an extensive area containing

major cities, Kobe and Osaka and their

surrounding satellite cities which constitute the

industrial, commercial and cultural center of

western Japan.

The areas most heavily damaged by this

earthquake extends in a belt-shaped zone along

the causative fault system with a length of 40km,

particularly the zones identified as JMA intensity

scale VII (equivalent to MMI=X). They extend

over the entire east-west length of the most

densely populated part of Hanshin (meaning

Osaka-Kobe) metropolitan region.. Three million

people in this region were seriously affected. A

free-field ground acceleration (pGA) exceeded

800cmfs2 in Kobe city and its response spectrum

was over 2000cmfs 2 at a damping coefficient of

0.05.

Table.Ll shows loss of human lives, and a

Table 1.1 A Summary of Damage Caused by the 1995 Kobe Earthquake (1995 Kobe Earthquake)

Human* Death: 6306 Missing: 2 Injured: 41,527

Housing and Buildings

Bridges **

Totally collapsed houses:Half and partially collapsed houses:Buildings:

Road (Hanshin Expressway): 67

100,300214,000

3,700

Railway: 32

Embankment and Landslides Embankment: 427 Landslides: 367

Water Customers without service: 1.2 million Restoration time: 40 days

Gas Customers without service: 857,000 Restoration time: 85 days

Electricity Customers without service:Outage of electric power:Restoration time:

2.6 million2836Mw7 days

Telecommunication Customers affected by Switchboard Malfunction: 235,000Damaged Cable Line: 19,300

Economic Impact Private properties:Transportation facilities:Lifelines:Others:Grand total:

¥6.3 trillion¥2.2 trillion¥0.6 trillion¥0.5 trillion¥9.6 trillion

***

Toll by Fire Defense Agency May 21, 1995

Collapsed and Extensively Damaged

1-1


summary of structural and functional disaster by

the Kobe earthquake to houses and buildings,

bridge, lifeline facilities and so on.

The first point to note about damage to civil

engineering structures is that elevated highway

bridge piers were completely destroyed. Although

there had been RC bridge piers damaged by

earthquakes in the past, this was the first

experience of total collapse in Japan. Most of the

seriously damaged piers were designed in

accordance with pre-1980 earthquake resistant

design codes. The piers of concrete structures

having low ductility and low ultimate strength,

were shear-fractured, resulting in such major

failures. Damage to RC piers designed in

conformance with the current earthquake resistant

design codes after 1980 was not so severe as to

result in bridge collapses.

Figure 1.2 Collapse of Bridge Piers of A

Elevated Highway Bridge (1995 Kobe

Earthquake)

Another point to note is the damage to steel

bridge piers. Many steel bridge piers buckled.

Most steel structures were designed by a method

where stresses in steel structural members fell

within an elastic region. The characteristics of

plastic deformation of steel structures had not

1-2

been· incorporated into design codes. This is one

of the technical subjects that the earthquake

showed needs to be promptly studied and

implemented.

Figure 1.3 Buckling of A Steel Pier of A

Bridge (1995 Kobe Earthquake)

Damage to large underground structures,

such as subway structures has also become a focus

ofattention. The severest dainage was caused at a

subway station in the downtown of Kobe city,

which is of box-type RC structure, where

reinforced concrete columns were shear- fractured

and an upper floor deck slab collapsed along with

the overburden soil. Severe damage to other

underground subway stations was also reported.

Besides subway tunnels, which were constructed

by the cut-and-fill method, many mountain

tunnels of railway and highway were also

damaged due to large ground motion in the near

field of the earthquake fault.

Another. typical characteristic of damage to

civil infrastructures caused by the Kobe

earthquake is collapses and large displacements of

quay walls. Numerous collapses of revetments and

quay walls had been reported in past earthquakes,

but most of them had not been designed to

withstand soil liquefaction and had been decaying.


This was the first time when recently constructed

quay walls were largely displaced by several

meters or collapsed. All the damaged quay walls

had been constructed using concrete caissons. The

result of the investigation into cause of the

damage to quay walls said that soft clay of the sea

bed largely amplified the earthquake motion and

the foundation ground of the caissons, which had

been constructed by replacing the original sea bed

of soft clay with liquefiable gravel sand,

weathered granite, also liquefied besides the filled

ground behind the quay walls.

Figure 1.4 Large Movement of Concrete

Caisson Quay walls (1995 Kobe Earthquake)

However, it should be noted that all the

so-called earthquake resistant quay walls mostly

survived. The construction of earthquake resistant

quay walls has been promoted nationwide, mainly

in major ports and harbors, through the lessons

leamed from the damage to quay walls in Akita

Harbor during the 1983 Nihonkai-Chubu

earthquake. The earthquake resistant quay walls,

which were designed by adopting a higher seismic

load than that for conventional quay walls, were

constructed to withstand liquefaction.

Damage to RC elevated railway bridge piers

1-3

of the Shinkansen (bullet trains) shocked not only

civil engineers, but also the general public. RC

bridge piers were shear-fractured and collapsed,

and girders fell. Fortunately, because the

earthquake struck 14 minutes before service hours,

no human life was lost. A serious issue has

surfaced of how to assure the safety of high speed

Figure 1.5 Soil Liquefaction of An Artificial

Island in Kobe (1995 Kobe Earthquake)

trains, including Shinkansen, against earthquakes

caused by inland faults directly below them.

Soil liquefaction was extensively caused in

the artificial islands and alluvial low lands in

Kobe and its neighboring areas, which resulted in

a significant damage to buried pipes and

structures of lifeline systems, and many port

facilities. Most of the artificial islands in Kobe

area was reclaimed from the sea by weathered

granite which contained large cobbles and fine

contents. This revealed a need of revision of the

method to evaluate the liquefaction potential of

gravel sand with fine contents.

The ground behind the quay walls moved

several meters towards the sea, resulting from the

large displacement of quay walls. These lateral

ground movement damaged the foundation piles

of bridges, buildings and industrial facilities.


Furthermore, large ground strain due to

liquefaction-induced ground movement ruptured

buried pipes of lifeline systems such as gas, water,

electricity and sewer. A great number of breakages

of buried pipes resulted in the out of service to

numerous customers during a long period. These

liquefaction-induced ground displacement had not

been taken in the consideration in the earthquake

resistant design codes before the 1995 Kobe

earthquake.

Figure 1.6 Fall of A Bridge Girder due to

Movement of its Foundation Caused by

Liquefaction-Induced Ground Displacement

When we learn the lessons from the Kobe

earthquake, we should keep in mind the fact that

some conditional factors mitigated the disaster.

For one example had the earthquake struck the

Shinkansen (bullet train) traveling on elevated

railway bridges one hour later, it would have run

off the rails and caused disastrous train accidents.

The same can be said of the collapse of subway

stations. Concrete slabs along with their

overburden soil collapsed onto subway tracks. If

subway trains had been stopped there or had

smashed into the collapsed sections, additional

serious damage would have resulted.

There were other factors that contributed to

lessening the secondary damage. One important

1-4

factor was that the earthquake struck early in the

morning. If the earthquake had struck a few hours

later during the rush hour, the results would have

been much more tragic. Another factor was that

dawn broke over the disaster-stricken area after

the earthquake. The daylight aided the evacuation

of victims and the rescue of people trapped under

collapsed houses. If the earthquake had struck at

midnight, the death toll would have been much

greater.

It is highly important to investigate into the

causes of damage to the structures and to apply

the results in future preventive measures against

earthquakes, but we should also pay our full

attention on the above-mentioned hidden lessons.

1.2 Key Concepts for Earthquake Resistant

Design.The JSCE (Japan Society of Civil Engineers)

organized a Special Task Committee of

Earthquake Resistance of Civil Engineering

Structures ill March 1995, about two months after

the Kobe earthquake, to discuss various subjects,

such as what an earthquake resistant capability of

civil engineering structures should be in the future

through the lessons from the Kobe earthquake.

The committee first discussed whether the strong

earthquake motions that had occurred in Kobe

area should be taken into account in the future

earthquake resistant design of civil engineering

structures. According to researchers on active

faults, in Japan the return period of the activity of

the earthquake fault is 500 to 2,000 years.

Assuming that the return period of the fault

activity is 1,000 years and the service life of civil

engineering structures is about 50 years, a

probability that the structures would undergo such

strong earthquake motions as those observed at

the Kobe earthquake during the serviceable life is


only five percent. The subject of the discussions

by the JSCE committee was how to treat great

disasters with such low probability of occurrence.

The JSCE proposed two key concepts for

earthquake resistant design of civil engineering

structures based on the discussions by the

committee. Those are two levels of ground

motions for earthquake resistant design and so

called performance-based design.

JSCE said that the resistance of civil

engineering structures against future earthquakes

should be examined by taking into the

consideration such strong earthquake motions as

observed during the Kobe earthquake in addition

to the ordinary earthquake motions that have thus

far been used for earthquake resistant design.

These two earthquake motions are respectively

called Level I and Level IT ground motions.

One of the reasons why JSCE said that the

Level IT ground motion should be taken into

account in the earthquake resistant design is

shown in Figure 1.7. This figure is a list of the

damaging earthquakes in the last century in Japan

and the numbers of casualties, and shows that

inland earthquakes of magnitude 7 and larger such

as the Kobe earthquake, which are surrounded by

squares in the figure, occurred 8 times and have a

probability of occurrence that can not be neglected

in terms of reformation of the design codes.

Figurel.7 also shows that the inland

earthquakes such as the Kobe earthquake resulted

in a greater number of causalities in comparison

with the plate boundary earthquakes in the pacific

ocean, if the 1923 Kanto and the 1900

Sanriku-Tsunami earthquake are excepted. In

these two earthquakes, the main causes of the loss

Name of M D ate Casualties (H, Kem!)da KyotoEarthquake 0 1000 2000 3000 4000 5000

Kumamoto 5.8 1889. 7.28

~~ 8.0 1891.10.28Tokyo 6.7 1894. 6.20

~20

Shonai 6.8 1894.10.22 7273

Sanriku Tsun. 7.1 1896. 6.15 31

~-------------~~~Rikuu 7.0 1896. 8.31 2091900

Gono 6.4 1909. 8.14

~Akita·-Senpoku 5.9 1914. 3.15

41Ch!iiwa-Bay 6.0 1922.12. 8

Great Kanto 7.9 1923. 9. 1 94

ita-t'Tafima 6.5 1925. 5.23~Oita Tango 7.5 1927. 3.7

ita Izu 7.0 1930.11.26~C;;;65······ ..·..···..·················J~~!'~!·

~Sanr-iku Tsun. 8.3 1933. 3. 3 2925

Oga-Hanto 7.0 1939. 5.1 r- 272

~ 7.4 1943. 9.10

~1------------ 3064

Tonankai 8.0 1944.12.7 27

lMikawa I 7.1 1945. 1.13-" ••••••••• 18~6Nanka! ---.. ·1961

8.1 1946.12.21 ...................... 144J

IFukui I P3769

7.3 1948. 6.2829

Tokecbf-oki 8.1 1952. 3. 4

Chile EQ Tsun. 8.5 1960. 5.231'-196a... -i-as - .

Niigata 7.5 1964. 6.16

t:::Tokachr-oki 7.926

- InlandE.Q.1968. 5.1652 ••••••• ' Plate Boundary (in Pacific Ocean) E.Q.Izu Hanto-roki 6.9 1974 5. 9 ;;::Izu-Oshima 7.0 1978. 1.14 30

_ - _ Plate Boundary (Tsunami)

Miyagiken-·oki 7.4 1978. 6.12 25

Nihonkat--Chubu~

287.7 1983. 5.26

INagano-Seibu I 6.8 1984. 9.14 104r------ 29

Kushiro-oki 7.8 1993. 1.15

~Hokkaido SI: 7.8 1993. 7.122.230 6308"

Hyogoken S 7.2 1995. 1.17

2000

Figure 1.7 Damaging Earthquakes and Number of Causalitiesin Last Century in Japan (c=J: Inland Earthquakes)

1-5


of human lives were the aftermath fire and the

tsunami, respectively.

However, JSCE's recommendation does not

mean that all structures should be designed and

constructed to sustain Level II earthquake motions.

It states that the earthquake resistant capability,

namely performance level of a structure should be

determined by comparing the importance of the

structure with the probability of occurrence of the

design earthquake motion. For instance, against

earthquake motions having a probability of

occurrence once or twice during the service life of

structures, e.g. Level I earthquake motions, the

earthquake resistant design should stipulate that

the deformation of structure falls within an elastic

limit and that any residual deformation does not

remain after the design earthquake. In contrast to

this, against very rare earthquake motions, e.g.

Level II earthquake motions, the performance

level of a structure should be changed according

to the importance of the structure. The

performance of structures after an encounter with

the design earthquake motion can be varied for an

example as follows; i) non-damaged and

functional, ii) slightly damaged but functional,

iii) heavily damaged and unfunctional, but

repairable, iv) collapsed and unrepairable.

The degree of importance of a structure is

determined by considering the following items;

i) effects of collapse of structures on human life

and survival, ii) effects on rescue and ambulance

operations and restoration activities immediately

after earthquakes, iii) effects on civic life after

earthquakes, iv) effects on economic activities

after earthquake, and v) effects on reconstruction

works.

The above-mentioned key concepts proposed

by JSCE were adopted in the National Disaster

Prevention Program in Japan which was newly

revised after the Kobe earthquake and were

strongly referred for the revision and development

of the earthquake resistant design codes.

1.3 Technical Subjects for Revision of

Earthquake Design Code

The adoption of the JSCE-proporsed key

concepts for earthquake resistant design raised

following technical subjects to be resolved for the

code developments.

i) Determination of Level II earthquake ground

motion.

ii) Evaluation of elasto-plastic behaviors and

ultimate strength of structures against the

Level II ground motion.

iii) Evaluation of residual deformation of earth

structures such as embankments, retaining

Probability of occurrence ofImportance of structure

design earthquake motion

I I~

Earthquake resistant capability

(Performance Level) of structure

Figure 1.8 Determination of Performance Level (Earthquake Resistant Capability)

1-6


walls and quay walls.

iv) Evaluation of liquefaction potential of

comparatively stiffer soil against Level II

ground motion

v) Effects of liquefaction-induced large ground

displacement.

How to determine the Level II ground motion

was one of the most important subjects in the

development of the design codes. There were

following three kinds of ideas;

i) Adoption of the maximum ground motion

recorded during past earthquakes including the

Kobe earthquake.

ii) Statistical approach of recorded and calculated

ground motion.

iii) Numerical Analysis of ground motion directly

from the design earthquake fault.

The first idea was introduced for the seismic

design specifications of highway bridges (Chapter

2) and the Level II ground motion was determined

based on the ground motions recorded during the

Kobe earthquake.

The second idea was adopted in the revision

of the design codes for the railway facilities

(Chapter 3) water facilities and gas supply

facilities (Chapters 4, 5).

The third idea where the :design ground

motion was numerically calculated from the fault

movement was also adopted for the railway

facilities and gas supply facilities.

The adoption of the Level II design ground

motion raised another Technical subjects. One is

how to estimate the behaviors of the structures in

the plastic region and their ultimate strength. For

an example, the earthquake resistant design of

steel structures has been generally made by the

allowable stress method. That is, the design is

made, not in a plastic region beyond an elastic

region. Research has hardly been done on the

1-7

behaviors of steel structures in plastic region. The

same can be said of the ultimate strength of buried

steel pipes of lifeline systems. If large ground

strain due to liquefaction-induced lateral ground

flow is incorporated into the earthquake resistant

design of buried pipes, strains of the pipes will

reach a plastic region. But a small amount of data

has been accumulated on the deformation

characteristics in a plastic region and ultimate

strength of buried pipes.

Further, evaluation of the and ductility of

earth structures, e.g. embankments, revetments,

retaining walls, and quay walls, is another subject

which needs research and development.

These above-mentioned technical subjects

have been progressively carried out after the Kobe

earthquake and the outcomes of the researches

was applied for the revision and the development

of the design code.

1.4 Diagnosis and Reinforcement of Existing

Structures

Although the future earthquake resistant

design of civil engineering structures will be

based on the concepts described above, an

additional problem is diagnosis and reinforcement

of existing structures. In large Japanese cities,

such as Tokyo and Osaka, there are countless civil

engineering structures similar to those damaged in

the Kobe area by the Kobe earthquake. Some of

them, e.g. highway bridges, Shinkansen lines,

subways, and quay walls, were constructed earlier

or have decayed more than those damaged in the

Kobe area. The earthquake resistant reinforcement

of these structures becomes an inevitable problem

if disaster preventive measures are taken by

predicting that earthquakes of a similar scale of

the Kobe earthquake will hit these cities.

Therefore, reinforcement of concrete piers of


highways and railways and concrete columns of

subways has been carried out by jacketing the

existing concrete with steel plates casting

additional concrete, and the other methods while

the effectiveness of those reinforcements was

confirmed by loading teats in the laboratory.

However, the diagnosis and the reinforcement of

the foundations of bridges and buildings against

the liquefaction-induced large ground

displacement has hardly been conducted.

As is clear from the damage caused by the

Kobe earthquake, most critical and urgent issue is

the reinforcement of structures on reclaimed lands,

for instance the Tokyo Bay and the Osaka Bay

areas, where in most of cases no soil improvement

has been taken against soil liquefaction, and a

huge number of buildings, bridges, and lifeline

facilities already exist there. It is urgently required

to develop technologies of soil improvement of

existing artificial grounds.

In addition, because reinforcement should be

undertaken in a proper order, it is also necessary

to develop a basic idea to decide the priority of

reinforcement. The previously mentioned

importance level of structures may be referred to

in deciding the priority of the reinforcement. That

is, the effects of structures on human life and

survival and on rescue and ambulance operations

and restoration activities immediately after

earthquake, as well as other effects:

1.5 Future Innovations of Design Codes and

Research Subjects

Most of earthquake resistant design codes for

civil engineering structures have been revised or

newly developed under the JSCE's key concepts

and based on the outcomes from the researches

after the Kobe earthquake. However, the following

technical subjects remains unresolved and needs

1-8

more detailed investigations in future.

i) Dynamic failure mechanism of steel and

concrete structures due to severe earthquake

ground motion, eg Level II ground motion,

shall be investigated through static and

dynamic loading tests of structural members

and large size structural models. Outcomes of

these studies are expected to give significant

information to establish new earthquake

resistant design method against extremely

severe earthquake ground motion.

ii) Mechanisms of large deformation and failure of

foundations against strong earthquake ground

motion and large ground deformation shall be

investigated, and effective countermeasures for

foundations against liquefaction and its induced

large ground displacement are required to be

developed.

iii) Mechanisms of occurrence of static large

ground deformation due to liquefaction shall

be studied by large scale shaking table test.

Studies on properties of perfectly liquefied soil

is essential for development of a rational

method for estimation of the ground

displacement. Furthermore, large scale shaking

table test on liquefaction-induced ground

displacement is expected to clarify the

mechanism.

iv)Reasonable techniques are expected to be

developed for diagnosis and reinforcement of

existing structures including foundations.

Furthermore, proper technology shall be

developed for the soil improvement of existing

liquefiable ground.


REFERENCES

1) Japan Society of Civil Engineers:

Proposal on Earthquake Resistance for Civil

Engineering Structures, 1996

2) Hamada, M.: Seismic Code Development for

Civil Infrastructures after the 1995

Hyogoken-nanbu (Kobe) Earthquake,

Proceedings of the 5th U.S. Conference on

Lifeline Earthquake Engineering, TCLEE,

Monograph No.16, pp922-929, 1999

3) Japan Road Association: Design

Specifications of Highway bridges, Part I

Common Part, Part IT Steel Bridges, Part ill

Concrete Bridges, Part IV Foundations, and

Part V Seismic Design, 1996

4) Seismic Design Code for Railway

Structures, published by MARUZEN,

Oct., 1999. (in Japanese)

5) Japan Water Works Association:

Seismic Design and Construction

Guidelines for Water Supply Facilities,

1997

1-9

2. 1996 SEISMIC DESIGN SPECIFICATIONS OF HIGHWAY BRIDGES

JAPAN ROAD ASSOCIATION

2.1 Introduction 2- 1

2.2 Damage Features of Bridges in The Hyogo-ken Nanbu Earthquake 2- 1

2.3 Basic Principle of Seismic Design 2- 3

2.4 Design Methods 2- 4

2.5 Design Seismic Force 2- 6

2.6 Evaluation of Displacement Ductility Factor of a Reinforced Concrete Pier 2- 7

2.6.1 Evaluation of Failure Mode 2- 7

2.6.2 Displacement Ductility Factor 2- 7

2.6.3 Shear Capacity 2- 8

2.6.4 Arrangement ofReinforcement 2- 9

2.6.5 Two-Column Bent 2- 11

2.7 Evaluation ofDisplacement Ductility of a Steel Pier 2- I I

2.7.1 Basic Concept 2- 11

2.7.2 Concrete Infilled Steel Pier 2- 12

2.7.3 Steel Pier without Infilled Concrete 2- 12

2.8 Dynamic Response Analysis 2- 13

2.9 Menshin Design 2- 14

2.9.1 Basic Principle 2- 14

2.9.2 Design Procedure 2- 15

2.9.3 Design of Menshin Devices 2- 15

2.10 Design of Foundation 2- 17

2.11 Design Against Soil Liquefaction and Liquefaction-Induced Ground Flow 2- 17

2.11.1 Estimation of Liquefaction Potential 2- 17

2.11.2 Design Treatment of Liquefaction for Bridge Foundations 2- 17

2.11.3 Design Treatment of Liquefaction-induced Ground Flow for Bridge Foundations 2- 18

2.12 Bearing Supports 2- 18

2.13 Unseating Prevention Systems 2- 19

2.14 Concluding Remarks 2- 20


2.1996 SEISMIC DESIGN SPECIFICATIONS OF HIGHWAY BRIDGESJAPAN ROAD ASSOCIATION

2.1 IntroductionHighway bridges in Japan had been

considered safe even against extremeearthquake such as the Great Kanto Earthquake(M7.9) in 1923, because various past bitterexperiences have been accumulated toformulate the seismic design method(Kawashima (1995)). Large seismic lateralforce ranging from O.2g to O.3g has beenadopted in the allowable stress designapproach. Various provisions for preventingdamage due to instability of soils such as soilliquefaction have been adopted. Furthermore,design detailings including the unseatingprevention devices have been implemented.

In fact, reflecting those provisions, numberof highway bridges which suffered completecollapse of superstructures was only 15 since1923 Great Kanto Earthquake. Based on suchevidence, it had been regarded that the seismicdamage of highway bridges had beendecreasing in recent years.

However, the Hyogo-ken nanbu Earthquakeof January 17, 1995, exactly one year after theNorthridge, California, USA, earthquake,caused destructive damage to highway bridges.Collapse and nearly collapse of superstructuresoccurred at 9 sites, and other destructivedamage occurred at 16 sites (Ministry ofConstruction, 1995a). The earthquake revealedthat there are a number of critical issues to berevised in the seismic design and seismicstrengthening of bridges in urban areas.

After the earthquake the "Committee forInvestigation on the Damage of HighwayBridges Caused by the Hyogo-ken nanbuEarthquake" (chairman : Toshio IWASAKI,Executive Director, Civil Engineering ResearchLaboratory) was formulated in the Ministry ofConstruction to survey the damage and clarifythe factors which contributed to the damage.

On February 27, 1995, the Committeeapproved the "Guide Specifications for

2-1

Reconstruction and Repair of Highway Bridgeswhich suffered Damage due to the Hyogo-kennanbe Earthquake," (Ministry of Construction1995b) and the Ministry of Constructionnoticed on the same day that the reconstructionand repair of the highway bridges whichsuffered damage in the Hyogo-ken nanbuearthquake should be made by the GuideSpecifications. It was decided by the Ministryof Construction on May 25, 1995 that theGuide Specifications should be tentatively usedin all sections of Japan as emergency measuresfor seismic design of new highway bridges andseismic strengthening of existing highwaybridges until the Design Specifications ofHighway Bridges was revised.

In May, 1995, the "Special Sub-Committeefor Seismic Countermeasures for HighwayBridges" (chairman KazuhikoKAWASHIMA, Professor of the TokyoInstitute of Technology) was formulated in the"Bridge Committee" (chairman : NobuyukiNARlTA, Professor of the Tokyo MetropolitanUniversity), Japan Road Association, to draftthe revision of· the Design Specifications ofHighway Bridges. The Special Sub-Committeedrafted the new Design Specifications ofHighway Bridges, and after the approval of theBridges Committee, the Ministry ofConstruction releasedit November 1, 1996.

This chapter summarizes the damagefeature of highway bridges by the Hyogo-kenNanbu earthquake and the new DesignSpecifications of Highway Bridges issued inNovember 1996.

2.2 Damage Features of Bridges in TheHyogo-ken Nanbu Earthquake

Hyogo-ken Nanbu earthquake was the firstearthquake which hit an urban area in Japansince the 1948 Fukui Earthquake. Although themagnitude of the earthquake was moderate(M7.2), the ground motion was much larger


Fig.2.1 Design Specifications Referredto in Design of Hanshin Expressway

Fig.2.2 compares damage of piers on theRoute 3 (Kobe Line) and Route 5 (Bay ShoreLine) of the Hanshin Expressway. Damagedegree was classified as As (collapse), A(nearly collapse), B (moderate damage), C(damage of secondary members) and D (minor

B

(b) Route 5

As

(a) Route 3

o

Although damage concentrated on thebridges designed with the older DesignSpecifications, it was thought that essentialrevision was required even in the recent DesignSpecifications to prevent damage againstdestructive earthquakes such as the Hyogo-kennanbu earthquake. The main points requiringmodifications were;(1) it was required to increase lateral capacityand ductility of all structural components inwhich seismic force is predominant so thatductility of a total bridge system be enhanced.For such purpose, it was required to upgrade

or no damage). Substructures of the Route 3and Route 5 were designed with the 1964Design Specifications and 1980 DesignSpecifications, respectively. It should be notedin this comparison that the intensity of groundshaking in terms of response spectra wassmaller at the Bay Area than the narrowrectangular area where JMA Seismic Intensitywas vn (equivalent to Modified MercalliIntensity of X-XI). The Route 3 was located inthe narrow rectangular area while the Route 5was located in the Bay Area. Keeping in mindsuch difference of ground motion, it is apparentin Fig.2.2 that about 14% of the piers on Route3 suffered As or A damage while no suchdamage was developed in the piers on theRoute 5.

Fig.2.2 Comparison of Damage Degree betweenRoute 3 and Route 5 (As: Collapse, A : NearlyCollapse, B : Moderate Damage, C : Damage of .Secondary Members, D : Minor or No Damage)

1990 DesignSpecifications

1964 or OlderDesign Specifications



than anticipated in the codes. It occurred veryclose to the Kobe City with shallow focaldepth.

Damage was developed at highway bridgeson Routes 2, 43, 171 and 176 of the NationalHighway, Route 3 (Kobe Line) and Route 5(Bay Shore Line) of the Hanshin Expressway,the Meishin and Chugoku Expressway.Damage was surveyed for all bridges onNational Highways, Hanshin Expressways andExpressways in the area where destructivedamage occurred. Total number of pierssurveyed reached 3,396 (Ministry ofConstruction, 1995a). Fig.2.1 shows DesignSpecifications referred to in design of the 3,396piers. Most of piers (bridges) which suffereddamage were designed according to the 1964Design Specifications or older DesignSpecifications. Although the seismic designmethods have been improved and amendedseveral times since 1926 based on damageexperience and progress of bridge earthquakeengineering, only a requirement for lateralforce coefficient was provided in the 1964Design Specifications or older Specifications.

2-2


correct mechanism ofa superstructure to

the "Check of Ductility of Reinforced ConcretePiers," which has been used since 1990, to a"Ductility Design Method," and to apply theDuctility Design Method to all structuralcomponents. It should be noted here that"check" and "design" is different; the check isonly to verify the safety of a structural memberdesigned by other design method, and iseffective only to increase the size orreinforcements if required, while the design isan essential procedure to determine the sizeand reinforcements,(2) it was required to include the groundmotion developed at Kobe in the earthquake asa design force in the Ductility Design Method,(3) it was required to specify input groundmotions in terms of acceleration responsespectra for dynamic response analysis moreactively,(4) it was required to increase tiereinforcements and to introduce intermediateties for increasing ductility of piers. It wasdecided not to terminate main reinforcements atmid-height for preventing premature shearfailure, in principle,(5) it was suggested to adopt multi-spancontinuous bridge for increasing number ofindeterminate of a total bridge system,(6) it was suggested to adopt rubber bearingsfor absorbing lateral displacement between asuperstructure and substructures. It was

important to considerforce transfer fromsubstructures,(7) it was suggested to include the Menshindesign (seismic isolation),(8) it was required to increase strength,ductility and energy dissipation capacity ofunseating prevention devices, and(9) it was required to consider the effect oflateral spreading associated with soilliquefaction in design of foundations at the sitevulnerable to lateral spreading.

2.3 Basic Principle of Seismic DesignTable 2.1 shows the seismic performance

level provided in the revised DesignSpecifications in 1996. The bridges arecategorized into two groups depending on theirimportance; standard bridges (Type-A bridges)and important bridges (Type-B bridges).Seismic performance level depends on theimportance of bridges. For moderate groundmotions induced in the earthquakes with highprobability to occur, both A and B bridgesshould behave in an elastic manner withoutessential structural damage. For extremeground motions induced in the earthquakeswith low probability to occur, the Type-Abridges should prevent critical failure, whilethe Type-B bridges should perform withlimited damage .

Table 2.1 Seismic Performance Levels

Importance of Bridges Design Methods

Type of Design Ground Motions Type-A Type-B Equivalent Dynamic(Standard (Important Static Lateral

Bridges) Bndges) Force Methods Analysis

Ground Motions with SeismicPrevent Damage Coefficient Step by Step

High Probability to Occur Method Analysis

Ground Motions Type-I or

with Low(Plate BoundaryEarthquakes) Prevent Limited Ductility Response

Critical Design SpectrumProbability Type-II Damage Damage Method Analysis

(Inlandto Occur Earthquakes)

2-3


In the Ductility Design Method, two typesof ground motions must be considered. Thefirst is the ground motions which could beinduced in the plate boundary-type earthquakeswith magnitude of about 8. The ground motionat Tokyo in the 1923 Kanto Earthquake is atypical target of this type of ground motion.The second is the ground motion developed inearthquakes with magnitude of about 7-7.2 atvery short distance. Obviously the groundmotions at Kobe in the Hyogo-ken nanbuearthquake is a typical target of this type ofground motions are called as Type-I andType-Il ground motions, respectively.

( Start )

Design forPrincipal

Loads

Seismic Design bySeismic Coefficient

Method

The recurrence time of the Type-IT groundmotion may be longer than that of the Type-Iground motion, although the estimation is verydifficult.

2.4 Design MethodsBridges are designed by both the Seismic

Coefficient Method and the Ductility DesignMethod as shown in Fig.2.3. In the SeismicCoefficient Method, a lateral force coefficientranging from 0.2 to 0.3 has been used based onthe allowable stress design approach. Nochange was introduced since the 1990Specifications in the Seismic Coefficient

:heck the Safety by>--~ Dynamic Response

Anal sis

UnseatingPrevention

Devices

Seismic Design by DynamicResponse Analysis (Type I

and II Ground Motions

Seismic Design byDuctility Design

Method (Type J andII Design Force)

Check the Safety byDynamic Response

Analysis (Type I and II

Ground Motion)

I

End

Fig.2.3 Flowchart of Seismic Design

2-4


Method.In the Ductility Design Method, assuming a

principle plastic hinge formed at the bottom ofpier as shown in Fig.4(a) and the equal energyassumption, a bridge is designed so that thefollowing requirement is satisfied.

Pa > he W (1)where

OR= CR (jJ. R-l) (l-r) a y (5)jJ. R = 1/2 {(he' W/Pai+ l ] (6)

in which a R = residual displacement of a pierafter an earthquake, a Ra = allowable residualdisplacement of a pier, r = bilinear factordefined as a ratio between the first stiffness(yield stiffness) and the second stiffness(post-yield stiffness) of a pier, CR = factordepending on the bilinear factor r, jJ. R =response ductility factor of a pier, and a y =yield displacement of a pier. The a aa shouldbe 11100 of a distance between the bottom of apier and a gravity center of a superstructure.

In a bridge with complex dynamicresponse, the dynamic response analysis isrequired to check the safety of a bridge after itis designed by the Seismic Coefficient Methodand the Ductility Design Method. Because thisis only for a check of the design, the size andreinforcements of structural members oncedetermined by the Seismic Coefficient Methodand the Ductility Design Methods can only beincreased if necessary. It should be notedhowever that under the following conditions inwhich the Ductility Design Method is notdirectly applied, the size and reinforcementscan be determined based on the results f adynamic response analysis as shown in Fig.2.3.The conditions when the Ductility DesignMethod should not be directly used include:(1) principle mode shapes which contribute to

(4)

(Z)khe

he =-.fZjJ.a-1

W = Wo--c» Wp (3)in which, Pa = lateral capacity of a pier, he =equivalent lateral force coefficient, W =equivalent weight, kne = lateral forcecoefficient, jJ. a = allowable displacementductility factor of a pier, Wu = weight of a partof superstructure supported by the pier, Wp =weight of a pier, and cp = coefficient dependingon the type of failure mode. The cp is 0.5 for apier in which either flexural failure or shearfailure after flexural cracks are developed, and1.0 for a pier in which shear failure isdeveloped. The lateral capacity of a pier Pa isdefined as a lateral force at the gravity centerof a superstructure.

In the Type-B bridges, residualdisplacement developed at a pier after anearthquake must be checked as

a R< a aa

where

Principal Plastic Hinge

(a) Conventional Design (b) Menshin Design (c) Bridge Supported by A Wall-type Pier

Fig.2.4 Location of Primary Plastic Hinge

2-5


bridge response are different from the onesassumed in the Ductility Design Methods,(2) more than two modes significantlycontribute to bridge response,(3) principle plastic hinges form at more thantwo locations, or principle plastic hinges arenot known where to be formed, and(4) response modes for which the equal energyassumption are not applied.

In the seismic design of a foundation, alateral force equivalent to the ultimate lateralcapacity of a pier Pu is assumed to be a designforce as

h p = Cdf PuIW (7)in which hp = lateral force coefficient for afoundation, Cdf = modification coefficient(=1.1), and W = equivalent weight by Eq.(3).Because the lateral capacity of a wall-type pieris very large in transverse direction, the lateralseismic force evaluated by Eq. (7) becomes in

most cases excessive. Therefore if a foundationhas sufficiently large lateral capacity comparedwith the lateral seismic force, the foundation isdesigned assuming a plastic hinge at thefoundation and surrounding soils as shown inFig.2A(e),

2.5 Design Seismic ForceLateral force coefficient he in Eq.(2) is

given ashe = cz : heO (8)

in which cz = modification coefficient for zone,and is 0.7, 0.85 and 1.0 depending on zone, andheo = standard modification coefficient. Table2.2 and Fig.2.S show the standard lateral forcecoefficients heo for the Type-I and the Type-Ilground motions. The Type-I ground motionshave been used since 1990 (1990Specifications), while the Type-Il groundmotions were newly introduced in the 1996

Table 2.2 Lateral Force Coefficient heo in the Ductility Design Method

(a) Type-I Ground Motions

Soil Condition Lateral Force Coefficient fuco

Group Ifuco=0.7 for T < 1.4 hco=0.876T2

/J for T > 1.4

(stiff)

Group II fueo=1.51TI/J fueo=0.85 fueo=1.16T2/J

(fueo > 0.7)(moderate) for T < 0.18 for 0.18 < T < 1.6 for T> 1.6

Group III beo=1.51TI/J fueo=1.0 fueo= 1.59T2/3

(beo > 0.7)(soft) for T < 0.29 for 0.29 < T < 2.0 for T> 2.0

(b) Type-Il Ground Motions

Soil Condition Lateral Force Coefficient fueo

Group I fueo=4.46T/J beo=2.00 beo=1.24T 4/J

(stiff) for T < 0.3 for 0.3 < T < 0.7 for T> 0.7

Group II heo=3.22T/J heo=1.75 beo=2.23T4/J

(moderate) for T < 0.4 for 0.4 ~ T < 1.2 for T> 1.2

Group III hco=2.38T/3 beo=1.50 beo=2.57T'3

(soft) for T < 0.5 for 0.5 < T < 1.5 for T> 1.5

2-6


accelerations excursion is longer in the Type-Iground motions than the Type-II groundmotions. As will be described later, such adifference of the duration has been taken intoaccount to evaluate the allowable displacementductility factor of a pier.

2.6 Evaluation of Displacement DuctilityFactor of A Reinforced Concrete Pier

2.6.2 Displacement Ductility FactorThe allowable displacement ductility factor

of a pier j.J. a in Eq.(2) is evaluated asa u - a y

j.J. a == 1 + Q:' a y (9)

in which Q:' = safety factor, a y == yielddisplacement of a pier, and a u = ultimatedisplacement of a pier. As well as the lateralcapacity of a pier Pa in Eq.(I), the a y anda u are defined at the gravity center of asuperstructure. In a reinforced concrete singlepier as shown in Fig.2.4(a), the ultimatedisplacement a u is evaluated as

a u== a y+ (et> u- et> y) Lp(h - Lp/2) (10)in which et> y = yield curvature of a pier atbottom, et> u == ultimate curvature of a pier atbottom, h == height of a pier, and Lp == plastichinge length of a pier. The plastic hinge lengthis given as

2.6.1 Evaluation of Failure ModeIn the ductility design of reinforced concretepiers, the failure mode of the pier is evaluatedas the first step. Failure modes is categorizedto 3 types based on the bending capacity andshear capacity of the pier as

1) Pu < Ps : bending failure2) Ps < Pu < Pso : bending to shear failure3) PsO < Pu : shear failure

in which Pu == bending capacity, Ps == shearcapacity in consideration of the effect of cyclicloading, and Pso == shear capacity withoutconsideration of the effect of cyclic loading.

The ductility factor and capacity of thereinforced concrete piers are determinedaccording to the failure mode as describedlater.

Specifications. It should be noted here that theheO at stiff site (Group I) has been assumedsmaller than the heO at moderate (Group II)

and soft soil (Group III) sites in the Type-Iground motions as well as the seismiccoefficients used for the Seismic CoefficientMethod. The Type-I ground motions wereessentially estimated from an attenuationequation for response spectra that was derivedfrom a statistical analysis of 394 componentsof strong motion records. Although theresponse spectral accelerations at short naturalperiod are larger at stiff sites than at soft soilsites, the tendency has not been explicitlyincluded in the past. This was because damagehas been more developed at soft sites than atstiff sites. To consider such fact, the designforce at stiff sites has been assumed smallerthan that at soft sites even at short naturalperiod. However being different from such atraditional consideration, the Type-II groundmotions were determined by simply takingenvelops of response accelerations of majorstrong motions recorded at Kobe in theHyogo-ken nanbu Earthquake. It wasconsidered appropriate to set realistic groundmotions.

Although the acceleration response spectralintensity at short natural period is higher in theType-II ground motions than in the Type-Iground motions, the duration of extreme

2.5u

- - - - - Group I..c.Y

- - - Group II Type I.....,2c -- ~ Group IIIa; I

U I --. - - - - - Group I<;:: - Group II Type II~ 1.5a;a Group 111

Ua;UL-

au.,

ro 0.5L-a; ---.....,ro

-..l0

0 2 3 4

Natural Period T (5)

Fig.2.S Type I and Type II Ground Motionsin the Ductility Design Method

2-7


(17)< 0.0184Ah

p s =

(

C eeC ell =

C ee +

(18)It is important to note here that the ultimate

strain c eu depends on the types of groundmotions; the c eu for the Type-II groundmotions is larger than that for the Type-Iground motions. Based on a loading test, it isknown that a certain level of failure in a piersuch as a sudden decrease of lateral capacityoccurs at smaller lateral displacement in a piersubjected to a loading hysteresis with morenumber of load reversals. To reflect such afact, it was decided that the ultimate straine eu should be evaluated by Eq.(18),

depending on the type of ground motions.

and a = 0.2 and j3 = 0.4 for a rectangularpier), and p s = tie reinforcement ratiodefmed as

sdin which Ah = area of tie reinforcements, s =space of tie reinforcements, and d = effectivewidth of tie reinforcements.

The ultimate curvature ¢ u is defmed as acurvature when concrete strain at longitudinalreinforcing bars in compression reaches anultimate strain e eu defined as

for Type I ground motions0.2 (J' ee

Edesfor Type II ground motions

(16)

(14)

(15)

e < E cu)

(12)

(13)

(J' ek 2

Ee e ee

Edes = 11.2p s (J' sy

in which (J' ek = design strength of concrete,(J' sy = yield strength of reinforcements, aand j3 = coefficients depending on shape ofpier ( a =1.0 and j3 =1.0 for a circular pier,

n=Ee E ee - (J' ee

in which a cc = strength of confined concrete,Ee = elastic modules of concrete, e cc = strainat maximum strength, and Edes = gradient atdescending branch. In Eq.(12), (J' cc, C ee andEses are determined as

a ee = (J' ek + 3.8 a p s (J' sy

C ee = 0.002+0.033 j3 P s (J' sy(J' ek

Lp= 0.2h - O.lD (O.lD < Lr < 0.5D) (11)in which D is a width or a diameter of a pier.

The yield curvature ¢ y and ultimatecurvature ¢ u in Eq.(10) are evaluatedassuming a stress-strain relation ofreinforcements and concrete as shown inFig.2.6. The stress (J' e - strain t: e relation ofconcrete with lateral confinement is assumed as

[

Ee e e{l _ 1 ~)n-l}n C ee

(0 < C e < e cc )(J'e=

ee - Edes(c e- c cc)(c cc< e

Stress (}sStress O'c

Strain e,

0.80' cc - - - - --I

_____ L _

I

III

II

I

II

II

r

£cu Strain E.c

(a) Reinforcing Bars (b) Concrete

Fig.2.6 Stress and Strain Relation of Confined Concrete and Reinforcing Bars

2-8


Table 2.3 Safety Factor a in Eq.(9)

Type of Bridges Type-I Ground Motions Type-II Ground Motions

Type-B 3.0 1.5

Type-A 2.4 1.2

Table 2.4 Modification Factor On Scale Effect for Shear Capacity Shared by Concrete

Effective Width of Section d (m) Coefficient Ce

d ;;:;; 1 1.0

d::::3 0.7

d::::5 0.6

d ~ 10 0.5

(19)

(20)

(21)

Therefore, the allowable ductility factor u a

depends on the type of ground motions; theu a is larger in a pier subjected to the Type-ITground motions than a pier subjected to theType-I ground motions.

It should be noted that the safety factor ain Eq.(9) depends on the type of bridges as

well as the type of ground motions as shown inTable 2.3. This is to preserve higher seismicsafety in the important bridges, and to takeaccount of the difference of recurrent timebetween the Type-I and the Type-IT groundmotions.

2.6.3 Shear CapacityShear capacity of reinforced concrete piers

is evaluated by a conventional method asPs :::: Sc + SsSc :::: 10 Cc Ce Cpt reb d

Ss > Aw a sy d (sin e+cos e)10 x 1.1Sa

in which Ps :::: shear capacity, Sc :::: shearcapacity shared by concrete, Ss :::: shearcapacity shared by tie reinforcements, t: c =shear stress capacity shared by concrete, Cc =modification factor for cyclic loading (0.6 forType-I ground motions, 0.8 for Type-II groundmotions), Ce = modification factor for scaleeffect of effective width, Cpt :::: modificationfactor for longitudinal reinforcement ratio, b, d

2-9

:::: width and height of section, Aw :::: sectionalarea of tie reinforcement, (J' sy:::: yield strengthof tie reinforcement, e = angle betweenvertical axis and tie reinforcement, and a =spacing of tie reinforcement.

The modification factor on scale effect ofeffective width, Ce, was based on theexperimental study of loading tests of beamswith various effective height and was newlyintroduced in the 1996 Specifications. Table2.4 shows the modification factor on scaleeffect.

2.6.4 Arrangement of ReinforcementFig.2.7 shows suggested arrangement of tie

reinforcement. Tie reinforcement should bedeformed bars with a diameter equal or largerthan 13 mm, and it should be placed in mostbridges at a distance of no longer than 150mm.In special cases such as the bridges with pierheight taller than 30m, the distance of tiereinforcement may be increased at height sothat pier strength should not be sharplydecreased at the section. Intermediate tiesshould be also provided with the same distancewith the ties to confine the concrete. Space ofthe intermediate ties should be less than 1m.


~~ uu u u u

p(

(

~ ~ ~~(b) Semi-square Section

(a) Square Section

(c) Circular Section (d) Hollow SectionFig.2.7. Confinement of Core-concrete by Tie Reinforcement

Lpn Lp

r:o--6--r--o---(c)}-+----------j--<!o»-O-.....-~-o

LPC

o o.

o Node

@ Plastic hinge

Lp Plastic Hinge Length

Rigid Member

Elastic Member

Fig.2.S Analytical Idealization of A Two-Column Bent

2-10


2.6.5 Two-Column BentTo determine the ultimate strength and

ductility factor for two-column bents, it ismodeled as the frame model with plastic hingesat the both end of lateral cap beam andcolumns as shown in Fig.2.8. Each elasticframe member has the yield stiffness which isobtained based on the axial load by the deadload of the superstructure and the column. Theplastic hinge is assumed to be placed at the endpart of a cap beam and the top and bottom partof each column. The plastic hinges aremodeled as spring elements with bilinearmoment-curvature relation. The location ofplastic hinges is half distance of the plastichinge length off from the end edge of eachmember, where plastic hinge length LP isassumed to be Eq.(ll).

When the two-column bent is subjected tothe lateral force in the transverse direction,axial force developed in the beam and columnsis affected by the aoolied lateral force.Therefore, the horizontal force-displacementrelation is obtained through the static push-overanalysis considering axial force N - moment Minteraction relation. The ultimate state of eachplastic hinges is obtained by the ultimateplastic angle e pu as

e pu = (¢ uI¢ y -1) Lp ¢ y (22)in which ¢ u =ultimate curvature and ¢ y =yield curvature.

(a) Fracture of Comers

The ultimate state of the whole two-bentcolumn is determined so that all 4 plastichinges developed reach the ultimate plasticangle.

2.7 Evaluation of Displacement Ductility ofA Steel Pier2.7.1 Basic ConceptTo improve seismic performance of a steelpiers, it is important to avoid specific brittlefailure modes. Fig.2.9 shows the typical brittlefailure mode for rectangular and circular steelpiers. The followings are the countermeasuresto avoid such brittle failure modes and toimprove seismic performance of steel piers:1) fill the steel column with concrete2) improve structural parameters related tobuckling strength

• decrease the width/thickness ratio ofstiffened palates of rectangular piers or thediameter/thickness ratio of steel pipes

• increase the stiffness of stiffeners· reduce the diaphragm spacing· strengthen comers using the comer plates

3) improve welding section at the comers ofrectangular section4) eliminate welding section at the comers byusing round comers

(b) Elephant Knee Buckling

Fig.2.9 Typical Brittle Failure Modes of Steel Piers

2-11


(24)

(23)

2.7.2 Concrete Infilled Steel PierIn a concrete infilled steel pier, the lateralcapacity Pa and the allowable displacementductility factor jJ. a in Eqs.(l) and (2) areevaluated as

Pu - PyPa == Py + --'-'----'-

(]I

-(1 O'u-O'y)~jJ.a- +(]I a y Pa

in which Py and Pu == yield and ultimate lateralcapacity of a pier, a y and a u == yield andultimate displacement of a pier, and (]I ==safety factor (refer to Table 2.3). The Pa andthe jJ. a are evaluated idealizing that a concreteinfilled steel pier resists flexural moment andshear force as a reinforced concrete pier. It isassumed in this evaluation that the steel section

be idealized as reinforcing bars and that onlysteel section resists axial force. A stress vs.strain relation of steel and concrete as shown inFig.2.10 is assumed. The height of infilledconcrete has to be decided so that. bucking isnot developed above the infilled concrete.

2.7.3 Steel Pier without Infilled ConcreteA steel pier without infilled concrete must

be designed with the dynamic responseanalysis. Properties of the pier need to bedecided based on a cyclic loading test.Arrangement of stiffness and welding at comermust be precisely evaluated so that brittlefailure should be avoided.

0.10iO. 05

0.04 0.06· 0.08

Strain t: s

ay

0Of.]

'"~l-<.....

1;1)

0 iEy

0.02

o

0.01 0.02 0.03 0.04 0.05

Strain t: s

(a) Steel (Tension Side) (b) Steel (Compression Side)

o. E

a= 2a,<!' x 0.00827 (2Ec )

0.00827

o! ! ! ! t

0.01 0.02 0.03 0.04 0.05

Strain t: c

(c) ConcreteFig.2.10 Stress-Strain Relation of Steel and Concrete

2-12


modification coefficient for damping ratiogiven as

Table 2.5 and Fig.2.ll show the standardacceleration response spectra (damping ratioh=0.05) for the Type-I and Type-Il groundmotions.

It is recommended to use at least threeground motions per analysis, and take anaverage to evaluate the response.

In the dynamic analysis, modal dampingratios have to be carefully evaluated. Todetermine themodal damping aratios, a bridgemay be divided into several sub-strucctures in

which energy dissipating mechanism isessentially the same. If one can specify a

2.8 Dynamic Rresponse AnalysisDynamic response analysis is required in

the bridges with complex dynamic response tocheck the safety factor of the static design.Dynamic response analysis is alas required as a"design" tool in the bridges for which theDuctility Design Method is not directlyapplied. In dynamic response analysis, groundmatins which are spectral fitted to thefollowing response spectra are used;

S I = cz ' CD' S I 0 (25)S II = cz ' CD • SilO (26)

in which S I and S II = acceleration responsespectrum for Type-land· Ty6e-II groundmotions, S I 0 and SilO = standard accelerationresponse spectrum for Type-land Type-Ilground motions, respectively, CZ = modificationcoefficient for zone (refer to Eq.(8», and CD =

CD =1.5

40hi + 1+ 0.5 (27)

Table 2.5 Standard Acceleration Response Spectra(a) Type-I Response Spectra SIO

Soil Condition Response Acceleration S10 (gal=cm/sec2)

Group I SIO=700 for Ti ~ 1.4 SIO=980!Ti for Ti > 1.4

SJo=1,505T."3 S1O=850 SJO=1,360!TiGroup II (SJO ~. 700)

for Ti < 0.18 for 0.18 ~ Ti ~ 1.6 for Ti > 1.6

SJO=1,511T."3 SIO=l,OOO SJO=2,000/TiGroup III (SJO ~ 700)

for 0.29 ~ Ti ~ 2.0 for Ti > 2.0for T; < 0.29

(b) Type-Il Response Spectra SilO

Soil Condition Response Acceleration Suo (gal=cm/sec2)

SII0=4,463Ti2/3 SII0=2,000513

SII0=1,104/TiGroup I

for Ti ~ 0.3 for 0.3 ~ T; ~ 0.7 for Tj > 0.7

Sno=3,224T/13 SII0=1,750 SJJO=2,371/T/'3Group I!

for T, < 0.4 for 0.4 ~ Ti ~ 1.2 for T ; > 1.2

3no=2,381Ti213 SII0=1,500 SII0=2,948/T;513Group III

for Ti < 0.5 for 0.5 ~ Ti ~ 1.5 for T; > 1.5

2-13


Fig.2.ll Type I and Type II Standard AccelerationResponse Spectra

damping ratio of each sub-structure for a givenmode shape, the modal damping ratio for i-thmode, hi, may be evaluated as

nL ¢ ij T.hij . Kj' ¢ ij

hi = j=l (28)<t:> iT'K' <t:> i

in which hij = damping ratio of j-thsubstructure in i-th mode, ¢ ij = mode vectorof j-th substructure in i-th mode, kj = stiffnessmatrix of j-th substructure, K= stiffness matrixof a bridge, and <t:> i = mode vector of a bridgein i-th mode, and is given as

cP iT = {¢ u", ¢ iZT, •••••• , ¢ inT

} (29)

Table 2.6 shows recommended dampingratios for major structural components.

2.9 Menshin Design2.9.1 Basic PrincipleImplementation of the Menshin bridges shouldbe carefully decided from not only seismicperformance but function for traffic andmaintenance point of view, based on theadvantage and disadvantage of increasingnatural period The Menshin design should notbe adopted at the following conditions;

1) sites vulnerable to lose bearing capacitydue to the soil liquefaction and the lateralspreading,

2) bridges supported by flexible columns,3) soft soil sites where potential resonance

with surrounding soils could be developed byincreasing the fundamental natural period,and

4) bridges with uplift force at bearings.It is suggested that the design should be

made with an emphasis on an increase ofenergy dissipating capability and a distributionof lateral force to as many substructures aspossible. To concentrate the hystereticdeformation at not piers but bearings, thefundamental natural period of a Menshinbridge should be about 2 times or longer thanthe fundamental natural period of the samebridge supported by the conventional bearings.It should be noted that an elongation of naturalperiod aiming to decrease the lateral forceshould not be attempted.

4

Type I

3

- - - - - Ground I

- - • Ground IJ

-- Ground III

----- Ground I

- - - Ground II Type U

2

Natural Period (5)

\\,,

\

\___ ..J.. __

\. <, "-- --- - -- - - - ....- , ...........' ... --.... ---

... -... """'----~~~~~~~~~~~j

r

, \, \,_-"", __" I .. \" I: I: I

r-r-r-r-r-r-it-r-r-r-r-c-:

2.5

-;) 2

-.. 1.5;:;'-''-'<'-'~

;::'-'c::

00

Table 2.6 Recommended Damping Ratios for Major Structural Components

Structural Elastic Response Nonlinear Response

Components Steel Concrete Steel Concrete

Superstructure 0.02 ~ 0.03 0.03 ~ 0.05 0.02 ~ 0.03 0.03 ~ 0.05

Rubber Bearings 0.02 0.02

Mensbin Bearings Equivalent Damping Ratio Equivalent Damping Ratioby Eq.(26) by Eq.(26)

Substructures 0.03 ~ 0.05 0.05 ~ 0.1 0.1 rv 0.2 0.12 ~ 0.2

Foundations 0.1 ~ 0.3 0.2 rv 0.4

2-14


(32)

reduced, as large as 30%, by the modificationcoefficient CE depending on the modal dampingratio of a bridge.

Modal damping ratio of a menshin bridge hfor the fundamental mode is computed asEq.(32). In Eq.(32), hsi = damping ratio of i-thdamper, hPi = damping ratio of i-th pier orabutment, hFui = damping ratio of i-thfoundation associated with translationaldisplacement, hF e i = damping ratio of i-thfoundation associated with rotationaldisplacement, Kn = equivalent stiffness of i-thpier or abutment, KFui = translational stiffnessof i-th foundation, KF e i = rotational stiffnessof i-th foundation, UBi =design displacement ofi-th Menshin device, and. H = distance from abottom of pier to a gravity center of a deck.

In the Menshn design, the allowabledisplacement ductility factor of a pier jJ. m inEq.(30) is evaluated by

2.9.3 Design of Menshin DevicesSimple devices stable against extremeearthquakes have to be used. The bearings haveto be anchored to a deck and substructures withbolts, and should be replaceable. The clearancehas to be provided

(33)

KBi KBi KBi'H2

L K B j'U B j 2(1 + -- + -- + )KPi KFui KFBj

smaller allowable ductility factor in themenshin design is to limit the hystereticbehavior of a pier at the plastic hinge zone sothat principle hysteretic behavior occurs at themenshin devices as shown in Fig.2.4(b).

hem =

h=

2.9.2 Design ProcedureMenshin bridges are designed by both the

a u - a yjJ. m = 1 + _---=--_-'---L.-

amoy

in which a m is a safety factor used inMenshin design, and is given as

jJ.m=2a O~

where a is the safety factor in theconventional design (refre to Table 2.3).Eq.(34) means that the allowable displacementductility factor in the menshin design jJ. m

should be smaller than the allowabledisplacemnent ductility factor u a by Eq.(2) inthe conventional design; The reason for the

Seismic Coefficient Method and the DuctilityDesign Method. In the Seismic CoefficientMethod, no reduction of lateral force from theconventional design is mae.

In the Ductility Design Method, theequivalent lateral force coefficient kbcm in theMenshin design is evaluated as

hcm(30)

.["2 jJ. m-1hcm = CE' hc (31)

in which hcm = lateral force coefficient inmenshin design, jJ. m = allowable ductilityfactor of a pier, CE = modification coefficientfor energy dissipating capability (refer toTable2.7), and knc =lateral force coefficient byEq.(8). Because the hc is the lateral forcecoefficient for a bridge supported by theconventional bearings, Eq.(31) means that thelateral force in the Menshin design can be

"'K 2(h hPi'KBi hFui'KBi hFBi'KBi'H2

L. B i ·u B i . B i + K + K + TT )Pi Fui ..L"'!t..~81

Table 2.7 Modification Coefficient for Energy Dissipation Capability

Damping Ratio for 1st Mode h Coefficient c E

h < 0.1 1.0

0.1 ~ h < 0.12 0.9

0.12 ;;;;;; h < 0.15 0.8

h ~ 0.15 0.7

2-15


(36)F(UBe) - F(-uBe)

2UBe~W

hs :::: 2 7C W (37)

use :::: cs : UB (38)in which F(u) :::: restoring force of a device at adisplacement U , UBe :::: effective designdisplacement, ~ W = energy dissipated percycle, W = elastic strain energy, 'and CB ::::coefficient to evaluate effective displacement(=0.7).

The equivalent stiffness KB and equivalentdamping ratio hs of a Menshin device areevaluated as

(35)UB ::::

between a deck and an abutment or betweenadjacent decks.

Isolators and dampers must be designed fora desired design displacement us. The designdisplacement UB is evaluated as

khem WuK:e

in which hem :::: equivalent lateral forcecoefficient by Eq.(3l), KB :::: equivalentstiffness, and Wu :::: dead weight of asuperstructure. It should be noted that becausethe equivalent lateral force coefficient hemdepends on the type of ground motions, thedesign displacement us also depends on it.

Curvature

VerticalDisplacement

Y : Yield- Mp: Plastic Moment

o

c v:EM. i---------o

;'8 My -_.:

OJ)c

"t:lCOJ~

(b)Vertical Force YS. VerticalDisplacement Relation

Vertical Force at Pile Top P...

Ultimate Bearing CapacityP" - -r-r- _

_____...L - PT'

Ultimate Pull-outForce

Curvature

C: CrackY: YieldU : Ultimate :

kHE

.....C l:E Mu - ••••••• _.- •••••• --- •••• -.-.------

o~

yMy.-.----

Cb.O Ma ---C:acOJ

~

Max. HorizontalReaction Force

:ti'\~: /:v~, ,

, 1 : I~, "

'I I I I I' ... _..l I ... _.J 1 __ .J

(a) Analytical Model

KVE,~~

,

Horizontal Displacement

OJ<.J1-0o~

c~<.J~ PHU

P::

BcoN'J:o:I: 0'---'---------

(c) Horizontal Force vs, (d) Moment vs. CurvatureHorizontal Displacement Relation of ReinforcedRelation Concrete Piles

(e) Moment vs. CurvatureRelation of Steel Pipe

Piles

Fig.2.12Idealized Nonlinear Model of A Pile Foundation

2-16


2.10 Design of FoundationThe evaluation methods of ductility and

strength of foundations such as pilefoundations and caisson foundations was newlyintroduced in the 1996 Specifications.

In a pile foundation, a foundation is soidealized that a rigid footing is supported bypiles which are supported by soils. The flexuralstrength of a pier defined by Eq.(7) shall beapplied as a seismic force to foundations at thebottom of the footing together with the deadweight superstructure, pier and soils on thefooting. Fig.2.l2 shows the idealized nonlinearmodel of a pile foundation. The nonlinearity ofsoils and piles is considered in the analysis.

The safety of the foundation shall bechecked so that 1) the foundation shall, notreach the yield point of a foundation, 2) if theprimary nonlinearity is developed in thefoundations, the response displacement shall beless than displacement ductility limit, and 3)the displacement developed in the foundationshall be less than allowable limit. Theallowable ductility and allowable limit ofdisplacement were commented as 4 indisplacement ductility, 40cm in horizontaldisplacement and a.025rad in rotation angle.

For a caisson type foundation, thefoundation is modeled as a reinforced concretecolumn which is supported by soil springmodel and the safety is checked in the sameway as the pile foundations.

2.11 Design Against Soil Liquefaction andLiquefaction-induced Ground Flow2.11.1 Estimation of Liquefaction Potential

Since the Hyogo-ken nanbu Earthquake of1995 caused liquefaction even at coarse sandor gravel layers which had been regardedinvulnerable to liquefy, a gravel layer wasincluded in the soil layers that requireliquefaction potential estimation. Soil layersthat satisfies the following conditions isestimated to be potential liquefaction layers:1) saturated soil layer which is located within20m deep under the ground surface and inwhich ground water level is within 10m deep.

2-17

2) soil layer in which fine particle content ratioFe is equal orless than 35% or plasticity indexIF is equal or less than 15.3) soil layer in which mean grain size Dso isequal or less than 10mm and 10% grain sizeDIO is equal or less than Imm.

Liquefaction potential is estimated by thesafety factor against liquefaction FL as

FL = RJL (35)where, FL = liquefaction resistant ratio, R =dynamic shear strength ratio and L = sharestress ratio during an earthquake. The dynamicshear strength ratio R may be expressed as

R = cw Rc (36)where, Cw = corrective coefficient for groundmotion characteristics (1.0 for Type-I groundmotions, 1.0-2.0 for Type-IT ground motions),and Rc = cyclic triaxial strength ratio. Thecyclic triaxial strength ratio was estimated bylaboratory tests with undisturbed samples byin-situ freezing method.

The shear stress ratio during an earthquakemay be expressed as

L = ru kne a via v' (37)where, ra = modification factor shear stressratio with depth, :he = design seismiccoefficient for the evaluation of liquefactionpotential, (J" v = total loading pressure, (J" v'

= effective loading pressure.It should be noted here that the design

seismic coefficient for the evaluation ofliquefaction potential :he is ranging from 0.3 to0.4 for Type-I ground motions, and from 0.6 to0.8 for Type-IT ground motions.

2.11.2 Design Treatment of Liquefaction forBridge Foundations

When the liqeufaction occurs, the strengthand the bearing capacity of a soil decreases. Inthe seismic design of highway bridges, soilconstants of a sandy soil layer which is judgedliable to liquefy are reduced according to theFL value. The reduced soil constants arecalculated by multiplying the coefficient DE inTable2.8 to the soils constants estimated on an


Table 2.8 Reduction Coefficient for Soil Constants due to Soil Liquefaction

Range of FLDepth from the Present Dynamic Shear Strength Ratio R

Ground Surface x (m) R ~ 0.3· 0.3<R

0~x;;;;'10 a 1/6FL~ 1/3 ----------------------- ----------------- ----------------

10<x ~ 20 1/3 1/3

a ~x;;;; 10 1/3 2/31/3<FL~ 2/3 ----------------------- ----------------- ----------------

10<x ~ 20 2/3 2/3

0 ~x;;;; 10 2/3 12/3<FL ~ 1 ----------------------- ----------------- ----------------

10<x ~ 20 1 1

assumption that the soil layer does not liquefy.

2.11.3 Design Treatment ofLiqeufaction-Induced Ground Flow forBridge Foundations

When the liquefaction-induced ground flowthat may affect bridge seismicity is likely tooccur, this influence was included in therevised Design Specifications in 1996. Thecase in which the ground flow that may affectbridge seismicity is likely to occur is generallythat the ground is judged to be liquefiable andis exposed to biased earth pressure, forexample, the ground behind a seasideprotection wall. The effect ofliquefaction-induced ground flow is consideredas the static force acting on structure. Thismethod premises that the surface soil is of thenon-liqeufiable and liquefiable layers, and theforces equivalent to the passive earth pressureand 30% of the overburden pressure areapplied to the structure in the non-liquefiablelayer and liquefiable layer, respectively.

The seismic safety of a foundation ischecked by confirming the displacement at thetop of foundation caused by ground flow doesnot exceed an allowable value, in which afoundation and the ground are idealized asshown in Fig.2.l2. The allowable displacementof a foundation may be taken as two times theyield displacement of a foundation. In thisprocess, the inertia force of structure is notnecessary to be considered simultaneously,

2-18

because the liquefaction-induced ground flowmay take place after the principle groundmotion.

2.12 Bearing SupportsThe bearings are classified into two groups;

the first is the bearings which resist the seismicforce of Eq.(2), and the second is the bearingswhich resist the seismic force considered in theSeismic Coefficient Method. The first and thesecond bearings are called as the Type-Bbearings and the Type-A bearings, respectively.Seismic performance of the Type-B bearings is,of course, much higher than the Type-Abearings. In the Type-A bearings, adisplacement limiting device, which will bedescribed later, has to be co-installed in bothlongitudinal and transverse directions, while itis not required in the Type-B bearings. Becauseof the importance .of bearings as one of themain structural components, the Type-Bbearings should be used in the menshinbridges.

The uplift force applied to the bearingsupports is specified as

Ru :::: R» - .r Rheq2 + Rvec{ (38)

in which Ru =design uplift force applied to thebearing support, RD = dead load ofsuperstructure, Rheq and Rveq are verticalreactions caused by the horizontal seismic forceand vertical force, respectively. Fig.2.13 showsthe design forces for thebearing supports.


1Center of gravity

±RV EQRHEQI

(i- )

jh B

!I

_----Jr==F=...Jc..J.. --'r-iL..Ji.-!:, r=..-==...-.L.l:.-, .:t.... _.

.-+

along the bridge axis, and is 0.0025, 0.00375,and 0.005 for Group-I, II and ill sites,respectively, L= distance which contributes tothe relative displacement of ground (m), and 1= span length (m). If two adjacent deck aresupported by a pier, the lager span lengthshould be 1 in evaluating the seat length.

Inthe menshin deisgn, in addition to theabove requirements, the following

Fig.2.13 Design Forces for Bearing Supports

each support against transverse response. Thedisplacement limiting device is not generallyrequired if the Type-B bearings are used. But,even if the Type-B bearing is adopted, it isrequired in skewed bridges, curved bridges,bridges supported by columns with narrowcrest, bridges supported by few bearings perpiers, and bridges constructed at the sitesvulnerable to lateral spreading associated withsoil liquefaction.

The seat length SE is evaluated asSE = UR +UG > SEM (39)SEM = 70 + 0.51 (40)UG =100' C G'L (41)

in which UR = relative displacement (cm)developed between a superstructure and asubstructure subjected to a seismic forceequivalent to the equivalent lateral forcecoefficient he by Eq.(2), UG = relativedisplacement of ground along the bridge axis,SEM = minimum seat length (cm), C G =ground strain induced during an earthquake

2.13 Unseating Prevention SystemsUnseating prevention measures are required

for the highway. bridges. The measuresrequired for the highway bridges are as:1) the unseating prevention systems have to beso designed that unseating of a superstructurefrom their supports can be prevented even ifunpredictable failures of the structural membersoccur,2) the unseating prevention systems areconsisted of providing enough seat length, afalling-down prevention device, a displacementlimiting device, and a settlement preventiondevice,3) enough seat length must be provided and afalling-down prevention device must beinstalled at the ends of a superstructures againstlongitudinal response. If the Type-A bearingsare used, a displacement limiting device has tobe further installed at not only the ends of asuperstructure but each intermediate support ina continuous bridge, and4) if the Type-A bearings are used, adisplacement limiting device is requested at

2-19


copnsiderations have to be made.1) To prevent collisions between a deck and anabutment or between two adjacent decks,enough clearance must be provided. Theclearance between those structural componentsSB shall be evaluated as

UB + LAbetween a deck and an abutment

CB"UB + LAbetween two adjacent decks

(42)

in which UB = design displacement of menshindevices (em) by Eq.(35), LA =:: redundancy of aclearance (generally + 1.5cm), and CB =modification coefficient for clearance (refer toTable 2.9). The modification coefficient CB wasdetermined based on an analysis of the relativedisplacement response spectra. It depends on adifference of natural periods 6. T = T. - T2 (T.> T2), in which Ti and T2 represent the naturalperiod of the two adjacent bridge systems.

2) The clearance at an expansion joint LE isevaluated as

LE = us + LA (43)

in which UB = design displacement of menshindevices (cm) by Eq.(35), and LA = redundancyof a clearance (generally -+- 1.5cm).

of the new Seismic Design Specifications ofHighway Bridges issued in 1996 as well as thedamage features of highway bridges in theHyogo-ken nanbu earthquake. The Hyogo-kennanbu earthquake was the first earthquakewhich developed destructive damage in anurban area since the 1948 Fukui Earthquake.Because it had been considered that suchdestructive damage could be prevented due tothe progress of construction technology inrecent years, it provided a large impact on theearthquake disaster prevention measures invarious fields. The "Part V Seismic Design" ofthe "Design Specifications of HighwayBridges" (Japan Road Association) was totallyrevised in 1996, and the design proceduremoved from the traditional Seismic CoefficientMethod to the Ductility Design Method. Therevision was so comprehensive that the pastrevisions in the last 30 years look minor.

Major point of the revision was theintroduction of explicit two-level seismicdesign consisting of the Seismic CoefficientMethod and the Ductility Design Method.Because the Type-I and the Type-Il groundmotions are considered in the Ductility DesignMethod, three design seismic forces are totallyused in design. Seismic performance for eachdesign force was clearly stated in theSpecifications.

The fact that lack of near-filed strong

Table 2.9 Modification Coefficient for Clearance CB

c. TIT, CB

o~~ TlTl < 0.1 1

0.1 ~ ~ TIT, < 0.8 -V2

0.8 ~ ~ T(I\ ~ 1.0 1

2.14 CONCLUDING REMARKSThe preceding pages presented an outline

2-20

motion records prevented to seriously evaluatethe validity of recent seismic design codes is


important. The Hyogo-ken nanbu earthquakerevealed that history of strong motion recordingis very short, and that no near-field recordshave yet been measured by an earthquake withmagnitude on the order of 8. It is thereforeessential to have enough redundancy andductility in a total bridge system. It is hopedthat the revised Seismic Design Specificationsof Highway Bridges contributes to enhanceseismic safety of highway bridges.

ACKNOWLEDGMENTS

Drafting of the revised version of the "Part VSeismic Design" of the "Design Specificationsof Highway Bridges" was conducted at the"Special Sub-committee for SeismicCountermeasures for Highway Bridges" andwas approved by the Bridge Committee, JapanRoad Association. The first and other authorsof this paper served as chairman andexecutive members in the SpecialSub-committee. The authors thank ail membersof the Special Sub-Committee and the BridgeCommittee.

REFERENCES1) Japan Road Association Design

Specifications of Highway Bridges, Part ICommon Part, Part II Steel Bridges, Partill Concrete Bridges, Part IV Foundations,and Part V Seismic Design, 1996

2) Kawashima, K.: Impact of Hanshin/AwajiEarthquake on Seismic Design andSeismic Strengthening of HighwayBridges, Report No. TIT/EERG 95-2,Tokyo Institute of Technology., 1995

3) Ministry of Construction: Report on theDamage of Highway Bridges by theHyogo-ken Nanbu Earthquake, Committeefor Investigation on the Damage ofHighway Bridges Caused by theHyogo-ken Nanbu Earthquake, 1995

4) Ministry of Construction: GuideSpecifications for Reconstruction andRepair of Highway Bridges WhichSuffered Damage due to the Hyogo-ken

2-21

Nanbu Earthquake, 1995

3. SEISMIC DESIGN FOR RAILWAY STRUCTURES

RAILWAY TECHNICAL RESEARCH INSTITUTE, JAPAN

3.1 Basic Principles of Seismic Design for Railway Structures

3.2 Setting of Design Earthquake Motions

3.2.1 Setting of Earthquake Motions for Bedrock

3.2.2 Setting of Design Earthquake Motions on the Ground Surface

3.3 Seismic Performance of Structures

3.3.1 Setting of Seismic Performance Levels for Structures

3.3.2 Consideration on the Damage Levels of Member, the Stability Levels of

Foundation as well as their Limit Values

3.4 Concept ofImportance Degree of Structure

3- 1

3- 3

3- 3

3-11

3-13

3-13

3-14

3-17

3.5 Evaluation of Surface Ground and Calculation of Displacement and Stress ofStructure3-17

3.5.1 Evaluation of Surface Ground 3-17

3.5.2 Calculation of Responses of Structures 3-24

3.6 Safety (Seismic Performance) Checking of Structures 3-25

3.6.1 Checking Damage Levels of Members 3-27

3.6.2 Checking Stability Levels of Foundation 3-27

3.6.3 An Example of Safety Checking of Pile Foundation 3-27

3.7 Conclusions 3-29


3. SEISMIC DESIGN FOR RAILWAY STRUCTURESRAILWAY TECHNICAL RESEARCH INSTITUTE, JAPAN

3.1 Basic Principles of Seismic Design for

Railway Structures

A new code, "Seismic Design Code for

Railway Structures" (in Japanese), drawn up by

Railway Technical Research Institute, has been

published recently, which reflects the recent

advances in earthquake engineering'{ In the code

some new thought for seismic design have been

adopted by drawing the lesson of the Hyogoken

Nanbu Earthquake of January 17, 1995 that

caused the devastating damage including the

large-scale cave-in of many railway structures. In

order to introduce a methodology for the seismic

design that can effectively prevent reappearance

of the kind of damage happened in the Hyogoken

Nanbu Earthquake, elucidation of the damage

mechanism has been conducted. As the results,

the following causes of the damage are inferred

based on the damage reconnaissance and

analysis".

CDMany of the structures damaged possessed the

seismic capacity that was designed by only

considering a horizontal design seismic

coefficient of 0.2. However, the acceleration

level of the Hyogoken-Nanbu Earthquake was

far over such a design level and caused the

large damage.

®Viaducts of the Shinkansen that suffered

serious damage including the collapsing of

structures, were originally designed to be less

safety against shear loads than bending loads.

This imbalance aggravated the damage degree

of the structures. This was partly due to the

fact that allowable stress against shear force

was set larger in the design code of those days.

@Some situation of the damage showed a great

gap in the damage degree between two

adjoining viaducts, where one side collapsed

3-1

totally and the other side with only cracks in

columns. This situation with different damage

pattern might be mainly due to the difference in

dynamic behavior of the surface ground, which

was inferred through the dynamic analysis by

considering both the properties of structures

and ground.

@As to the damage of cut and cover tunnel, both

bending and shear stresses occurred in columns,

but since the shear strength was lower than that

of bending which is same as the case of

viaducts, the shear failure occurred and caused

the collapse under the weight of overburden.

The facts above indicate the following

procedures are important to seismic design.

CDTaking inland earthquakes into account

®Evaluating the safety of members by

considering the failure modes of structures

@The necessary to use dynamic analysis

methods and consider the dynamic behavior of

surface ground in response analysis of

structures.

Moreover, the level of design earthquake

motion has become dramatically large because of

consideration of the inland earthquakes.

Generally the return period of the intense

earthquake may be several hundred years long.'

Therefore, it is reasonable to abandon the elastic

design method and adopt the performance-based

design method in which the seismic performance

of structures is evaluated and the damage of

structure is allowable in some extend, but never

the collapse.

Seismic design of a railway structure should

therefore be carried out according to the

following procedures. Firstly, from the

viewpoint of damage control, the degree of

damage to a structure (seismic performance)

should be identified. Secondly, the responses of


the surface ground are analyzed by inputting the

design earthquake motion in the base ground.

Thirdly, the response waves of the surface

ground are inputted to the structure and the

responses of the structure are analyzed. Finally,

basing on the obtained responses of the structure

the seismic performance can be checked.

There are two types of design earthquake

motion are determined in this code. One is the

so-call L1 earthquake motion, which has a

recurrence probability of a few times during the

service life of the structure. .The other is L2

earthquake motion with high intensity, which is

caused by a near-land-large-scale interplate

earthquake or an inland earthquake near to the

structure. Comparing with Ll earthquake, the

occurrence probability of L2 earthquake is low.

For the earthquake motions, by considering the

damage of members and stability of the

foundations, the seismic performance of a

structure is set to 3 grades corresponding to the

presumed levels of repair or reinforcement that

may be required following an intense earthquake.

In the seismic design, responses of a structure

resulting from an earthquake should be made to

satisfy the seismic performance objective.

Which performance the structure should be

endowed with basically depends on the

importance of the structure.

As the reasons described above, in order to

check the seismic performance properly, a

dynamic analysis method for calculating the

responses of a structure is generally adopted in

seismic design. However, some times a static

analysis method is also used depending on the

type of structure. The procedure of seismic

design for bridges or viaducts based on the

approaches above is shown in Fig.3.1.1.1.

As what indicated in the figure, there are two

types of approaches can be used for the seismic

design. One is the simplified method (nonlinear

spectrum method) that can be easily applied for

the calculation of the responses of a structure by

i) selecting the soil profile type based on site

geological conditions; ii) using the demand yield

strength spectrum that is calculated with the

earthquake motion corresponding to the soil

profile type selected. The other is the detailed

Setting inputearthquake motion

Evaluation ofsurface ground

Calculation ofresponses of structures

Examinationof seismic performance

Selection of Ll , L2 earthquake motions(Spectrum I, Spectrum II)

Selection of earthquake motionsaccording to Soil Profile Type

Simplifieddynamic analysis

(Nonlinear spectrum method)

Members : Damage LevelFoundation; Stability Level

Fig.3.1.1.1 Procedure of seismic design for bridges or viaducts

3-2


method (time-history dynamic analysis method)

with which the time history of responses of the

ground and structure can be analyzed detailed.

For a common structure, the nonlinear

spectrum method is suitable. However if a

structure can not be modeled as a system with

single degree of freedom, as described later, the

detailed analysis method should be applied to.

In the following pages, major procedures for

the seismic design, such as the setting of design

earthquake motions, the analysis of

displacements and stresses of structures, and the

checking of structural safety are described.

3.2 Setting of Design Earthquake Motions

3.2.1 Setting of Earthquake Motions for

Bedrock

(1)Types and Determination of Design

Spectra

As what described previously, in order to

consider the effects of surface ground to the

responses of a structure, either LIar L2

earthquake motion is set on the surface of

bedrock.

Ll earthquake motion has about the same level

as the acceleration spectrum corresponding to the

high quality ground that used to be adopted in

the allowable stress design. The maximum value

of the response acceleration is 250 gal

corresponding to the damping coefficient of 5%.

L2 earthquake motion is classified into the

following 3 types.

CD SpectrumI acceleration spectrum

corresponding to the near-land interplate

earthquakes of magnitude 8.0 and epicenter

distance of 30 to 40 kilometers.

In addition, the inland active fault, which will

cause an earthquake of magnitude less than 6.5, is

difficult to be found since its size is not big

enough to reach the ground surface. According to

the historical earthquakes, this type of earthquake

3-3

may happen in most areas of Japan. Consequently,

the motion due to this type of earthquake is also

covered by Spectrum I, therefore this spectrum is

regarded as the minimum earthquake motion to be

verified in the seismic design.

® SpectrumII : acceleration spectrum based

on the statistic analysis of the earthquake data

recorded in the past inland earthquakes caused

by active faults.

@ Spectrumill: also representing the

motions caused by active inland faults, but based

on the analysis of the active faults, if such a

model of active fault is available.

motion from the 3 types above is a difficult, but

important task in the seismic design, because the

presumed earthquake may be affected by a great

amount of uncertainty.

It is desirable to determine the design

earthquake motion for a specific site according to

the risk factors such as the return period of

earthquake from certain seismic faults. However,

the return period of earthquake related to an

inland active fault is not accurate enough at

present, when compared with the service life of

structure. Therefore, an extreme event associated

with an inland active fault should be taken into

account, unless it is evident that the fault will not

move during the life of structure.

To determine the design earthquake motion of a

site, the geological and seismological information

on inland active faults, historical activities of

earthquakes around the site and interplate

earthquakes near land must be analyzed

carefully'). A general flowchart is given in

Fig.3 .2.1.1.

There are a number of ways to define the design

earthquake motion. The design earthquake motion

is defined below by the response spectra of

acceleration on a free surface of bedrock, the

shear wave speed of which is over 400m/s. The

choice of bedrock is to avoid the influences from


local effects of specific site on the ground motion,

such as the amplification due to the soft surface

soil and irregular topography of ground. The

influence due to geological conditions is very

remarkable, as recognized in seismic records, and

can be evaluated by calculating the responses of

surface soil using a proper numerical model of

surface ground with the design earthquake motion

as the incident motion. A corresponding artificial

seismic wave can be generated by adjusting

Fourier amplitudes of the wave according to the

objective response spectra of acceleration and

modeling the phases to reflect the non-stationary

property of earthquake motions.

Which spectrum should be used as the design

earthquake motion depends on the results of

investigation of inland active faults. There could

be three possibilities shown following from the

investigation (Fig.3.2.1.1).

The first (the left route in Fig.3.2.I.1), if there is

no active fault near the site, the earthquake motion

of Spectrum I is to be used as design earthquake

TNo

Analysis withsource model?

Yes,

Doubtful

No

Determine localseismic risk factor

,Spectrum I modified

by risk factor

Computation of groundmotions

~-----'

I,Determination of

spectrum ill

I,Compared withodified spectrum

,Artificial wave

I,c?

Speetrum Ilattenuated with

distance

Determine localseismic risk factor

,Spectrum TI

modified by riskfactor

Fig.3.2.1.1 General flowchart to determine the design earthquake motion

3-4


motion after modified by the risk factor of the

area.

The second (the middle route in Fig.3.2.1.1),

there are cases where one or more active faults

existing near the site. When the parameters of

seismic source for the faults can be properly

decided, the design earthquake motion can be

determined by the fault analysis with source

model (Spectrum III). Otherwise, the earthquake

motion of Spectrum IT attenuated according to the

distance between the fault and site, will be used as

the design earthquake motion. Because the power

of the motion decreases as the distance between

attenuated results of Spectrum IT and III should be

compared with that of Spectrum I modified by the

risk factor of the area, then the larger one will be

taken as the design earthquake motion.

The third (the right route in Fig.3.2.1.1), there

are sites where the existence of active fault is very

doubtful and difficult to confirm due to very deep

sedimentary deposit, or there exists a complex

tectonic structure beneath the site, such as the

Kanto area in Japan where three plates encounter

with each other. Hence, the design earthquake

motion is Spectrum IT modified by the risk factor

of the area.

(2)Near-Source Earthquake Motions Induced

by Inland Active Faults

There are still many problems to be solved

when using a seismic source model of fault to

predict the earthquake ground motion at a site for

the purpose of seismic design, such as the

distribution of the asperity on the fault plane, the

start point of rupture, etc. To consider these

uncertainties of it, it is effective to evaluate the

earthquake motion near inland fault from

statistical analyses of near-source strong seismic

records observed in recent years. Below

summarized is a method to determine Spectrum IT

based on strong seismic records.

1) Seismic records

Table 3.2.1.1 shows the list of records observed

in recent earthquakes in the United States and

Japan, Hyogoken-Nanbu (1995,M7.2), Coyote

Lake (1979, MS.9), Loma Prieta (1989,M7.1),

Landers (1992, M7.5) and Northridge (1994,

Table 3.2.1.1 Near-source seismic records from recent earthquakes

Max. Ace. (gal) ] "StU

r!:::LL aOl .8Ol

Ol '-'_ c

-"" Ol o Ol Ol Ol 0

'" ""0""0 c.. c> o > :.= Soil condition

No::::>

3 3 .>, c; c Ol ro0- Nameof seismic record '5> .c ro ro -2:£; NS EW ~ c -05 05 ""OOl at the positionof seismometerffi ....J 0

C._0 C en

....J Ol""O ::::>.0W Cij 05 eo

> Ol o'5 '"0- 0UJ U

679.8 302.6 135.208 11.64 3.24 GL-83 Vs=450 (m/s)

2 86.0 109.3 134.783 32.75 27.08 GL·100 Vs=460 (m/s)

3 293.9 319.8 135.442 34.57 24.65 GL-97.0 Vs=455 (m/s)

4 Hyogoken- 272.0 306.5 135.240 14.99 6.90 GL·9.5 0.5m(240m/s) layeroverVs=590 (m/s)

5 Nanbu 185.3 200.4 135.427 38.03 25.03 GL-30 Vs=780m/s

6 445.9 425.3 135.296 20.00 12.38 GL-33 Layerof N=18 aboveGL-45

7 683.6 600.9 135.344 29.93 16.88 GLO.O Nover63, 1.5m surfacelayerwithN=5

8 510.7 584.2 135.250 16.52 7.53 GLO.O Vs=300m/s, 4msurface layerVs=200m/s

9 Coyote Lake 314.6 408.8 121.484 1.0 GLO.O Rock

10 433.1 401.5 122.06 18.01 12.19 GLO.O LimestoneLoma Prieta

11 426.6 433.6 121.572 26.56 12.21 GLO.O Franciscan Sandstone

12 Landers 268.3 278.4 116.314 16.90 10.79 GLO.O Shallow alluvium overgranite bedrock

13Northridge

GLO.O Thin alluvium oversiltstone

::;] GLO.O IRock

3-5


5430.40.30.210

0.1

4000

2000

1000800

600---.'"2 400eo<::»

::::0.~ 200'-<1)

a:5<:)o 100~ 80

60

40

20

0.5 0.6 0.7 0.80.9 1 Period (sec) 2

Fig.3.2.1.2 Acceleration response spectra of observed records at near-source area of inland earthquakes

M6.7). The records are chosen to satisfy the

following requirements.

CDThe soil condition at the station of seismometer

meets the condition of the aforementioned

bedrock.

®The maximum acceleration is greater than

lOOgal.

@The Closest Distance to Fault is less than 30km.

The list shows that the records of Hyogoken

Nanbu Earthquake are all within the ground.

Theoretically, deconvolution shall be carried out

to separate the incident wave from the record. The

original records are used here instead, because it

is difficult to get a result that is reasonably closer

to incident wave than original record, as there are

a number of unsolved problems in the

deconvolution analysis for strong ground motion.

Besides, the influence of the surface soil would

not be too strong since the shear wave speed of

soil at all sites is higher than 450m/s anywhere,

except at the Great Bridge of East Kobe.

The acceleration response spectra of the

selected records are illustrated in Fig.3.2.1.2. It

can be found that the response accelerations vary

from 200(gal) to 3000(gal) in the range of short

period and from tens of gals to lOOO(gal) in the

range of long period. As the soil conditions at the

observation stations have been carefully chosen,

this wide variation may be attributed to the

following.

CDDifferencein the mechanism of seismic sources

®Difference in the propagation of seismic waves

@Influence of irregular topography

The influence of irregular topography can be

avoided only by selecting records according to the

geological condition if available. Through a

careful investigation, it is found that the records at

Tarzana, Northridge earthquake (1994, M6.7),

3-6


N INx-2 ="d 2x:-2 "d2eq i..J I 1 .t... I

j=l j=l

New Kobe substation and Takaratuka, Hyogoken

Nanbu earthquake (1995,M7.2), are influenced

strongly by special topography4),S), so that these

records are excluded from the statistical analyses.

As to the influence of the propagation of

seismic waves, the profiles in deep ground in the

range of several kilometers as well as the Q factor

(quality factor) are considered to be very

important, but they are out of the scope of this

study. However, a number of attenuation

functions of ground motion have been proposed,

in which influences on propagation are all

considered in an average sense. By using the

recorded earthquake motions to a same distance

from the seismic source so that the variation of

ground motion due to propagation can be

minimized. The rest variation of ground motion

in statistics is attributed to the properties of the

seismic source or other unclear reasons.

2) Compensation by attenuation function

Among the attenuation functions proposed, the

measurement of the distance between the site and

the seismic source is very important to decide the

near-source strong ground .motion, where the

extent of fault plane must be considered properly.

To satisfy the above requirement, the Closest

Distance to Fault (CDF) has been widely used

recently. The following is an attenuation function

of response spectra of ground motion based on

CDF which is proposed by Fukushima6).

logS(T) =aj(T)M~ - az(T)Mw + b(T)· R

-log(R +O.025xlOo.4zMw) + [,cj(T)lj

(3.2.1.1)

in which M w , R and T are the moment magnitude,

the Closest Distance to Fault and the period,

respectively; al, az and b are coefficients of

regression; Cj is the coefficients related to site

properties.

3-7

On the other hand, Ohno et al.7) proposed

another type of attenuation relation based on

Equivalent Hypocentral Distance, this is

determined by the energy radiated from the finite

fault plane.

logS(T) = a(T)Mw -logXeq - b(T)Xeq + c(T) +&(T)

(3.2.1.2)

(3.2.1.3)

where x ; denotes the Equivalent Hypocentral

Distance; N, Xi, and d, are the number of small

site and the center of the area i, and the seismic

moment on the area i, respectively.

The Closest Distance to Fault and Equivalent

Hypocentral Distance given in Table 1 for every

site of record are calculated according to the fault

models published by USGS for earthquakes in

USA and by Irekura for Hyogoken-Nanbu

Earthquake, respectively.

There is an important phenomenon for the

ground motion in near-source area, in that it tends

to saturate as the site is getting close to fault

presumably for the following reasons. Firstly,

most of active fault planes are nearly vertical to

the ground surface. Secondly, the thickness of the

crust of the earth is from 15 to 20km. In

consequence, the size of the fault in the horizontal

direction will increase as the scale of the

earthquake gets larger, so that the affected area

becomes larger too. However, the intensity of

ground motion at the near-source area will not

increase because the energy does not concentrate

but widely spreads on the whole plane of the fault.

Since we need to infer the ground motion right

above the fault, we can omit the influence of the

magnitude while taking into account only the

distance between the site and the fault,

compensating for the observed records by the

aforementioned attenuation relation.


Fig.3.2.1.3 Observed spectra compensated

with the attenuation relation of

Equivalent Hypocentral Distance

4 5.4 0.5 o.e0.70.80.91 Period (sec) 2

2D

40

100.1

1000800

600

"lO t:<_..... ·"-_"";'''''';;;Mm",-H-.p..:;c~H~~~~~-H~ _=1.]: 200 ---=~~~ =~:::::;:3100 _1oI._a.-Jo

~ 80 =:-='~':If]l-.~.~"')j_··_$o.:r._...._1o.r.... ,..1_ ••__e-UCl:ca ..__ -to_C_UOC"I__~l_.c.u;w

=:=~::-c..&)QH-+-t-++++----1----+":-H_1.'.t_NS__,.. .....w

_","l,I.~

After all fault models are examined, the

Equivalent Hypocentral Distance of destination is

taken as 12lan in this study. With the attenuation

relation using the value of Equivalent Hypocentral

Distance, all the acceleration response spectra of

observed records are then compensated, the

results are shown in Fig.3.2.1.3.

Because, overall, the compensation from the

attenuation function gives a ground motion closer

to the fault than original records, all spectra

become larger. The upper limit is about 2000gal,

except those of SGK EW97 and Gilroy#l

Gravilan Coll.EW records. As expected, the

Fig.3.2.1.4 Comparison of the statistical results

based on the seismic records in USA

and Kobe, respectively

Fig.3.2.1.5 Comparison of the statistical

results using Closest Distance to Fault

and equivalent hypocentral distance

attenuation relations

100.1

4 5

4 50.2 0.3 0.4 0.5 0.6 0.7 OBO.g 1 Pedodfsec) 2

0.2 0.3 0.4 0.5 0.6 0.70.80.91

.' 1~'~r~""

i---.~_.~I"'- '" f·...... .... " .. ,,,·..•......·i·.

...~'., ~" I

,.~~

~ ...--... mean(Kobe)", ·1·:··.. ··.~..Iv-,

-_.. 90% unsurpass (Kobe) ", "~ \

=......... mean(US) >----- llnsurpass (US)

-,- -mean I

90% Unsur ss

I~ 1'--.. rt- IV' J--..... i-

---I ~ '"<,

1'----. 1"'-r-. "\

1

2D

20

100.1

100

80

60

2000

40

1000800

600

400

200

4000

2DOO

1000BOO

BOO

40

4000

mean value becomes smaller for all periods.

When compared with those in the short period,

the improvement in the long period is slight, to

imply the existence of dominating effects from

the seismic source and the structure in deep

ground.

Fig.3.2.1.4 compares the statistical results based

on the seismic records of USA and Kobe. They

satisfactorily agree with each other for the period

up to 1.0 second. For the period longer than 1.0

second, the records at Kobe give larger response

spectra This difference would be a major cause

of larger deviation of total statistical results in the

long period range. Meanwhile, it can also be

found that the statistical results become smoother

as the number of records increases.

The attenuation function based on CDF is also

used, where the distance of destination is taken as

2lan. The point of 2km from fault is the place

right above it, because little portion of energy will

be radiated from the range within 2km from the

ground surface, even though the fault reaches and

appears on the ground surface. There is not much

difference between the mean of response spectra

and that based on the Equivalent Hypocentral

Distance, but the values of 90% unsurpassed

probability show a little difference (Fig.3.2.1.5).

This illustrates that the statistical result of ground

np.v1:ltlcm of crrrrrmri monon frorn thp. staristir-al-_ ..--...-- --- 0---- ~- ---- ---- -- ~---~---

3-8


Fig.3.2.1.6 Response spectra of acceleration

for design earthquake motion straightly

above an inland fault (Spectrum II)

log plot to define the response spectra of

acceleration for the design earthquake motion

(Fig.3.2.1.6) called Spectrum II. Its values

corresponding to the ranges of period are shown

below.

CDllOOgal at O.ls in period

®1700gal between 0.2s and 0.7s in period

@154galat 5.0s in period.

This spectrum express the motion just above a

fault straightly. Therefore, it's values can be

reduced by the attenuation relationship according

to the distance between the seismic fault and the

site. Here Formula (3.2.1.2) based on the

Equivalent Hypocentral Distance is recommended.

(3)Earthquake Motions due to Near-Land

Interplate Earthquakes

In the codes of seismic design used before the

Hyogoken-Nanbu earthquake, the seismic motion

of interplate earthquake was provided. The values

of the response spectra of acceleration for the

design were about 1000gal on a standard ground

surface. In order to treat the earthquakes due to

inland faults and interplates on a same basis, the

same methods for statistical analysis and

compensation carried out so far are applied to the

seismic records of interplate earthquakes. A brief

outline about the determination of Spectrum. I is

summarized bellow.

0,2 0.3 0.4 0.5 0.6 0.70.80.9 1 Penod (sec) 2

~yI

I

-,

I-,

IDamping ratio h=5% I

II5

'000.1

~ sao

$500

".,g 400

~'E 300

:<

2"'0

,,'"900800700

motion right above the fault is almost independent

of attenuation relation of the Closest Distance to

Fault or the Equivalent Hypocentral Distance.

3) Spectrum for earthquake motion straightly

above the inland fault

In view of the limited number of records

adopted at present as well as unknown properties

of earthquakes in the future, it is wise and

reasonable to determine the design seismic motion

according to a certain unsurpassed value, rather

than by taking the envelope of the maximum

values.

To what degree the unsurpassed probability

should be taken is very important but difficult to

determine. It usually depends on a subjective

judgment. For railway structures, the following

considerations are necessary.

CDRailways are means of mass transportation

directly related to the safety of passengers.

@ A failure at one point of a railway system will

affect the whole route, and it is very costly and

impossible to have a bypass for the same

railway.

@The seismic records used are limited possibly

with unknown factors.

In the light of above considerations, a high

unsurpassed probability is strongly expected, but

the value 90% is believed to be acceptable and

adequate when the accuracy of the whole process

of seismic design is taken into account.

It is not difficult to get the value of a certain

unsurpassed probability if we assume that the

response spectra at the given period is normally

distributed. The 90% unsurpassed value is given

in thick dot line in Fig.3 .2.1.3. Due to the

influence of the records at SGK (Hyogoken

Nanbu earthquake) and Gilroy Gavilan Coll.

(Lama Prieta earthquake), the apparent value near

0.3s in period is over 2000gal, which may be

attributed to some local effects of two sites.

Therefore, we use three straight lines on the log-

3-9


Table 3.2.1.2 Seismic records from recent interplate earthquakes in Japan<; = ....

" "o .b '" <5= ~

~,,~

.8 §g 2;.0,,-... ~~ " a

<.l~ '" Direction ofEarthquake No Recorded site Latitude Longitude <; 2 ::c~ § 2 .~

.b~ o ."§ 0 '- recordsc~ c

" c '" Cl0

o .£ U;

:~0 '" .-

~0, ;>-0c-, .::;

::r: C' 0 0~ U P-.

Tokacbi-Oki (May 16,1968) 1 Hacbinohe 40.55 . 141.483 179.4 130 88.6 GL NS,EW

Off NemnroPen. (June 17,1973) 2 Otanoshike Brg. 43.0083 144.271 136.9 163.7 109.7 GL LG

3 Kaihoku Brg. 38.445 141.313 81.6 70.2 56.5 GL LG,TROff Miyagi Pref. (June 12,1978)

86.8 71.64 Ofunato- Bochi 39.00 141.733 101.7 GL N41W,E41N

W off N Tohoku (May 26,1983) 5 Kamitorizawi Brg. 42.1014 140.563 231 190.8 144.5 GL LG,TR

6 Urakawa 42.158 142.781 151.6 174.4 149.1 GL NS,EW

7 Hanasaki Port 43.2800 145.589 109.4 156.4 131.3 GL N20E,E20S

8 Tokachi Port 42.2889 143.324 106.5 141.7 121.8 GL NS,EWKusiro-Oki (Jan. 15, 1993)

142.27929 HirooBrg. 143.319 107.5 142.4 122.4 GL LG,TR

10 Otanosbike Brg. 143.0083 144.271 19.8 105.2 100.1 GL LG,TR

11 Chiyocla Brg. 42.9197 143.389 81.5 123.3 108.2 GL LG,TR

12 Muroran Port 42.3167 140.967 153.3 149.0 129.3 GL NS,EWHokkaido Nansei-Oki (July 12,1993)

13 Kamitorizawi Brg. 42.1014 140.563 124.6 120.1 91.4 GL LG,TR

Hokkaido Tohoku-Oki (Oct 4,1994) 14 Hanasaki Port 43.2800 145.589 168.4 123 58.5 GL N20E,E20S

0.1

Distance (km)

~Hypocentrai distancer---

SiI EquivalentHypocentral distance

o Closest distance to fault r-----

I~ n I ~ ~ ~

II 11m ,~ .~~ II~~ n~111 ~ ~

16

14

12

Fig.3.2.1.7 Interplate earthquakes occurred in

Japan recently

Fig.3.2.1.8 Distribution of distance between

seismometer and seismic source

1) Seismic records of interplate earthquakes

The recent interplate earthquakes occurred near

Japan are shown in Fig.3.2.1.7, from which over a

hundred records with the maximum acceleration

larger than 100gal have been collected. The

distances between the site of seismometer and the

source are mostly from 100 to 200km

(Fig.3.2.1.8).

From these data, 27 records have been chosen,

according to the following requirements, their

detail information is given in Table 3.2.1.2.

(DBoth Equivalent Hypocentral Distance and

Closest Faults Distance are within 200

kilometers.

®The soil condition of the observation station is

good.

@There is no unnatural peak existing in the

Fourier spectra of the records.

The acceleration response spectra of those

records are shown in Fig.3.2.1.9.

3-10


2) Spectrum compensated for interplate

earthquakes

The attenuation relations used for inland

earthquakes are also used here. As we have to

take the scale of earthquake into consideration

because the earthquake motion to be inferred is at

the site a little far from the seismic source, the

motion becomes very sensitive to the scale of

earthquake.

After investigating the effect of scale of

earthquake with the different attenuation

functions .based on the Equivalent Hypocentral

Distance as well as the Closest Distance to Fault

respectively, the final compensated result is

obtained. Through analysis of the characteristics

of this compensated spectrum, the Spectrum I is

defined by two straight lines in log-log plot as

-S-ZS2NS

~ktro~~~~~ =~~EW_It!o~n~s

_IXH_W_OO-l_W

~,,=----"~"-+--+--1 =~~~~

+--f'-,:-"'f-.j =;:~_P_61Il:tS_00'_00....

~l&"+'rl--i =~:~;:=~:-oo-6U>_oo-ou_OO_IIHA-OO-11lUl-OQ-11HA._OO_11Wl_CXi_lJlU..-oo-1.ltlB-!'-ss.,..s_1"_llol'lW_~'l"':lltS

-J,(-I4f38W

aa (1.<1 1).5 G.6 U-1l>.8'O.9 1 Period (sec) 2

shown in Fig.3.2.1.10, which possesses the

following values.

(1) 1100gal between O.ls and LOs in period

(2) 154gal at 5.0s in period

For comparison, Spectrum IT is also plotted in

Fig.3.2.1.10. In order to make the design

simplified, the values of Spectra I are defined as

the same as the Spectrum IT for the period longer

than 1.0 second. When compared with the

original records in, Fig.3 .2.1.9, the design spectra

are larger in the period longer than 0.2 seconds.

In general, the vibration of high frequency

decreases much quicker due to the damping in

structure and soil as ,:veil as the displacement

associated with it is small too. Therefore, the

design, spectra defined here does not

underestimate the actual ground motion for the

seismic design.

Besides, the level of Spectrum I, even when

multiplied by the smallest risk factor, can cover

the ground motion due to an earthquake with the

magnitude less than 6.5 which may occur inland

without making its fault reach the ground surface.

This can be easily verified through the attenuation

function given above, where the depth of a fault

center is assumed as 10km from the ground

surface".

Fig.3.2.1.9 Acceleration response spectra of

observed records of interplate

earthquake near Japan

3.2.2 Setting of Design Earthquake Motions

on the Ground Surface

Fig.3.2.1.10 Comparison of the spectra

between the interplate and inland

While calculating responses of a structure to

earthquake motions, the structure can be

modeled as an overall system including the

surface ground where the foundation is

embedded, then inputting the previously

mentioned design earthquake motion into the

bedrock, and using a dynamic analysis method to

perform a nonlinear analysis which can take the

effect of soil-foundation-superstructure

interaction into account. .This kind of procedure,

however, at this time is considered overly

4 ,

I.......... "1"" .....

.... ... ... -

··j··l,

II

Spectrum I "' I

I I·...... Spectrum If <, II r-,

I I i I I'r-,

, ,, I I I II I I

10090eo1060

"0.1

l~

~:3500.9 .400

] 300

1i.:: 200

'000

3-11

3000

2000

r--,

catlJ)"-";:::

.9 1000....., 900('j..... 800(l)

Q) 700o

600()

<C 500(l)UJ;::: 40000..UJ(l) 300~

200


G2l.--/ I ""'- 2"3

~v-;>~, ,

~q5" --------- -.- -.- K~l~~------

- ....... .j-.......... I~-....... I"" _._.u.. ...... ..~z~:.~.~ .. 'G6..' .. r-,~

- .-' .».' -- r-, <, -, " ---', , ,--" --~ -- .' -.. ...... <, -, ,"", ,

-, ,-- .' - <, -, !'-.-'-'>-, ,.- .' .- -, , ,-. ... - .., .... .G7 -, <, ~" '

,.' " '

. ,.. -- "GO"" ".~'-, ",.. , . , -, ,.... -- --. " .--

~~ ",.. J" Gl,. ....'\.........---

~~.-

""~,

"'"'-. ""-~~

,~~I

~c:;0.1 0.2 0.3 0.4 0.5 0.6 0.7 0.8 0.9 1 2 4 5

Period (sec)

Fig.3.2.2.1 Design response spectra of acceleration on ground surface for Spectrum II (damping

coefficient of 5%)

complicated and impractical for general use. As

a general rule, in order to simplify the design

procedure, the foundation of a structure will be

replaced by supporting springs and the

superstructure modeled as a multiple mass

system. In this case, the earthquake motions on

the ground surface are needed, which can be

calculated from dynamic analysis of the ground.

But, in reality, there are difficulties in this

dynamic analysis of surface ground such as

setting of relationships between the strain and

shear modulus of ground, damping coefficient of

soil and so on. To overcome such difficulties,

design earthquake motions on ground surface

corresponding to various types of soil profile

were investigate in an amount of parametric

studies. As the results, the acceleration response

spectra on ground surface due to Spectrum I and

Spectrum II are determined. Fig.3.2.2.1 gives the

design response spectra of acceleration on

ground surface for Spectrum II, which are

corresponding to the soil profile types from GO

3-12

Table 3.2.2.1 Soil Profile Types

Soil Soil ProfileProfile Period (sec) Name/GenericType Description

GO - Hard Rock

Gl - Bedrock

G2 -0.25 Diluvium

G3 0.25-0.5 Dense Soil

G4 0.5-0.75 Dense to Soft Soil

G5 0.75-1.0 Soft Soil

G6 1.0-1.5 Very Soft.Soil

G7 1.5- Extremely Soft Soil

to G7. Moreover, the soil profile types indicated

in the figure are categorized based on the natural

periods of ground that are calculated with the

velocities of elastic shear wave in surface ground.

The relations between the soil profile types and

the natural periods of ground are summarized in

Table 3.2.2.1.


In summary, there are 8 types of soil profile

used in this code. With respect to each soil

profile, the design response spectra of

acceleration on ground surface are determined

corresponding to the L1 earthquake motion,

Spectrum I and Spectrum II of L2 earthquake

motion.

3.3 Seismic Performance of Structures

3.3.1 Setting of Seismic Performance Levels

for Structures

Corresponding to the presumed levels of repair

and reinforcement of structures that may be

required after an intense earthquake, the seismic

performance can be categorized into 3 levels as

follows.

CDSeismic Performance I (SPI): capability of

maintaining the original functions without any

repair and no excessive displacement

occurring during an earthquake

@Seismic Performance II (SPIT): capability of

making quick recovery of the original

functions with repairs after an earthquake

@Seismic Performance ill (SPill): capability of

keeping the overall structure in place without

collapse during an earthquake

These performance levels are mainly defined

by the ease degree of recovery of the structures

after an earthquake. Therefore, the relationship

between the levels of earthquake motions and

seismic performances has been established as

follows.

For L1 earthquakes, the structural seismic

structures designed.

For L2 earthquakes, SPII should be satisfied by

the structures with greater importance, and SP ill

by other structures.

Furthermore, the seismic performance levels

are also connected with the state of damage of

member as well as the stability of foundation

Seismic Performance I (SPI)Capability of maintaining the

original functions without anyrepair and no excessivedisplacement occurring during anearthquake

Seismic Performance II (SPII)Capability of making quick

recovery of the original functionswith repairs after an earthquake

Seismic Performance III (SPill)Capability of keeping the

overall structure in place withoutcollapse during an earthquake

Damage Level 1: no damage

Damage Level 2: damage that may require repairdepending on situation

Damage Level 3: damage requiring repair

Damage Level 4: damage requiring repair, andreplacement of members depending on situation

Stability Levell: no damage (loading smaller thanbearing capacity)

Stability Level 2: damage requiring repair depending onsituation

Stability Level 3: damage requmng repair, andcorrection of structure depending on situation

Fig.3.3.1.1 Relationship among seismic performance levels, damage levels of member and

stability levels of foundation (bridges and viaducts)

3-13"


Reinforcing bar yieldingin axial direction

Cracks occurringEnvelop curveof test results

Maintaining yield load

Skelton curvefor analysis

Deformation

Fig.3.3.2:1 Relationship of lateral load-deformation relationship for reinforced concrete

member, with a generai ievei of compressive axiai force

which are constituted in the overall structure.

Since the damage level of member and the

stability level of foundation will influence the

structural seismic performance level much, how

to determine them properly is important. ill this

code, the damage level for each member which

composes a structure is set properly by

considering the role played by the member for

the overall structure. ill regard to the stability of

foundation, as it has a big impact on

displacement of a structure, it should be

determined by considering the bearing capacity

or the deformation of the foundation involved.

Fig.3.3.1.1 shows the relationships among

seismic performance levels required for bridges

and viaducts, the damage levels of member, and

the stability levels of foundation.

3.3.2 Consideration on the Damage Levels of

Member, the Stability levels of

Foundation as Well as Their Limit

Values

(1)Damage Levels of Member

It is considered appropriate to determine a

damage level to a member by considering the

3-14

relation among the property of the member, state

of damage, and repairing methods. Moreover the

relationship between the damage levels and the

displacements on the load-displacement curve

should also be taken into account. As an

example, the following shows how to set the

damage levels for a member of reinforced

concrete.

ill case the bending failure mode occurs firstly

under the condition that the exerting compressive

axial force is of a general level, the load

deformation relation of the member is shown in

Fig.3.3.2.1. It is considered that some physical

phenomena reflecting the stress-strain condition

of the member, as shown in this figure, occur at

the changing points of the envelop curve.

Taking this member's characteristics into

consideration, each damage level of the member

is determined corresponding to the deformation

range as the following.

CDDamage Levell: before the point of B

®Damage Level 2: from B to C

@DamageLevel3:fromC to D

@Damage Level 4: after D

Once the relationship between the damage

level and the deformation is established, the


Table 3.3.2.1 Relationship between the damage levels of member and rotational angles

------------ Limit Value of Rotational Angle

Damage Level 1 eyd : Yielding rotational angle of member

Damage Level 2 emd : Rotational angle of member corresponding to the maximumdeformation resulting from the peak lateral loading

Damage Level 3 end Rotational angle of member corresponding to the maximumdeformation being able to resist the yield lateral load

Damage Level 4eud: Rotational angle of member for limiting the excessive deformation

in axial direction

P y: Yield bearing capacity

. Pm: Maximum bearing capacityCOy: Yield displacement

Om: Displacement corresponding to

maximum loadou : Ultimate displacement

Level 3Level 2Level 1

p

p Bm ...........•................._._..•......_ :.:; _-------.

p Ay _ .

Oy Om Ou

Fig.3.3.2.2 Imagine of load-displacement curve as well as stability levels offoundation

value of deformation becomes a suitable index

for checking the damage level, which may be

directly calculated from a response analysis. If

the member's nonlinear behavior is evaluated

with a mechanical model of bar, generally, the

rotational angle or the curvature for the section

of plastic hinge is taken as the index for the

member checking. The relationship between

them is shown in Table 3.3.2.1.

(2)Stability Levels of Foundation

Since the stability levels of foundation have a

great impact on the seismic performances of

overall structure, how to determine them

properly is important. In order to ensure the

seismic performance for an overall structure, the

stability levels of foundation should be

determined in term of two aspects. One is the

damage levels with respect to the stability of the

foundation itself. The other is the damage level

to the members constituting the foundation. For

the latter one, the procedure to determine the

damage levels of member is same as what

described previously. As to the procedure for

determination of damage levels to the foundation

stability, the following items should be taken into

account.

CDThe effects on the usage property of structure

3-15


due to the displacement of foundation

®The variation of bearing capacity of the

foundation after an earthquake

As indexes for evaluating these items, response

ductility ratio as well as residual displacement of

foundation should be used. The former is

defined as the ratio of the foundation's seismic

response displacement to yield displacement that

is determined by the load-displacement curve of

the foundation. Fig.3.3 .2.2 gives a general

illustration of the load-displacement curve as

well as the stability levels of the foundation.

Using the indexes of displacement in this figure,

the stability levels of foundation can be

determined as follows.

G)Stability Levell: In principle, load acting on

the foundation should be less than its yield

bearing capacity and no excessive

displacement occurs. Stress resultant of

members composing the foundation should not

exceed yield strength.

@Stability Level 2: Either subgrade supporting

the foundation, members composing the

foundation or both are deformed plastically, but

yet maintain sufficient bearing capacity. No

displacement detrimental to maintenance of the

structure's functions nor residual displacement

should be allowable after an earthquake.

@Stability Level 3 : Sufficient bearing capacity

should be maintained to protect the structure

from collapse by damage of the bearing

subgrade or members.

8j: Damage parts

Fig.3.3.2.3 Illustration of damaged parts of a rigid frame viaduct

Table 3.3.2.2 An example of the relationship among the limit values of structure's seismic

performance levels, member's damage levels and foundation's stability levels (rigid frame

viaduct)

Structure SPI SPIT SPill

Superstructure Girder and Underground Beam 1 2 3

Damage Level ofOther Beam 1 3 4Member

Column 1 3 3

Stability Level of Foundation 1 2 3

3-16


Besides the values of stability level are set

corresponding to the types of foundation.

(3)Limit Values

Based on the consideration explained above,

the parts where the damage may occur to a rigid

frame viaduct are illustrated in Fig.3.3.2.3, and

an example of the relationship among the limit

values of structure's seismic performance levels,

member's damage levels and foundation's

stability levels is shown in Table 3.3.2.2.

3.4 Concept of Importance Degree of

Structure

Determination of the importance degree of a

railway structure requires consideration various

factors, for example, the possible influences on

human life, society, neighborhood, operating

speeds and timetable of trains, and the difficulty

degree of recovery in case of damage. Based on

this concept, greater importance has been given

to the following structures.

(DStructures of the Shinkansen bullet lines and

those of passenger railway lines in major

metropolitan cities

@Structures whose recovery after an earthquake

is considered very difficult, for example a cut

and cover tunnel, etc.

3.5 Evaluation of Surface Ground and

Calculation of Displacement and Stress

of Structure

According to what shown in Fig.3.1.1.1, the

procedure for seismic design of a viaduct is,

inputting the L2 earthquake motions on the

bedrock firstly, evaluation of surface ground,

calculation of response of the structure and

evaluation of its seismic performance. In this

case, since the L2 earthquake motions are so

intense, both the ground and the structure are

expected to behavior strongly nonlinearly.

3-17

Therefore, how to evaluate the nonlinear effects

of ground and structure becomes an essential

task in seismic design.

3.5.1 Evaluation of Surface Ground

Characteristics of the surface ground must be

carefully analyzed because of its big impact on

the seismic performance of the structure to be

built. Generally there are 3 kinds of problems

that may be encountered and difficult to handle

in design practice: irregularity in topography or

geology, liquefaction, and soft or very soft soil

profile. In this section, the consideration and

analysis approaches adopted in the code to deal

with these special kinds of surface ground are

described.

(1) Irregular Surface Ground

From the past damage reconnaissance after

earthquakes, it is often observed that severe

damage happened on a ground with irregularity in

topography or geology. The cause for this

phenomenon is obvious that the superposition of

reflection waves resulting from the irregularities

of surface ground make the response amplified.

At this time even though there are some analysis

methods with rigorous numerical models may

evaluate such irregularity effects precisely, the

necessary of large amount of precise input

I Groundmodel for 2D analysis IInclination e

'I """'_"""-,.. IIFig.3.5.1.1 Ground models used for 10 and 20

analyses


parameters makes such analysis impractical for

general use. Therefore, a simplified method that

can estimate the amplification of earthquake

motion caused by irregularity and satisfy the

accuracy for seismic design is needed.

The incident wave is input at the bottom of the

hard layer and the response analysis method used

in the investigation is FEM. The time-history

Fig.3.5.1.2 Responses of horizontal

acceleration obtained from 20-FEM

analysis (normalized by peak value of

input wave)

Fig.3.5.1.3 Responses of horizontal

acceleration obtained by subtracting

values of 10 from 20 (normalized by

peak value of input wave)

H=20(m)

H~m)

201510Time(s)

Time(s)

l,x=<sU( ,.,"

,

I~~I~ iT> ,- ,. <I"(}In H " .,,'.. '.,

,•••••••

,,"'"

,.. '. ... ' .: .-:,'., ... "

, -:,

~bv

'.' ... '.AUlm

6Qlin... TT ..:

.: ,.,

.'.:,' .. " <

•••••••• :,... '. ,.

, ... ' ... ,.., ... .:... ,'·"1"'" ,

",". IT160:in ...,

"'....

"

.'.<..'.,' '. "~ .

,20

2) Effects of geological irregularity

In order to elucidate the mechanism of

amplification of earthquake motion due to the

geological irregularity, some numerical

investigations have been conducted as follows.

Firstly, the responses of 2D and 1D modeled

grounds were calculated, respectively. Then the

differences of the response between the 2D and

ID models were extracted by subtracting the

results of 1D from those of2D. These differences

represent the effects of geological irregularity,

because the responses due to the laterally

propagating waves that rebound on the inclined

boundary of hard layer are included in the results

of2D.

1) Ground models

Since ground motion amplification is affected

by various factors, such as the scale of irregular

shape and the characteristics of input motion, etc.

it is almost impossible to take all the factors into

account in the response analysis of surface ground.

For this reason, a ground model with rather

simple irregularity is considered in this

investigation. As what illustrated in Fig.3.5.1.1,

there are. two types of models prepared for

analysis, one is the two-dimensional (2D), and the

other is the one-dimensional (1D). For 2D

analyses, the property of surface ground is

modeled by the 3 key parameters: the inclination

angle ((J) of hard layer (bedrock), the thickness

(If) of soft layer, and the impedance ratio (IC ) of

the two layers. In ID analyses, all conditions

such as the properties of soil profile and the

thickness of soft and hard layers are set equal to

those of the corresponding 2D models as shown in

the Fig.3.5.1.1.

3-18


®Determination of a andLJ. t

It is easy to be conjectured that the coefficient

a: and LJ. t in Formula (3.5.1.1) are dependent on

f'(t)a=-,-,

f(t)

the Fourier spectra of earthquake

motion on the surface of irregular

ground;

the amplitude ratio between the

horizontally propagating wave

l' (t) and the vertically

propagating SH wave J(t);

the Fourier spectra of SH wave

J(t) ;

the delay time between J' (t) and

J(t) ;

G(CtJ, x)= F(CtJ)+ a .F(w)- e-imAt

=F(CtJ). ~ +a .e-i~t}= F(w)·r;(w)

(3.5.1.1)

(3.5.1.2)

a:

G(CtJ,x)Where,

LJ.t

3) Simplified methodology for evaluation of

geological irregularity

CDFormulae for estimating ground motion

Based on the results of numerical investigation

above, the influence of the irregularity upon the

earthquake motion on the ground surface is

possible to be modified with the following

expressions.

response acceleration on the ground surface

calculated by 2D FEM are shown in Fig.3.5.1.2,

whose incident wave is a SH Ricker wavelet with

a predominant period of O.8sec that is the same as

the fundamental resonance period of the soft layer.

Since the soft layer is rather thick, the

amplification property of the irregular ground is

obvious with the normalized peak value of 2.4 in

the vicinity of the edge of basin (x=40m). The

duration time of response becomes longer at

places remote from the edge of basin, These

characteristics are attributed to both the thickness

of soft layer and the irregularity in geology. Then,

the responses purely caused by the irregularity can

be obtained by subtracting the results of ID from

those of 2D (See Fig.3.5.1.3). The response

waves shown in Fig.3.5.1.3 are caused by a

laterally propagating wave that is generated at the

edge of the basin. Since the phase velocity of this

wave is nearly equal to that of Raleigh wave, the

major component included in the laterally

propagating wave is presumed to be Raleigh wave.

From these results of the numerical investigations,

a phenomenon is revealed that the earthquake

motions on the surface of the ground with

irregularity are synthesized from two parts, one is

the SH wave propagating directly form the

bedrock, the other one is Raleigh wave

propagating horizontally.

+

Fig.3.5.1.4 Definition of the parameters for irregular ground

3-19


the properties of geometry and material of the

irregular ground. Through an amount of

parametrical studies, the relationship among a

and the geometrical parameters e(the inclination

angle of hard layer), H (the thickness of soft

layer), and the material parameter K (the

impedance ratio of soft and hard layers) is

empirically expressed as the following".

a = 0.3xexp( - 7~OJxJr xexp(-O.44X)

(3.5.1.3)

formulae. Fig.3.5.1.5 gives the comparison of

Fourier amplitudes calculated by 2D-FEM and the

results of Gem, x) obtained by the empirical

formula (3.5.1.1). As to the time-history

responses of acceleration, the comparisons

between the two methods are shown in Fig.3.5.1.6.

The good agreement between them proves that the

accuracy degree of the empirical formulae is

sufficient to the level of seismic design.

Accordingly, in practice it is adequate to apply the

simplified methodology for general use.

(3.5.1A)

Where, X = x I H represents the normalized

distance from the origin at the edge point as

shown in Fig.3.5.1A.

With the same procedures the empirical formula

for delay time,.,d t is obtained as follows.

----

20

'1/ V :

Time(sec)10

1500 ,-----.,-----.,------,--------,:; tx=O(m)----.: 1\ --;;---J--.. _.=9 ~. . I !lA1 .n "" , -i\l\rl~~ 0 ;..,/ .. V..... ~

·1500 ~~~= method

fr~-lrS}--+~__4'1500~' . . ~

:iJ.000 I-······..;-·~ ..eeo....2:ao

r;.;. 500 1-'-"';-'.".

1500 x=20(m)

---2D FEM. - Proposed method .-+...,..-t-t-t-tf-l

o'---'--'--'--'-'---'---"-''-=-'.......0.1 0.5 1 5 10

Frequency(Hz)

Fig.3.5.1.5 Comparison of Fourier amplitudes

between the 2D-FEM and the empirical

method (x=20 m)

Fig.3.5.1.6 Comparison of time-history

responses between the 2D-FEM and

the empirical method (x=O, 20, 40 m)

the

the

in Fig.3.5.1.4,illustrated

shear velocity of the hard layer;

shear velocity of the soft layer;

travelling velocity of the

horizontally propagating wave

within the range where the hard

layer slants;

phase velocity of the Raleigh wave;CJ!...OJ) :

Where, as

meanings of the main parameters are as

following.

Vb

CDAdequacy of the simplified methodology

In the code, the simplified methodology

described above is proposed for evaluation of the

ground irregularity in general use. According to

this methodology, in a general case a 2D response

analysis of irregular ground can be omitted and

the irregularity effect is taken into account by

modifying the response of ID analysis with the

empirical formulae. Therefore, it is necessary to

grasp the calculation accuracy of the empirical

3-20


Therefore, the hitherto applied relationship

1000

1-.~_1-.1••1.1J.bll-'-s-arc promoted

10 100

Nnber of cycle(Nc)

o~~10 ~ • Or<SQ'X;~ ... Dr=::70$

~\ .Or-SO'!.

+ '1\• Or-9O$

~r-,-,+ 4ill

:r. --::::I:-l" ~\~ .I- + I • +

+

•,.~ !, r-t---•

I I0.0

1

O.S

1.0

.€.~ 1.5

2.0

2.S

liquefaction judgement

Since the intense level of the earthquakes which

occurred before the Hyogoken-Nanbu Earthquake

was not so high, the relative density of the soil

profile incurred liquefaction was low.

Accordingly the relationship between R and Nc

was obtained based on experimental results that

correspond to the values of relative density below

60%. Furthermore, this relationship was

determined independent to the relative density of

soil profiles.

After the Hyogoken-Nanbu Earthquake, the

intensity level of design earthquake as well as the

density level of the soil profile needed

Liquefaction is a very serious problem to

consider in seismic design. During the past

earthquakes, there were an amount of damages to

infrastructures caused by liquefaction or

subsequent lateral flow. Therefore, for the

ground with liquefaction possibility, if any

financially feasible measure is available, such as

ground improvement that can prevent

liquefaction to happen, it should be implemented.

If not, the overall structure, including the

superstructure, should be taken care of

comprehensive measures to prevent collapse or

other disastrous damage against excessive

response the structure may incur due to

liquefaction or lateral flow.

In this code, the procedures for liquefaction

judgment as well as decrease in coefficient of

subgrade reaction to consider the effect of

liquefaction and subsequent lateral flow are

determined.

(2) Liquefied Surface Ground

1) Liquefaction judgement

In liquefaction judgement for railway structure

design, the following expression is applied.

Fig.3.5.1.7 Relationship between the ratio of

liquefaction strength (R) and the

number of cycles (Nc)

Fig.3.5.1.8 Relationship between the ratio of

liquefaction strength and the number of

cycles (DA=10% to15%)

0.01 1000100

NootIer of. eyels(Ne)

10

\[1\ ~ I~I\

DA-"""2Q%

DA=5~ 1\

'---......

~I--~DA=5~

t--t--t-0.5

3.0

1.0

2.0

2.5

t!.~ 1.5

( 3.5.1.5)

Where,

FL factor of liquefaction resistance;

R ratio of liquefaction strength;

L maximum shear stress ratio;

The ratio of liquefaction strength (R) is

determined by correcting the standard values of

liquefaction strength ratio that are obtained from

cyclic triaxial tests or in-site tests. In this

correction the concept of accumulated damage

index is introduced to reflect the irregularity effect

of earthquake motion. Therefore, the relationship

between R (ratio of liquefaction strength) and N,

(number of alternative cycles of earthquake

motion) becomes required.

3-21


20<P,

Structures are designed by seismic defomation method.

+-- ground surface propertiesfor- seismic design

~ range to reduce the coefficient~ of subgrade reaction

--ground surface propertiesfor seismic design

__ground surface pruper-t.iesfor seismic design

_ range to reduce the coefficientof subgrade reaction

Ground surface properties for se iseric design set up theshallowest layer to reduce the coefficient of subgrade reaction.

Ground surface properties for seismic design set up thedeepest layer to reduce the coefficient of subarsde reaction.

ground surface

range for lowering the coefficients of subgrade

and the parameters concerning ground properties

is judged by using the liquefaction coefficient, an

index of the degree of liquefaction, for different

values of liquefaction resistance given by the

Fig.3.5.1.9 Range to reduce the coefficient of

subgrade reaction and ground surface

properties for seismic design

2) Reducing the coefficient of subgrade

reaction as the effect of liquefaction

Basing on some researches, in this code the

between R and N; corresponding to the low

relative density is considered too conservative, in

which it mistakenly leads to judgement that even

soil at a high relative density free from

liquefaction will liquefy.

On the other hand, the influence of liquefaction

on the dynamic response of structure is taken into

account by reducing the coefficient of subgrade

reaction according to the situation ofliquefaction.

Accordingly, the decrease in coefficient of

subgrade reaction should be formulated varying

with the degree of liquefaction and the depth from

ground surface.

Fig.3.5.1.7 shows the relationship between R

and N, corresponding to different relative density

obtained from cyclic triaxial tests of dense sandy

soil. This result reveals a fact that the sandy soil

with relative density below 50% its relationship

between R and N, can be considered independent

to relative density, but if the density over the

value of 50% the relationship should be

determined by taking the effect of relative density

into account.

So far, the double amplitude (DA) of axial strain

used as index for liquefaction judgement is set to

5% as the critical value. This is proper to L1

design earthquake motion, but for L2 earthquake

the critical value of DA is promoted to 10% or

15% by considering the structural seismic

performance levels and the return period of

earthquake'?',

In the strain range of 10 to 15%, dense soil will

show cyclic mobility or positive dilantancy,

recover the effective stress and present high

stiffness against cyclic shear stresses. In this

situation, the dynamic shear strength ratio

becomes larger as shown in Fig.3.5.1.8, which

means that the soil will not liquefy.

3-22


Fig.3.5.1.10 Relationship between excess

pore water pressure and shear strain

3) Liquefaction-induced lateral flow

From mechanism elucidation of foundations

damaged by liquefaction-induced lateral flow of

ground in past earthquakes, it is understood that

loads existing at foundation are composed of drag

force due to liquefaction layer and load due to

ground displacement of non-liquefaction layer.

However, this kind of investigation needs precise

analysis models and sophisticated technology for

numerical computation. It is impractical for

seismic design, especially, estimating the drag

force is very difficult.

Since the most important task in seismic design

structure, the methods for evaluating the drag

force and the load due to ground displacement in

this code, are determined based on the

consideration that the calculated response

displacements of structure can fit well to those

obtained by experiments. The illustration of this

concept is shown in Fig.3.5.1.11, where the

effects of the lateral flow are expressed with the

loads exerted to the upper and lower parts of the

foundation, respectively. The upper part load is

transferred equivalently from the displacement of

the non-liquefaction layer through the spring

constants of subgrade. The lower part load is due

to the lateral flow of liquefaction layer.

Displacements of structure calculated with this

model are a little bit larger than those obtained in

experiments. Therefore, this design methodology

is considered in the safety side.

(3) Surface Ground with Soft Soil Profile

The amplification property of surface ground

with soft soil profile has been testified in many

past earthquakes. This property will cause big

effects on structure design. For this reason,

response analysis of surface ground under an

intense earthquake should be conducted carefully

and precisely.

For dynamic analysis of surface ground, a shear

(3.5.1.6)

110

~2 -1 010 10 10

Shear strain (%)

r20

PL =Jo (1 - FL )wdzW =1O-0.5z

o -410

in which PL is the liquefaction potential; FL is

the factor of liquefaction resistance; and z is the

depth in meters.

Fig.3.5.1.9 shows the range to reduce the

coefficient of subgrade reaction and ground

surface properties for seismic design by referring

to the calculation result of the liquefaction

potential of a few ground models. This

formulation considers sudden changes of ground

condition for liquefaction.

Moreover, this covers soft ground at the

liquefaction potential of less than 5, and structures

are designed by the seismic deformation method.

Because the ground strain considered by the

seismic deformation method is 0.1%, this is a case

of liquefaction potential of less than 5. Namely,

this is a state where the strain has risen a little in

excess pore water pressure in Fig.3.5.1.10. It is

the state of just before liquefaction.

equation .(3.5.1.6), where the weighting

coefficient for the depth is set to reflect the effect

on structures.

3-23


stress-strain model used should be satisfied to the

following conditions.

The model,

(Dcan express the stress-strain relationship ('" "-'

r ) for various geo-materials ranging from soft

clay to hard rock over a wide pre-failure strain

range;

®has a minimum possible number of parameters

to describe the model, each possessing clear

physical meaning;

@can express the damping-strain relationship (h

"-' r ) over a wide strain range obtained from

laboratory tests;

@)can reflect the concept of failure strength;

@can easily be applicable to seismic design.

In this code, a model of shear stress-strain

satisfy the conditions above is proposed. This

model fits dynamic deformation characteristics

obtained from laboratory tests, such as G/Gmax~

r , hr- r relationships over a wide strain range,

and reflects failure limit stress r » Furthermore,

the adequacy of the model was examined by

model ground test with shaking table.

(" ....,! j

non effective I non-Iiquifaction layerranp of lateral flow

3.5.2 Calculation of Responses of Structures

Dynamic analysis should be the main method

for seismic design of bridges. In this case, how

to setting the nonlinear behavior for structural

members is very important. In the code, the

member's non-linearity is prescribed according

to what shown in Fig.3.3.2.1 and Fig.3.3.2.2. In

addition to the skeleton curves, the hysteresis

loops for determining damping constant are also

required. In the code, they are given with respect

to the types of material and foundations.

Moreover, it is very convenient to use nonlinear

spectrum method (the simplified dynamic

analysis method as shown in Fig.3.1.1.1) to

calculate the ductility ratio of structure for a

general case.

Fig.3.5.2.1 gives an example of the demand

yield-seismic-coefficient spectrum that is applied

in nonlinear spectrum method. The spectrum is

applicable to a general structure, and the

procedure for making out it is: i) modeling the

structure to a single-degree-of-freedom system,

ii) calculating the maximum nonlinear response

displacement of the structure under the design

earthquake motion; iii) plotting the relationship

load as lateral flow of non-Iiqufaction layer

model of analysis

Fig.3.5.1.11 Illustration of design methodology for lateral flow induced by liquefaction

3-24


between the yield seismic coefficient and the

natural period corresponding to each ductility

ratio.

For such structures as multiple-spanned

bridges, structures with long natural periods, or

new types of bridges whose behavior cannot be

expressed with the system of single degree of

freedom, detailed dynamic analysis method using

the model of multiple degrees of freedom should

be chosen.

With regard to foundation structures, dynamic

response analysis should also be chosen as main

way for design. In case of surface ground with

soft soil profile, the ground displacement

resulting from an earthquake is generally beyond

negligible levels, especially when the earthquake

is intense the ground displacement may cause

severe damage to a deep foundation embedded.

In this code, therefore, it is prescribed that the

effect on deep foundation due to ground

displacement should be taken into account by

using so-called seismic deformation method.

Until now only this code has the stipulation,

and in other codes deep foundations are designed

merely against seismic inertial force. However,

ground displacement caused by an earthquake

will generate curvature of deep foundation and

subsequently bending moment along its whole

length, which makes stress resultant increase

within the foundation's members. Therefore the

design method considering only initial force is at

unsafe side, especially in case of intense

earthquakes.

The seismic deformation method prescribed in

the code is a cost-effective one that can

conveniently combine the both effects coming

from the inertial force of superstructure and the

displacement of ground according to relationship

between the natural periods of structure and

ground.

3.6 Safety (Seismic Performance) Checking

of Structures

In checking seismic performance of a structure,

the prescribed procedure in the code specifies

that responses calculated as in Section 3.5 should

satisfy the limit values of the member's damage

levels and the foundation's stability levels, both

mentioned in Section 3.3. The flowchart for

5

InputWave: L2 Earthquake Motion; Objective Structure: Surperstructures ofRC or SRC

-cQ)

'(3

~ooo

:~ 0.5 -------------------_-----~--§--§~~~~ ----------------------------------------pT~rOr i . '

(j) ---------------------------:---------------,----------,--------,------r-----'- -, Nonlinear Behavior: Clough model>=~ I-----------------------~-------------.L-------j--------L-----:-----f--..L ~::s:x~::::~;~~~:~~::~~g Stiffness: 0.1 .

~ I Type ofSoil Profile: G3 j ii' Initial Damping Coefficienth=0.04/T, (0.10~h~0.20)0.1 1 i0.1 0.5 1

Equivalent Natural Period (sec)

Fig.3.5.2.1 An example of Demand-Yield-Seismic-Coefficient Spectrum (Earthquake Motion:

Spectrum II, Surface Ground: G3 Type)

3-25


Setting design earthquake motion and soil profile

•Modeling structure foranalysis

i.-Static nonlinear analysis (Pushoveranalysis): ~ Setting nonlinear property for

Lateral load-displacement relationship """"" members and subgrade

~Grasping seismic performance ofstructure

(khy: yield seismic coefficient; Teq: equivalent natural period; deformation property)

~Calculating response ofstructure;

I~Demand-yield-seismic-coeffcient spectrum

Ductility ratio or Detailed dynamic analysis method

~Checking seismic performance of structure

(Members: damage level; Foundation: stability level)

Fig.3.6.1.1 Procedure of seismic performance checking for bridges and viaducts

such a procedure is shown in Fig.3.6.1.1.

Static nonlinear analysis method (pushover

Analysis method), in the code, is stipulated to

apply in the checking process. The procedure of

pushover analysis is, i) modeling overall

structure (from superstructure to foundation) to a

frame structure, and subgrade supporting the

foundation to a system composed of springs; ii)

setting the strengths and deformation behaviors

for the structural members and the subgrade

reaction according to what described previously;

iii) calculating the displacement of structure by

increasing seismic load step by step and plotting

the relationship between the seismic load and the

displacement. In this way, the failure process of

the overall structure can be grasped by indicating

the various critical steps in the load-displacement

3-26

curve. Such critical steps include the steps

where the structural capacities reach to the limit

values of yield, maximum and ultimate. The

ultimate displacement can be determined by

comparing the calculated displacement with the

limit values listed in Table 3.3 .2.1. For the

superstructure and foundation, when the

member's capacity of whichever reaches to the

limit value of ultimate state, the displacement is

determined as the ultimate displacement for the

overall structure.

Therefore, if the value of the ultimate

displacement determined as above is larger than

response displacement calculated by a dynamic

analysis method, it means that the structural

seismic performance designed satisfy the

objective of seismic performance level, and a


safety judgement is obtained. Furthermore, the

judgement of each member's damage level and

foundation's stability level should be conducted

by checking the deformation state of the step in

the pushover analysis, whose displacement is as

same as that calculated by the dynamic analysis

method. The main contents about this checking

are described as follows.

the allowable values of residual displacement

should be limited within a small range.

All the items above are checked according to

the results obtained by the static nonlinear

analysis.

3.6.3 An Example of Safety Checking of Pile

Foundation

3.6.1 Checking Damage Levels of Members

In the code, the following items are stipulated

for the checking of stability levels of foundation.

CDThe response ductility ratio of foundation;

®The damage levels of the members composing

the foundation;

@The residual displacement of foundation.

The residual displacement above is taken as a

main index for checking the Seismic

Performance II. That is to say in order to make

quick recovery of the function for train operation,

In checking the damage levels of members

made of concrete, failure mode should be judged

at first, namely, if shear stress calculated is

smaller than shear strength when bending

strength is reached, the failure mode is defined as

bending failure mode, inversely shearing failure

mode. In the code, it is stipulated that the real

strength of reinforcing bar should be used in the

failure mode judgement.

In case of bending failure mode, the damage

levels can be judged with the deformation results

calculated from static nonlinear analysis. For the

case of shearing failure mode, however, the

judgement can only be conducted according to

the strength. That is to say the deformation

behavior of the member with shearing failure

mode should be set to linearity in the overall

structural model for the static nonlinear analysis.

3.6.2 Checking Stability

Foundation

Levels of

3-27

(1) Seismic Performance Levels of Pile

Foundation

The seismic performance levels of pile

foundation are determined by the stability levels

or pue foundation, The stability levels of pile

foundation are determined by considering the

strength and deformation properties of subsoil and

pile members. Table 3.6.3.1 shows the definition

of the state of pile foundation corresponding to

the seismic performance.

(2) Pushover Analysis

Table 3.6.3.1 State of pile foundation corresponding

to the seismic performance levels

Seismic StabilityPerformance Level of State of Pile Foundation

Level Foundation

SPI Level 1 Pile foundation do not yield.

Although pile foundationSPII Level 2 yields, it maintains a

sufficient bearing capacity.

Although pile foundation

SPill Level 3reaches the ultimate state,super structure does notcollapse.

1) Structural analysis model

In the pushover analysis, super structures and

pile foundations are modeled as a overall

structural system (Fig.3.6.3.1), which includes the


Kv : Vertical subgrade reaction of pile pointKsv : Vertical subgrade reaction of pile surfaceK.h : Horizontal resisitance of pile.Khf : Horizontal resisitance of footing

(a) Pier type

EIs

Elp

I

(b) Rigid frame type

K.h

------i

iLKhf

I

Ell

Elr

ili !Elp i

I

2) Characteristics of ground resistance

The property of ground resistance of pile

foundation is assumed to be represented by an

elasto-plastic model (bilinear type). Fig.3.6.3.2

shows an example of the ground resistance model

that becomes plastic when the subgrade reaction

of each ground resistance reaches the upper limit.

nonlinear properties of both the subgrade and

structures. The springs expressing the subgrade

reaction are attached to the nodal points, and the

parts connecting the pile to the spread footing and

the pile to the embedded lateral beams are

assumed to be rigid.

Fig.3.6.3.1 Structural model for viaducts

(c) Horizontal on pile surface

Fig.3.6.3.2 Models for ground resistance

criterions. In this case, the yield point can be

determined by taking into account the causes

which intensify the displacement rapidly in the

load-displacement curve.

liwDisplacementof pile skin

Rp: Design point bearing capacityof single pile

Re: Design skin friction capacityof single pile

Pe : Effective resistance earth pressure

Indentation side

Vertical subgrade reactionof pile skin

---'----I-U n hPulling side

(b) Vertical on pile surface

Displacementof pile point

IihHorizontal displacement

Effective resistanceearth pressure

Indentation sideRp f--~---

0'---'-------

Vertical subgrade reactionof pile point

(a) Vertical atpile tip

Pulling side

Horizontal resiaitance ofpile I

3) Yield point of pile foundation

Yield point of a pile foundation is established

according to the load-displacement curve of an

overall structure, where the displacement

increases rapidly mainly because of the subgrade

reaction reaching the upper limit values or the

stiffness of pile members decreasing due to the

strength yielding. However, the yield point where

the displacement rapidly increases in the 1000

displacement curve varies for different types of

foundations. This makes it difficult to judge the

yield point from i) the degree to which the

subgrade reaction exceeds the upper limit values

and ii) the number of members damaged over the

total number of members.

In order to investigate the causes of yield point,

some common prototype pile foundations were

chosen for trial designing. As a result, it was

confirmed that the yield point appears when i) the

subgrade reaction yields at the outermost edge of

the indentation in side of pile group and ii) half of

the total number of pile members yields.

In the code, therefore, it is stipulated that the

yield point of pile foundation with a common

shape can be determined as the point when it

reaches one of the states shown in Table 3.6.3.2.

If a pile foundation has too many piles, it is

difficult to determine the yield point by these

3-28


Table 3.6.3.2 Yield point definition for pile foundation

Subgrade in the When the vertical resistance of pile

indentation-in head atthe outermost edge reach the

side of pile groupupper limit value of design verticalcapacity

Subgrade in theWhen the vertical resistance of the

pulling-out side ofhead of a half (ignoring fractions) of

pile grouptotal piles reach the upper limit ofdesign pull-out resistance

Pile membersWhen the strength of a half (ignoringfractions) of the total piles yield

(3)Response Analysis of Pile Foundation

To check the stability levels of pile foundation,

the response values of pile foundation due to the

design earthquake motion should be calculated

firstly. Then the stability level can be determined

by comparing the response values with the

indexes of ductility, damage level and response

displacement. The response analysis should be

conducted by using the dynamic analysis method

which is chosen by the designer out of the

following by taking into account the ground and

structure conditions.

CDNon-linearspectra method

®Analysis method with springs supporting

foundation

@Analysis method considering the soil-pile

structure interaction

For the method CD or ® above, the procedure of

pushover analysis is needed. But for the method

@, only the member's properties and the

properties of ground resistance as illustrated in

Fig.3.6.3.2 are needed.

(4) Checking Stability Levels of Pile

Foundation

1) Response ductility ratios

In the code, the safety checking of pile

foundation is stipulated to check the ductility ratio

3-29

of the foundation. Table 3.6.3.3 gives the limit

values of response ductility ratio corresponding to

various stability levels for cast-in-place pile,

which are prescribed in the code. Furthermore,

the limit values of ductility ratio are based on the

results of loading experiments. If there is the

sufficient strength left for pile members, the limit

values can be determined by other methods while

taking the damage process into account.

Table 3.6.3.3 Stability Levels and Limit Values of

Ductility RatioLimit value of ductility factor

jL L

Stability Stability Stabilitylevel 1 level 2 level 3

Cast-In- 1 5 8Place Pile

2) Damage levels of members

In the seismic design, it is necessary to confirm

that the demanded damage level of each pile

member is satisfied. Referring to some studies'",

it is understood that even when the damage level

of one part of a pile group exceeds the damage

level 1 or 2, the strength remaining for the overall

structural system is enough. Therefor, in the code,

the limit values for the damage levels of pile

members have been relaxed.

3) Response displacement

It is confirmed that the values of response

displacement or residual displacement should be

less than the limit values corresponding to various

stability levels.

3.7 Conclusions

The outline of the new seismic design code for

railway structures has been described above.

Because of the limited space in this article, only

the basic principles and some.important advances

for the seismic design are introduced.

The adequacy of seismic design methodology


should be confirmed through precise analysis of

real damage examples incurred in past

earthquakes. The methodologies introduced here

are based on the results of damage analyses

concerning to the Hyogoken-Nanbu Earthquake.

Since these damage analyses are still being

conducted by each organization, currently, some

new knowledge or consideration may be

obtained in the near future. Consequently, by

absorbing that information, the current seismic

design methodology can become as perfect as

possible.

Moreover, the methodology for seismic designL_~____ __.4-L__ _ 1.: __ .4-_..1 1-. ~_ _.&ucco.urcs 1 <1Wc;l I,;VllljJlll,;<1lc;U Uc;l,;<1U1\C' Vi.

consideration of the both non-linearity corning

from the structures and the subgrade. In order to

avoid meaningless complication, the described

approaches taken in the seismic design are the

essential ones that can express the damage levels

of structures. Therefore, by using these

approaches the state of damage to designed

structures during an intense earthquake can be

predicted corresponding to the seismic

performance levels.

At last, there is a notice that the precision of

the input parameters concerning structures and

subgrade and the computing accuracy should be

appropriate to the execution of computer. Even

though the level of design method is promoted, a

design using incorrect input data can not be

considered as a good one.

REFERENCES1) Seismic Design Code for Railway Structures,

published by MARUZEN, Oct, 1999. (in

Japanese)

2) Akihiko Nishimura: "Earthquake resistant design

for Railway Structures", Quarterly Report of

RT.RI., VOl.37, No.3, pp.128-138, 1996.

3) "Proposal on Earthquake Resistance for Civil

Engineering Structures", Special task committee

3-30

of earthquake resistance of civil engineering

structures, Japan

4) "Report on the investigation of disaster of

Earthquake in Hanshin-Awaji", Committee on the

investigation of disaster of Earthquake in

Hanshin-Awaji, The Japanese Geotechnical

Society. (in Japanese)

5) Gotou, Y, Ejiri, J.: "The characteristics of

amplification at the Tarzana observation station in

Northridge earthquake", Proceedings of

Amplification of Ground Motion on Soft Ground

Symposium, Japan, 1994.

6) Yoshimitsu Fukushima: "Empirical prediction for_.&...... .J __ .l.': J:1 __ .l._..J __ "-L ...: __ l1\UVllg gIVUllU lllVUVll lCJ.lCl,;lCU Vll WCVIC'U'-'''-l

backgrounds of source and propagation of seismic

wave", ORr Report 93-07, Ohsaki Research

Institute, March 1994. (in Japanese)

7) Susumu 0000, Katsuya Takahashi: "Evaluation

of strong-motion attenuation relation using near

source data in California", Proceedings of the 9th

Japan Earthquake Engineering Symposium, 1994.

(in Japanese)

8) Haibo Wang, Akihiko Nishimura:

"Determination of design seismic motion by

considering inland and interplate earthquakes",

Quarterly Report of RT.RI., Vol.40, No.3 ,

pp.130-138, 1999.

9) Yoshitaka Murono, Akihiko Nishimura:

"Characteristics of Local Site Effects on Seismic

Motion, --Non-linearity of Soil and Geological

Irregularity--", Quarterly Report of R T.RI.,

Vo1.40, No.3, pp.139-l45, 1999.

lO)Ryo Sawada, Akihiko Nishimura: "Design

Method of Structure Considering Liquefaction and

Subsequent Lateral Flow" , Quarterly Report of

RT.RI., VolAO, No.3, pp.146-l51, 1999.

11)Kimura, Okoshi, et al : An Experimental Study

on The Ductility of Pile Foundations, Journal of

Study Engineering, Vol.44A, 1998.3 (in Japanese)

4. EARTHQUAKE RESISTANT DESIGN OF PORT FACILITIES

BUREAU OF THE PORTS AND HARBORS, MINISTRY OF TRANSPORT

4.1 History or Revisions of Design Codes 4- 1

4.2 Damage to Port Facilities by Past Earthquakes 4- 3

4.2.1 Gravity Type Quaywalls 4- 3

4.2.2 Sheetpile Bulkheads 4- 4

4.2.3 Pile Supported Piers 4- 6

4.2.4 Breakwaters 4- 6

4.3 Evaluation of Seismic Performance 4- 7

4.3.1 General 4- 7

4.3.2 Seismic Performance Requirement for Port Facilities 4- 7

4.3.3 Pseudo-static Method 4- 8

4.3.4 Earthquake Response Analysis 4- 9

4.3.5 Seismic deformation method 4- 16

4.4 Earthquake Load 4- 18

4.4.1 Design Seismic Coefficient 4- 18

4.5 Lateral Earth Pressure and Water Pressure during Earthquake 4- 22

4.5.1 General 4- 22

4.5.2 Apparent Seismic Coefficient (Seismic Coefficient of Submerged Soil Layer) 4- 22

4.5.3 Dynamic Water Pressure During Earthquake 4- 22

4.6 Liquefaction Prediction/Determination Method 4- 22

4.6.1 General 4- 22

4.6.2 Grain Size Distribution and SPT-N Value 4- 22

4.6.3 Undrained Cyclic Triaxial Test and Seismic Response Analysis (Sensitive

Assess Method) 4- 24

4.7 Seismic Design of High Seismic Resistant Quaywalls 4- 25

4.7.1 Evaluation of Seismic Performance of High Seismic Resistant Facilities 4- 26

4.7.2 Design Seismic Coefficient of High Seismic Resistant Quay Walls 4- 26

4.8 New Seismic Design of Open Piled Piers 4- 31

4.8.1 General 4- 31

4.8.2 Seismic Performance Requirements 4- 31

4.8.3 Design Earthquake Forces 4- 32

4.8.4 Structural Analysis Procedures 4- 32


4. EARTHQUAKE RESISTANT DESIGN OF PORT FACIILITIESBUREAU OF THE PORTS AND HARBOURS MINISTRY OF TRANSPORT

4.1 History of revisions of design codes

Having been established in 1951, the Port and

Harbour Law in Japan has been revised many

times so far. The important revision in view of

the design of port and harbour facilities was made

in 1974, in which it was noticed that the port and

harbour facilities must be constructed, maintained

and rehabilitated in accordance with the Techni

cal Standard of Port and Harbour Facilities. In

1975 the engineering requirement was established

as the Ordinance of the Ministry of Transport and

it was prescribed in the ordinance that the faciliti

es in ports and harbours must be stable against

the loads such as earthquake loads, dead weights,

wave forces, impacts due to ships andso on.

The Technical Standard of Port and Harbour

Facilities was established in 1973 as the order of

the Director General of Bureau of the Ports and

Harbours, Ministry of" Transport, in which the

details on earthquake resistant design, such as

design procedures, factor of safety and allowable

stresses, were specified.

In 1979 the Technical Standard of Port and

Harbour Facilities and its Commentary was com

piled under the supervision of the Bureau of the

Ports and Harbours, Ministry of Transport, and

has been revised in every ten years after the 1979

edition.

Seismic stability of the port and harbour

structures was to be examined only by the

pseudo-static method in the 1979 edition and

1989 edition of the Technical Standards. The

pseudo-static method is called the seismic coeffi

cient method, and the earthquake load is obtained

by the multiplication of the design seismic coeffi

cient and the vertical load. The design seismic

coefficient is obtained by the multiplication of the

regional seismic coefficient, the factor for subsoil

condition, and the importance factor. Those three

4-1

factors were classified into three groups respec

tively, with the regional seismic coefficient

ranging from 0.05 to 0.15, the factor for subsoil

condition ranging from 0.8 to 1.2, and the im

portance factor ranging from 0.5 to 1.5. The re

sultant value, the design seismic coefficient, was

rounded off to the nearest 0.05 or 0.00. As to the

design of the reinforced concrete structures, al

lowable stress method was applied.

Significant modification had not been made as

to the earthquake resistant design procedures ever

since the first edition thus far, however, the pro

cedure of assessing the liquefaction potential was

not stated in the 1973 edition, and was firstly

stated in the 1979 edition.

In 1999, the order of the Director General of

Bureau of the Ports and Harbours was repealed

for variety of reasons, and the Ministry of Trans

port notified the new detailed Technical Standard.

In the new Technical Standard, some significant

revisions have been made based on the outcome

of the recent research after the 1995 Hyogoken

Nambu earthquake. Those are summarized as

follows:

(1) Principles of design

The concept of performance-based design

has been introduced. The principles are:

CDAil the structures must be stable against

the level 1 earthquake motions whose re

turn periods are about 75 years.

®High seismic resistant facilities should

keep the required performance against the

level 2 earthquake motions whose retum

periods are over some hundred years.

(2) Seismic coefficient method

CD The regional seismic coefficient and the

importance factor have been modified,

while the factor for subsoil condition has

remained as it was. The number of region

al groups for the regional seismic coeffi-


cient has come up to five, coefficient

ranging from 0.08 to 0.15. Range of the

importance factor has become from 0.8 to

1.5. In addition, the resultant value has

been considered down to three decimal

places.

(2) The equation for the apparent seismic

coefficient, which is the seismic coeffi

cient used for the calculation of earth

pressure below groundwater level, has

been modified.

® Consideration ofthe dynamic water pres

sures acting at the front of vertical walls

has been stated.

(3) Assess of earthquake-resistant performance

Assessing way of the earthquake-resistant

performance in view of level 2 earthquake

motions has been introduced.

(4) Assess of liquefaction potential

Assessing way of liquefaction potential has

been modified.

(5) Design method of open piled piers

Modified pseudo-static design method, whi

ch is called the modified seismic coefficient

method, has been introduced for the design

of open piled piers.

(6) Design method of reinforced concrete struc

tures

Limit state design method has been intro

duced, and safety factors for the design have

been established.

The history of revisions of design codes IS

summarized in Table 4.1.1.

Table 4.1.1 Summary of history of revisions of design codes

1973 1979 1989 1999

Earthquake design level One level Two levels

Seismic coefficient method 0 0modified

Performance-based design principles - 0

Assess of liquefaction potential - 0 0modified

Design of open piled piers Seismic coefficient Modified seismic coefficientmethod method

Design of reinforced concrete structure Allowable stress method Limit state design method

4-2


20.0

~t""'r',...,,-------.-------- muJl~ I: \\

L WL. ± 0.0 \:;.J---l....' -------.-, ~

- - - Before earthquake~ ---- After earthquakea

+3..tl2 = ,.,,--=-;;rIi "", ...vH.W.L+1.§ I :.. _ - - - - - .....i L.W.L+0.5 0.43 I ........... I I .....

I Caisson I "-~<I I Rubble "-I L B H I Backfill -, .....I 15.0x12.0XO.6 I >-

-9.11 I 1. "\) -- --_-..1---------.....:...< »>:- <:__-J~6_ B.u.!?ll!.el'1E.ull d__ .»>: Unit (m)

The 1993 Kushiro-Oki earthquake

A typical cross section of a gravity type quay

wall at Kushiro port is shown in FigA.2.2. As

shown in the figure, a caisson wall was put on a

firm foundation with SPT N-values ranging from

30 to 50, with a loose backfill. Shaken with a

peak bedrock acceleration of O.28g, residual dis

placement of the caisson walls ranged from Om to

OA3m, on average 0.24m.

FigA.2.1 Cross section and deformation of a

quaywalI at Gaiko District in Akita port

13.0

Concrete Caisson

the evidence of ground liquefaction and the

ground liquefaction behind the caisson might

have a major effect on the deformation of the

caisson and the settlement at the apron._______. before

-- after

FigA.2.2 Cross section and deformation of aquaywall at Kushiro port (West port DistrictNo.2 West quaywaIl-9m)

Port facilities ill Japan has been suffering

severe damage by earthquakes, such as the 1964

Niigata earthquake, the 1968 Tokachi-Oki earth

quake, the 1978 Miyagiken-Oki earthquake, the

1983 Nipponkai-Chubu earthquake, the 1993 Ku

shiro-Oki earthquake, the 1993 Hokkaido

Nansei-Oki earthquake, the 1995 Hyogoken

Nambu earthquake, etc. Earthquakes that induce

severe damage for port facilities have been occur

ring approximately once in five years in Japan.

The details of the damage caused by those earth

quakes were carefully surveyed and summarized

in the reports.

The observation of the strong-motion earth

quake at major ports in Japan has been conducted

since 1962 and strong ground motions by these

earthquakes were recorded at various ports.

Therefore, the relationship between ground mo

tion and damage of port facilities has been ex

amined carefully since 1962. Although the

mechanism and pattern of the damage depend on

the type of facilities, strong ground motion char

acteristics and geotechnical properties of founda

tion have a major effect on the extent of damage.

In this section, typical damage of various types

of port facilities and its mechanism are summa

rized considering the ground motion characteris

tics and geotechnical background.

4.2 Damage to port facilities by past

earthquakes

4.2.1 Gravity type quaywalls

The 1983 Nipponkai-Chubu earthquake

Figure 4.2.1 shows a cross section of a quay

wall at Gaiko district in Akita port. A typical

feature of the damage was a large settlement at

the apron in an order of 1.0 to 1.5m, and the cais

son wall inclined toward the sea by 1.6 degree.

Maximum horizontal displacement at the top of

the caisson was lAm. Observed was 0.22g of

maximum acceleration in Akita port. There was

4-3


- - - before-- after

The 1995 Hyougoken-Nambu earthquake

Gravity type quaywalls in Kobe port slid to

offshore side 1m to Sm and subsided 1m to 2m,

subsidence behind the quaywalls of 3m to 4m due

to the lateral deformation of the quaywall as indi

cated in FigA .2.3. As shown in this figure, a cais

son wall was put on loosely deposited decom

posed granite. A peak acceleration of 0.55g at a

depth of GL-32m was recorded at the Port Island

vertical seismic array site in Kobe port.

\7±O.

+2.37r,---~~~-------~~~r------I \I ~\ _

Tie Rod 1=11.0

EE

~ 0

c:: 0~

tti "S-Ols:Ul

~~

FigA.2.3 Cross section and deformation of aquaywall in Kobe port (RC-5, Rokko Island-14m)

Alluvial Clay LayerBackfillin Sandfor Replacing Clay Layer

'V-34.00~-36.00

Sand Drain

Unil(m}

FigA.2A Cross section of a sheetpile bulkhead

in Yamanoshita Revetment in Niigata port

T:'.: ~l'" C _L ........: 1:' .....L _

.rlgUlC '"t •.L . .J snows a (.;1U::;::; ::;C(.;l.!UH Ul anouier

sheetpile bulkhead in Yamanoshita wharf Con

struction of this wharf was completed about one

year before the earthquake. The earthquake re

sistance design of the wharf was carried out using

the design seismic coefficient of 0.12. As seen in

the figure, no appreciable damage was observed,

except for a local sinking of the fill behind the

anchor plate.

4.2.2 Sheetpile bulkheads

The 1964 Niigata earthquake

The majority of quaywalls in Niigata port were

sheetpile bulkheads. A typical damage of the

sheetpile bulkheads was their swelling and tilting

toward the sea. This type of damage was ob

served mostly in bulkheads with poor anchor re

sistance. In such cases, the swelling of bulkheads

was accompanied by a horizontal shear at a joint

of the top concrete and the upper end of sheet

piles.

A cross section of a sheetpile bulkhead in

Yamanoshita Revetment is shown in FigA.2A. A

characteristic feature of the damage was an over

all settlement. A face line of the walls swelled

more or less toward the sea and some of the top

concrete blocks sank completely under the water.

4-4

Tie Rod

FigA.2.5 Cross section of a sheetpile bulkhead

in Yamanoshita wharf in Niigata port

The 1968 Tokachi-oki earthquake

As shown in FigA.2.6, the Konakano No.1

quaywall in Hachinohe port was heavily damaged

by the earthquake. The walls tilted 5 degrees and

swelled toward the sea by O.6m at maximum due

to insufficient anchor resistance. Tension cracks

in the direction parallel to the face line and set-


- _. before- aIler

-5.0 VV

Tlrrber PileI

15.0

__ . [+250

::: 1

1\ \TIe Rod Tumbuclde

""s:-4.5 ~

]~

I~~L- 068.0

H.WL+~ ~itfLWL ±O.ro

-10.0- -1.lV..t.?;t--

FigA.2.8 Cross section of a sheetpile bulkhead

in Hanasaki port

The 1973 Nemuro-hanto-oki earthquake

As shown in FigA.2.8, the sheetpile bulkhead

was severely damaged by the earthquake. Ac

cording to the investigation after the earthquake,

the tie rods were not cut and the damage was es

timated to have been caused by the decrease of

anchoring capacity due to the seismic effect.

The 1983 Nipponkai-chubu earthquake

The severe damage occurred on the sheetpile

bulkhead at Ohama NO.2 wharf of -10m depth.

Typical features of damage in the quaywall were

a large settlement at the apron and a tilting of the

coping. Through the investigation after the earth

quake, the sheetpile damage was summarized as

shown in Fig.4.2.9. These damages were estimat

ed to be caused mainly by liquefaction of the

backfilling sand.

+2.0

LWL ±o.oo

+2.73

- - - before-- after

-7.00~

-12.00

-14.50

tlement in an order of several 10cm occurred in

the backfill surface. The maximum acceleration

of the earthquake was observed to be 0.26g in

this district.

- - - l:efore-after

FigA.2.6 Cross section of a sheetpile bulkheadin Konakano No.1 quaywaII in Hachinohe

port

The sheetpile bulkhead with batter anchor

piles, the quaywall of Kitahama pier in Hakodate

port, was damaged by the earthquake as shown in

FigA.2.7. The fixation point of sheetpiles and an

chorpiles was broken and the face line of the

quaywall swelled toward the sea by 59cm at

+3.00

HWL +1.04 ~ ,L WL ± O.00 "1\\~:;:Ll:u"

maximum.

Fig.4.2.7 Cross section of a sheetpile bulk

head in Kitahama pier in Hakodate portFig.4.2.9 Cross section of a sheetpile bulkhead

at Ohama No.2 pier in Akita port

4-5


4.2.3 Pile supported piers

The 1964 Niigata earthquake

The severe damage was observed on the tres

tle type quaywalls at Rinko district in Niigata

port. The ground having consisted of very loose

sandy alluvial layer, a typical feature of damage

in this area was a large settlement. The quaywall

shown in FigA.2.10 sank completely under the

water.

+2.40

l7 +0.00

-1.50

IIIIIIIIII,

FigA.2.10 Cross section of a trestle type pier

in B Berth in Niigata port

Composite breakwaters consisting of concrete

caissons and the foundation rubble in Kobe port

suffered damage as shown in Fig.4 .2.12. These

breakwaters were constructed on loose decom

posed granite, which was filled into the area after

the excavation of the original alluvial clay layer.

TIle breakwater settled about 1.4 to 2.6m through

the earthquake. The horizontal displacements of

the breakwater, however, were less than tens of

em. The mode of deformation suggests that the

caisson was pushed into the rubble foundation

and the rubble was also dragged down and

pushed into the loose deposit beneath it.

- - - before-after

-·16.7

FigA.2.11 Cross section and deformationlfailure of a pile supported pier at Kobeport

before-afler

Unit(m)

Clay

.... Backfill Soil, after c; /".... -!-'~s Excavating Clay Layer 'j'> /"

........... ///

......<::~::. -40.00.:::':'-;;':'-1-15}0- J

.g L.W.L

Clay

The 1995 Hyogoken-Nambu earthquake

A pile supported pier suffered damage at Ta

kahama wharf in Kobe port. The horizontal

residual displacement of the pier ranged from 1.3

to 1.7m. A typical example of the cross section

and deformation of the pile supported pier is

shown in FigA.2.ll. As shown in this figure, the

pier was constructed on a :firm foundation deposit

consisting of alternating layers of Pleistocene

clay and sandy gravel. The steel piles having a

diameter of 700mm buckled at the pile heads ex

cept for the piles located most landward. A crack

was observed at the connection of the pile cap

and the concrete beam located most landward.

4.2.4 Breakwaters

The 1995 Hyogoken-Nambu earthquake

Fig.4.2.12 Cross section and deformation of a

breakwater at Kobe port (Breakwater No.7)

4-6


4.3 Evaluation of Seismic Performance

4.3.1 General

In the design of port facilities, the effect ofearthquakes should be taken into account sothat they possess appropriate amount ofseismic resistance.

Explanation

(1) Earthquake resistant design should beapplied to port facilities as explained :in thischapter. Seismic resistance of bridges, oil

be examined based on other appropriateregulations and guidelines.(2) In the examination of seismic resistance,following factors should be taken into account.

(a) Seismicity of the region, target earthquakeand target ground motion.

(b) Subsoil conditions.(c) Importance of the facility, which should be

determined based on various factorsincluding it's role in the society or economy.

(d) Seismic resistance of the facility.(3) Following factors should be examined toassure the seismic resistance of the facility.(a) Stability of the whole structure.(b) Stability of the subsoil against failure.(c) Effects of liquefaction on the stability of

subsoil and upper structure.(d) Stress of the members of the structure.(e) Relative displacements between various

portions of a structure, between structuresor between structure and soil. This factormay be important for the purpose ofmaintaining the functions of the structureafter the earthquake.

(4) At Kobe Port, the type of structures werequite uniform during the 1995 HyogokenN anbu earthquake. This is why almost all ofthe structures suffered similar damage. If thetype of structures had been more diverse, theamount of damage for each structure should nothave been uniform because their responsecharacteristics should have been different.

4-7

Based on this experience, in the choice of thestructural type of port facilities, it isrecommended to adopt various type ofstructures as long as possible.

Related information

Seismic performance of port facilities shouldbe examined with pseudo-static method,earthquake response analysis and/or seismicdeformation method depending on the dynamiccharacteristics of the structure.

Seismic resistance of structures which arerelatively rigid and will not show muchamplification during earthquake, should be

design seismic coefficient designated in 4.4 and4.7. Gravity type quay wall is a typical exampleof such structure.

For structures which has a small dampingfactor and a natural period close topredominant period of ground motion or for thestructures which has a relatively long naturalperiod, modified pseudo-static method shouldbe applied, taking into account the dynamiccharacteristics of the structure. The applicationof modified pseudo-static method to the designof piled piers is explained in 4.8.

The seismic resistance of buried linestructures such as tunnels and pipelines shouldbe examined with seismic deformation methodbecause the safety of these structures arecontrolled by the deformation of surroundingsoil.

If the facility is especially important or thetype of structure is rare and there is no similarconventional structure, it is recommended thatit's seismic resistance should be examined byusing earthquake response analysis togetherwith . conventional pseudo-static method,modified pseudo-static method or seismicdeformation method. The earthquake responseanalysis should be based on appropriatemodeling of related conditions including thestructure and the earthquake.

4.3.2 Seismic performance requirement for portfacilities.


the importance of the liquefaction mitigation.

Seismic coefficient

(Level-Iground motion)

ISeismic coefficient -Regional seismic coefficient

X Factor for subsoil condition

x Importance factor

I

Cross section of the facility

IAssessment of liquefaction

and mitigation

I

Figure 4.3.1 Design process applied to

all port facilities

Detailed design

(2) During the 1995 Hyogoken-Nanbuearthquake, the lateral movement of the groundcaused significant damage to piles. In theearthquake resistant design of port structures,

the mitigation of liquefaction is always requiredwhen necessary. Therefore, it is only in verylimited case that that the liquefaction or relatedlateral movement of the ground is allowed and

used as a given condition of the design ofstructures. In these limited cases, the designshould be performed adequately based on theearthquake response analysis as a part of the

examination of earthquake resistance of soil

structure system or based on the references

regarding lateral movement of the ground.

Explanation

(1) Port facilities should sustain their structuralstability and maintain their functions for alevel-1 ground motion, which by definitionoccurs with high probability during thefacility's duration.

(2) High seismic resistant facilities, which areespecially important and require highseismic resistance, are allowed to suffer onlyslight damage for a level-2 ground motion,which by definition occurs with relativelylow probability during the facility's durationbut which is very intense. In other words,high seismic resistant facilities should beprepared for rapid restoration to sustaintheir intended functions after a level-2ground motion.

In the seismic design :of port structures, a

level 1 ground motion, which has a returnperiod of 75 years and a level 2 ground motion,

which is a ground motion due to intra-plateearthquake with a return period of more thanseveral hundred years or a ground motion dueto a subduction zone earthquake, should betaken into account.

High seismic resistant facilities include highseismic resistant quay walls, which arespecially designed for the transportation of

emergency cargo or for the maintenance ofeconomic or social activity just after theearthquake, and the revetments of the disasterprevention base, which is intended to keep thesafety of the citizen just after the earthquake.

While 'to maintain their functions' means tosustain their structural stability, 'to sustain

intended functions' means to suffer only a slightdamage and to be prepared for a immediaterestoration.

Related information

(1) Fig. 4.3.1 shows the design procedure

required for all of the port facilities. In this

procedure, after determining the structural

parameters, the evaluation of liquefaction

potential and the mitigation of liquefaction is

requested. This is based on the appreciation of

4.3.3 Pseudo-static method

4-8


(1) In principle, seismic load for portstructures with relatively short naturalperiod and relatively high damping' factorshould be designated as a design seismiccoefficient for pseudo-static approach. In thiscase, the design seismic coefficient designatedin 4.4 and 4.7 should be used. Seismic inertiaforce should be the larger of the following (a)and (b) and should be assumed to act on thegravity center of the structure.(a) (Seismic force)::::

(Self weight) x (Seismic coefficient)(b) (Seismic force)=(Self weight +

Surcharge) x (Seismic coefficient)(2) For structures for which pseudo-staticmethod is not applied, seismic load should bedesignated in an appropriate manner, takinginto account the characteristics of the

I structure.

Explanation

(1) For quay walls and other similar portstructures, pseudo-static method is applied asfor other wide range of structures1) • Becausenatural periods of these structures are generallyhigher than predominant periods of groundmotions, the response of these structures duringearthquake are similar to those of rigid bodieson a rigid table. In this case, it is assumed thatthe seismic load is proportional to the structure'sweight, The seismic coefficient is defined as theseismic load divided by the weight. In pseudostatic method it is assumed that the seismic loadacts as if it were a static load at the gravitycenter of the structure.(3) Because the seismic load is assumed to act asa static load in the pseudo-static method, it isnecessary to take into account the differencebetween the real phenomena and theassumptions in the method. To appreciate thisdifference, the safety factor and the allowablestress for dynamic loads are different from thosefor static load.(3) It is preferable to examine the seismicresistance of those structures which has a longernatural period compared to predominant periodsof ground motion or for which the distribution ofacceleration is not uniform along the height. Inthis case, seismic load should be assumed to be

the product of the weight of the portions of thestructure and the seismic coefficient of theparticular portion depending on the responsecharacteristics of the structure. In modifiedpseudo-static design of port structure, designseismic coefficient designated in 4.4 is used forthe calculation of seismic load. Therefore, theonly difference of modified pseudo-static methodcompared to original pseudo-static method liesin the computation of the distribution of seismiccoefficient along the height of the structure. See4.8 for details of the modified pseudo-staticmethod for piled piers.(4) The effect of the vertical component of groundmotion depends on the type of structure and on

strict to consider vertical seismic coefficient, thevertical seismic coefficient is not required to beconsidered in the design code because offollowing reasons. First, it is preferable to avoidthe complexity of the computation. Second,according to the observation of ground motion,the vertical component is usually smaller thanthe horizontal component except for near-sourceregion. Thirdly, the horizontal design seismiccoefficient designated in 4.4 includes the effect ofvertical seismic ground motion-.Because of thesereasons, the consideration of horizontal designseismic coefficient is sufficient for the design ofusual port structures.

4.3.4 Earthquake response analysis

If the facility is especially important or thetype of structure is rare and there is nosimilar conventional structure, it isrecommended that it's seismic resistanceshould be examined by using earthquakeresponse analysis together with conventionalpseudo-static method or seismic deformationmethod.

Explanation

(1) General explanation

4-9


Recently, new type of port facilities orextremely large port facilities have beendesigned and constructed. On the other hand, itis sometimes required to construct portstructures at a site with a poor subsoilconditions. Furthermore, as explained in 4.7, itis requested to examine whether a high seismicresistant facility will maintain their functionsafter a near-source ground motion such as theground motion at Kobe Port during the 1995

Hyogoken-Nanbu earthquake. It isrecommended to examine earthquake resistanceof structures-by conducting earthquake responseanalysis to understand the performance ofstructures during earthquake more precisely jf-LL _ .L- .f! _.J.. ~~_~_ :. ~_~ :..J'! .L-l- __ .L-..._w ...... J... ... .; ....Lilt: LYIJl::: UJ. ::;L.I:UCLUJ.-t: L:5 ilew U.I: .ll I,llt: ::;L.l:U{.;l>UJ.1:: -'-'"

especially important.

(2) Implementation of earthquake responseanalysis.

When earthquake response analysis isconducted, first, appropriate method should beselected. Then the structure should be modeledfor that particular method and the materialproperties should be determined. Furthermore,input ground motion should be determined. Thevalidity of the results should be examinedcarefully in the light of the limitation of themethod, the limitation of the modeling and theaccuracy of the material property;

Related Information

(1)Input ground motion

(a) In the design of structures, it isrecommended to determine input ground motionbased on past observations or earthquakeresponse analysis of the ground. Strong motionrecords at Japanese ports have beenaccumulated and published since 19632

).

(b) For determining peak amplitude andwaveform of the ground motion, the size and thesource-mechanism of the earthquake, thedistance and the site effects should be taken intoaccount.(c) When past records are used, if the soil

conditions at the observation site and theconstruction site are different, the surfacerecords should be deconvolved to obtain incidentwave at the bedrock, which can be used as aincident wave to compute surface motion at theconstruction site. For this process, responseanalysis of the ground based on multiplereflection theory can be used. Multiple reflectionmethod, however, is based on equivalent-lineartheory. Therefore, the method can be appliedonly when the shear strain in the soil is lessthan 1%. It should be noted that if the targetground motion is of level-2, the method is notapplicable in many cases.(d) To determine the peak amplitude or the~.~••_-J _~-<-:~_ .f'~~...~..~ ..1-.~ .. ~.~_~ ~..~.;.~~ ~~ th\~.!l.uu.uu .1..llV W.Vll , .!.a.\';iJUJ..~ ltL!.aL vv,ac; ~LoGUJtJ\A. J.L.I. "oJ!

should be considered. Following equation hasbeen presented to estimate peak amplitude ofground motion at engineering-orientedbedrock",

Log l oAcoR=O.55M-loglO(X+O.0050 X lOD.05M)

- O.00122X+O.502. (4.3.1)

Log l A MAC=O.53M-loglO<X+O.0062 X lOo.53~

- O.00169X+O.524. (4.3.2)

LoglOV=0.48M-loglO<X+O.014 X 100.43M)

- O.00060X-O.324. (4.3.3)

Here, AcOR is the corrected peak groundacceleration (Gal), ASMAC is the peak groundacceleration measured with SMAC-t-ypeaccelerograph (Gal), V is the peak groundvelocity (kine), M is the magnitude, X is theclosest distance from the fault to the site (km).

Strong motion observations. at Japanese portshave been conducted with SMAC-type and ERStype accelerographs, Because of the differentcharacteristics of these types of seismographs,they give different waveforms. SMAC-typeaccelerograph gives smaller peak groundacceleration. Therefore accelerograms fromthese two different manner should be treated ina different way. In the standard process of the

4-10


strong motion observation at Japanese ports,records are corrected for their characteristicsof the accelerographs and published as"corrected records". Corrected records can betreated in the same way irrespective of the typeof accelerographs. This is the reason whyusually corrected records are used in theearthquake response analysis.

Although almost all of the accelerographs atJapanese ports are of ERS-type, most of the pastearthquake records were obtained by SMACtype accelerographs. Therefore, past researchwere based on the SMAC records. Because thedetermination of design seismic coefficient ofhigh seismic resistant quay walls and

given in FigA.3.2 and expressed in EqA.3A.

Log1oA=3.159+0.234M- l.4781oglDX. (4.3.4)

Here, A is the peak ground acceleration ofengineering-oriented bedrock (Gal), M is themagnitude, X is the shortest distance from thesite to the fault (km).

(2)Methods for earthquake response analysis

Methods for earthquake response analysis canbe divided into two categories, that is, numericalanalysis with computers and vibration tests.

Figure 4.3.2 Attenuation relations for

Pacific side of Tohoku

500 1000

M :r6.0

It J.iI:: 7. .:1.

oM' 7.2x M>6.7.c. M-6.5-6.0a M = .5.9_5..5• M·5.4-~.O

.. M ::4.9-4,0

( I , I I I Ii100

Fault distance [km]50

500

1000.-

100..-.--;

,,~Q.<:00.-

J5~J10

(a) Numerical analysisTable 4.3.1 shows various methods for numericalearthquake response analysis.1) Effective stress analysis and total stress

analysisWhen the soil is liquefied, the pore water

pressure is induced in the soil and the effectivestress decreases. As a result, rigidity anddamping of the soil change. Effective stressanalysis can treat these situations and the

liquefaction assesment is based onresearch, SMAC PGA is used for theseexaminations.(e) Most commonly used seismograms in thedesign of port structures are shown in FigA.7.3.These waveforms, however, are equivalent toSMAC-type accelerographs. In the earthquakeresponse analysis, corrected waveforms shouldbe used.(f) Alley observations of strong motion have beenconducted, which are useful in measuring the

. strain of the ground during earthquake. Ingeneral strong motion accelerations are directlyobserved and the displacement can be obtainedby integrating the records. This integrationoften fails in error because of the error duringthe digitization. A method to avoid the errorduring the integration was presented.Displacement waveforms based on this methodare displayed in reference 2).(g) In pseudo-static design, the verticalcomponent of ground motion is usually neglected.In the earthquake response analysis, however,vertical component of ground motion should besometimes taken into account. The peak verticalground motion divided by peak horizontalground motion usually ranges between 1/31/23),4) .

(11) According to the examination of past strongmotion records, the attenuation of peak groundacceleration is dependent on the region. Theregional attenuation was examined for thePacific coast of Tohoku area. The results are

4-11


Table 4.3.1 Methods of numerical earthquake response analysis

Treatment of excess I Effective stress Total stresspore water pressurei

Dimension I I-D, 2-D. 3-D

I ,Modeling i

Multiple reflection model. :MDQIi'. FEMII

Material Linear. Equivalent-linear, Nonlinear

Domain Time domain, Frequency domain

'"

O~ 07I

05I

I'J

0.3,Time

01 0.2

-liH"~i (rnJ (kN/JJI") j fmJ'$J i '!='~======S=4;=C:;:;~=::::::~==-. 6 ! 19.5 i12~-Ii I 4119.3 "Ii 200 I: I! I I 'I t a ! 16.1 '~O!

J

efl : : ;1 14 JO.7! '20 i, I I

! i iI " 1 1a.5 I '20 I

I .

!

Ir(a) Material properties (b) Reflection and transmission

Figure 4.3.3 Multiple reflection model

4-12


excess pore water 'pressure in the soil can becomputed directly, On the other hand, in totalstress analysis the pore water pressure is notcomputed and the effect of pore water pressureon the response is not considered. Therefore, inthe case of high pore water pressure (the porewater pressure ratio of 0.5 or greater), theresults of total stress analysis is not accurate. Inthe practical design, however, total stressanalysis is often utilized because of it'ssimplicity. In many cases, stress or accelerationgiven by effective stress analysis is smaller thanthose given by total stress analysis.2) Dimension of the domainThe dimension ranges from 1 to 3. In general theresponse of horizontally layered soil is treated asa I-D problem. On the other hand, structure-soilsystem such as quay walls which satisfiesplane-strain condition is 'treated by 2-D analysis.Although there are some cases in which 3-D

analysis is more appropriate, 3-D analysis ismainly used for especially important structuresor research purposes because of the limitationsof computers.3) Modes for computation

a) Multiple reflection modelIn this model, the soil layers are consideredto be horizontally homogeneous as shown inFigA.3.3 and vertical incidence of a shearwave is assumed. In this method, stressstrain relation is usually assumed to beequivalent-linear. SHAKE5) adopts thisalgorithm.b) MDOF modelIn this model, the soil is considered to be acombination of masses, springs and dampersas shown :in. Fig.4.3A. The algorithm of thismethod is simple. It is also possible toconsider nonlinear restoring force.c) FEM, etc.

In this model, the soil is divided mto finiteelements as shown in FigA.3.5. This methodis applicable not only soils but also manytypes of materials. The main feature of thismethod is that the 2-D characteristics of thesoil is easily taken into account. Practicalprogram for this method includes FLUSH6

),

BEAD?) and FLIPS).

d) Cantilever model

Structure (or soil) is modeled as a cantileverwith a constant or linearly varying materialproperties. Shear beam model is mostcommonly used. Information regarding theshape of the structure, density, rigidity anddamping is necessary for computation.

4) Evaluation of material properties.In the earthquake response analysis, modelingof the soil nonlinearity is also important. In thelow strain range, stress-strain relation of the soilis linear. In the middle or high strain range,however, this relation is nonlinear, In such cases,nonlinearity of the soil has to be taken intoaccount. Today, besides linear analysis,equivalent-linear analysis, which uses materialparameters corresponding to the level of strain,and nonlinear analysis, which reproduces actualstress-strain relation at large strain to someextent, have been developed and used. The effectof the deference of modeling among theseanalysis can be summarized as follows. Fig.4.3.6 shows the comparison among linear,equivalent-linear (SHAKE), Bi-linear, Tri-linear,Hardin-Drnevich and Ramberg-Osgood modelsas applied to I-D soil response problem in whichthe amplitude of input ground motion is 100Gal.G/Go- 'l' and h- 'Y curves from these models areset to be consistent with experimental results atthe strain level of 0.3%. The figure shows thedistribution of peak acceleration, peak stressand peak strain. In this range of the inputmotion amplitude, the models which considerssoil-nonlinearlity has a tendency to give smallerresponse acceleration / stress and larger strain.Difference between equivalent-linear model andnonlinear model are small as long as the peakstrain is less than 1%. In the case of level-2ground motion, however, peak shear strain oftenexceeds 1%. At our current state of knowledge itis difficult to determine whether abovementioned tendency applies to' such cases.

In the earthquake response analysis, G/Go- "I .

and h- 'Y curves from the models should beconsistent with the results of experiment at thestrain level of concern. In general, masing-rule

is used for representing the stress-strain loop. Ithas been revealed that this rule overestimates

4-13


HVs {ml si

37 , '5.6

M 5

M 4 InM 3 K4

M 2 K 3

M I K2

~::!:~C!!/){'------=-I

~=:::::j

I

I 7.5 /6.7

(a) Material properties (b) Multiple degree of freedom system

Figure 4.3.4 Multiple degree of freedom (J\1DOF) system

Cohesive soil

(.!jilli : rn)

Sand drain

2-l.lY./

7-:tlJXl-3S.0':

30.50

Basement of crane Backfill soilGravel /'"1.< I , v-6.0

Gravel ~=='------7

11.J

Replaced soil

9-14.50,,;/1 l;

v-IH.sn

v H.W.L+1.7m'¥¥ LW.L""*"O.om

Cohesive soil

(a)"Cross section of the target structure

Inclination4.1"

\l Horizontal displacement3.5m

\'ertical displacement I 5m

, ,I ,I I I I

T ............. ...,..."'T'"-r ~ ..... _ J +4.0m

-!--'_ -H-\-;i..(:· eJ.:I.:.l''- ~U X t______ • __ L~

...... ,._ _ _ _ ....... _to .. _ "_.1

- --- .. -"'-- -~~-- - - - - - -...- ~ ..", I

,, ,

i"I ,

I I I ' iI i I I I iI I 1 I ; ; I

I ,

I I ;

Iii iI I ;

(b) Results of FEM analysis'

Figure 4.3.5 FEM analysis

4-14


Cl mcx =' 100Gai

S - 252 NS Bose

4000--QLin0---08- LLr--6T- LV---'V H - Do-'~R - 0--SHAKE

300200

Acceler-ation a (Gal)

100, r~ ~

l ~P'" <, i'\',B ..------;-- i

I

1'1)'/ Ii 'x -'- I> , :

i..: T~ I ,

'1. ,/ / II

I: 'I 1\'-"<., !\ I

% . ;

j),o'

>!::;l>,.u

'" Vi\I

" I

:Ii I, i I

a12.3

-73.6

E

o

S-252 NS Base

Omex = IOOGal

1.5o--oLino--oS - Lis--'£:' T - Ly--"7 H - D<>---¢ R - 0--SHAKE

Shear strain y (%)

05 1.0'$>-.

0 -0 --~

~~---=-~-~ iI

~.,---:~ i !. "",4 i

~ q !

~ 9 ii ;. I

q71 Lf.~ !I

JI: :

~ C! -,!WoP i

;)#>'1\ !

~.->J~ ,

If' i,J:r !

Iti'f i i

12.3

-73.6

Shear stress -r: (Him')

s- 252 NS Base

Omex ='IOOGol

0--0 Lino--oS-LLr---6. T - Lv---v H - Do-.-¢ R - a-- SHAKE

5040302010a;:a... o., , I I~ -c; I

-\'R~ .~ ! I

.~-~ ~'::J

~ I!

c>- .i\\ "",- '<f ( :;.;0 ,

I .,~ ....- il;;:(, L>: ?J;~ '-[ -, ii f'.. i

~ 1"< ~=t ? I'-< \7

__ 0

I: QI \ItS ~~"" I

<s. \Zl' u.:.!

I ~>~ L~i

i ~L:,.+-\ I 'Qy

i , «~\I \----=-~ --¥! ~ -. , 0

12.30 0

-73.6

Figure 4.3.6 Comparison ofthe modeling of soil noninearitv (l)

4-15


the damping in the case oflarge strain. Today anapproach to mitigate this discrepancy has beenpresented, in which the thickness of stressstrain loop is controlled to give more realisticdamping factors". FLIP is one of the programswhich use this approach.

Fig. 4.3.7 compares the results of SHAKE,DESRA (hyperbolic model) and CHARSOIL(Ramberg-Osgood model) for the same groundmotion (El Centro-1940NS, PGA=O.lG) and forthe same soil layers'?', SHAKE gives largersurface acceleration and shear stress. Todaythere is a consensus to think that equivalentlinear analysis such as SHAKE gives safe-sideresults, although the situation will depend onthe soil and ground motion characteristics.5) TIme domain and frequency domain.

Nonlinear analysis including effective stressanalysis is usually implemented in time domain.If the excess pore water pressure is small (porewater pressure ratio less than 0.5), effectivestress analysis may be implemented infrequency domain in a similar manner asequivalent-linear analysis.6) Effect of water

In FEM programs, the effect of water shouldbe properly taken into account by using fluidelement. For example, FLIP has a fluid elementby which sea water can be treated as a noncompressional fluid.

(b) Vibration testsThis is a test in which model soil-structure

system is subject to ground motion. This is aconvenient method to understand the globalperformance of soil-structure system. High skill.is required, however, to conduct vibration tests.Vibration tests include log shaking table tests,centrifuge tests and in-situ vibration tests.

(3)FOl' dynamic characteristics of the structure(vibration mode, natural period and damping), itis convenient to refer to the results of in-situmeasurement and/or numerical analysis.

(4) Earthquake response analysis requires theevaluation of nonlinear material properties.

Effective stress analysis is a method to considerthis nonlinear properties relatively accurately.Now effective stress analysis has been proved tobe a efficient method to evaluate seismicperformance of structure including residualdeformation and residual stress. On the otherhand, equivalent-linear analysis has been usedwidely because of it's simplicity, Materialproperties for this analysis should bedetermined by conducting experiment or byreferring to past analysis.

(5) The effect of water should be taken intoaccount if the structure has an interface withwater.

(6) Sometimes large and temporal responseappears in the results of earthquake responseanalysis. These phenomena can be evaluated byreferring to the past design of similar structureor related research results.

4.3.5 Seismic deformation method

Because the deformation of line structuresetc. buried in the soil during earthquake iscontrolled by surrounding soil, it is preferablethat such structures should be designed byusing seismic deformation method.

Explanation

(1)In the examination of earthquake resistanceof line structures such as tunnels or oilpipelines, the relative displacement of theground is important. The relativedisplacement is dependent on thecharacteristics of ground motion and the soilconditions.

(2) Besides line structures buried in the soil,seismic deformation method has been appliedto dams. Seismic deformation method can beapplied to structures other than linestructures as long as the residualdisplacement of the structure can beappropriately evaluated.

4-16


Period (,)

'mol (kg/lem')

jC 01 0.2 03 0.4 05

'"~"-,'C, CHARSOIL J",5 -,- ,', !I \\ /SHAKE

I

-= \\ [I

J \\

~I

~. ».~ 10DESRA/ \,

\\

I2 \\\ \,

15 "

Figure 4.3.7 Comparison of the modeling of soil nonlinearity (2)

References ........nnnntOl"...... "' .......... 1:-' .................... of

1) "Earthquake Resistant Design of CivilEngineering Structures" by N. Mononobe,1952 (Revised Edition in Japanese).

2) "Annual Report on Strong-motionEarthquake Records In Japanese Ports(1995 & 1996) ,i by Yukihiro Sato, Koji Ichii,Susumu Iai, Yuko Hoshino, Yoko Sato,Masafumi Miyata and Toshikazu Morita,Technical Notes of the Port and HarbourResearch Institute, No.909, 1998 (in

Japanese with English abstract).3) "Relation between Seismic Coefficient and

Peak Ground Acceleration Estimated fromAttenuation Relations" by A Nozu, T. Uwabe,Y. Sato and T. Shinozawa, Technical Note ofthe Port and Harbour Research Institute,No.893, 1997 (In Japanese with Englishabstract).

4) "Characteristics of Vertical Components ofStrong Motion Accelerograms and Effects ofVertical Ground Motion on Stability ofGravity-type Quay wall" by T. Uwabe, S.

Noda and E. Kurata, Report of the Port andHarbour Eesearcb Institute, Vol. IS, No.2,197G.

5) "SHAKE-A Computer Program for

Earthquake Response Analysis ofHorizontally Layered Sites" by P.B. Schnabel.

J. Lysmer and H.B. Seed, Report No. EERC

72-12. Col. of Eng., University of Californiaat Berkeley, December 1972.

4-17

Approximate 3-D analysis of Soil-StructureInteraction Problems" by J. Lysmer, T.Udaka, C.F. Tsai and H.B.Seed, ReportNo.EERC 75-30, University of California atBerkley, 1975.

7) "Coupled Hydrodynamic Response

Characteristics and Water Pressures ofLarge Composite Breakwater" By T. Uwabe,S. Noda, T. Chiba and N. Higaki, Report ofthe Port and Harbour Research Institute,Vo1.20, No.4, 1981 (in Japanese with English

abstract).8) "Strain Space Plasticity Model for Cyclic

Mobility" by S. Iai, Y. Matsunaga and T.Kameoka, SOlIs and Foundations, Vo1.32,

No.2, pp.1-15.9) "Modeling of Stress-Strain Relations of Soils

in Cyclic Loading" by K. Ishihara, N.Yoshida and S. Tsujino, Proceedings of the{fh Conference on Numerical Methods inGeomechenics, Nagoya, Vol.L, 1985, pp.373

380.10) "Comparison of Dynamic Analysis for

Saturated Sands" by W.D.L. Finn, G.R.Martin and M.K.W. Lee, Proc. of ABeLEarthquake Engineering and Soil Dynamics,

VoLI. pp.472-491, 1978.


4.4 Earthquake Load

4.4.1 Design Seismic coefficient

(1) For pseudo-static design of port structures, horizontal design seismic coefficient should bedetermined with following equation.

Seismic coefficient =Regional seismic coefficient.x Factor for subsoil condition X Importance factor(4.4.1)

Horizontal design seismic coefficient should be rounded to obtain two places of decimals. Standardvalues for regional seismic coefficient are:ReID-on A: 0.15Nemuro, Kushiro, 'Iokachi and Hidaka districts of Hokkaido, Saitama, Chiba, Tokyo (Except forHachijo and Ogasawara Islands), Kanagawa, Yamanashi, Shizuoka, Aichi, Gnu, Fukui, Shiga, Mie,Nara, Wakayama, Osaka and Hyogo.Recion B: 0.13Pacific side of Aomori, Iwate, Miyagi, Fukushima, Ibaragi, Tochigi, Gunma, Nagano, Kyoto, Kochiand Tokushima.Region C: 0.12Iburi, Oshima and Hiyama districts of Hokkaido, Aomori (except for Pacific side), Nata, Yamagata,Niigata, Toyama, Ishikawa, Tottori, Hiroshima, Ehime, Oita, Miyazaki, Amami Islands ofKagoshima and Kumamoto.Region D: 0.11Abashiri, Goshi, Ishikari, Sorachi, Rumoi and Kamikawa districts of Hokkaido, Okayama, 'Iottori,Kagawa, Nagasaki (except for Goto, Iki and Tsushima Islands), Saga, Kagoshima (except for AmamiIslands) and Okinawa (except for Daito Islands).Region E: 0.08Sorachi district of Hokkaido, Hachijo and Ogasawara Islands of Tokyo, Yamaguchi, Fukuoka, Goto,Ik:i. and Tsushima Islands of Nagasaki and Daito Islands of Okinawa.

Factor for subsoil condition should be determined as shown in Table 4.4.1 and 4.4.2.

Table 4.4.1 Factor for subsoil condition==--====--===--=====Classification l"t kind

Factor 0.8 1.0 1.2

Table 4.4.2 Classification of subsoil

Thickness ofQuaternaryDeposit

less than 5m5-25mmore than 25m

GravelSand

orclay

4-18

Soft ground


Importance factor should be determined according to Table 4.4.3.

Table 4.4.3 Importance factor

Category

Factor

Special

1.5

A

1.2

B

1.0

C

0.8========--===--===

Category Special: The structure has significant characteristics described by items (1)~(4) ofcategory A.Categorv A: (1) If the structure is damaged by an earthquake, a large number of human life andproperty will possibly be lost. (2) If the structure is damaged by an earthquake, economic or socialactivity of the region will be severely suffered. (3) The structure will perform an important role inthe reconstruction work of the region after the earthquake. (4) The structure handles a hazardousor a dangerous object and it is anticipated that the damage of the structure will cause a great lossof human life or property. (5) If the structure is damaged, it is supposed that the repair work isconsiderably difficult.Category B: The structure does not belong to categories Special, Anor C.Category C: The structure does not belong category Special nor A and is easy to repair or, even ifthe structure is damaged by an earthquake, the effect on economic or social activity is small.

(2) If vertical seismic coefficient is required in the pseudo-static design, the vertical seismiccoefficient should be determined appropriately, taking into account the characteristics of sn..uctureand subsoil.

Explanation

(1) In general, factors that has to be consideredin the determination of design seismiccoefficient are regional seismicity, subsoilconditions, dynamic characteristics of thestructure and the importance of the structure.Most of the port structures, however, haverelatively short natural period and relativelylarge damping factor. Therefore, in general, thedesign seismic coefficient for pseudo-staticdesign is determined without considering thedynamic characteristics of the structure.(2) When the design seismic coefficient can beaccurately determined by investigating regionalseismic activity; characteristics of groundmotion, site response, etc., it is preferable to usethis design seismic coefficient instead of thevalue designated here. For example, when thedesign ground motion is determined based onthe information regarding regional seismicactivities or based on strong ground motion

4-19

observations or when seismic response analysisof the structure is conducted, design seismiccoefficient can be determined based on theseresults.(3) Th determine importance factor of thestructure, it is necessary to consider not onlythe purpose, type or size of the structure butalso social or economic aspects of the structure.Following factors also should be taken intoaccount.

1. The extent of damage in the futureearthquake, the difficulty of restorationwork 01' the residual strength after theearthquake.

2. The cargo-handling capacity of theother facilities of the same port.

Therefore, it is possible to use differentimportance factors for the structures for thesame cargo in the same port when desired.

(4) When computing seismic load, it is notallowed to subtract buoyancy from the weightof the structure. In the computation of soil


pressure, however, unit weight is usuallymodified to include the effect of buoyancy.Therefore, in the computation of soil pressure,apparent seismic coefficient should be used asdescribes in 4.5.(5) For structures other than high seismicresistant quay walls, the upper limit of designseismic coefficient should be 0.25 for severalreasons. First, in the past, the upper limit ofdesign seismic coefficient was 0.25. Second,there has been no port structures with designseismic coefficient of 0.25 that sufferedsignificant damage. Thirdly, high seismicresistant quay wall has been constructed inmany ports.

Related information

(1) Design seismic coefficient for modifiedpseudo-static method should be determinedbased on the response characteristics, responsespectrum of ground motion and the relationbetween response acceleration and designseismic coefficient. The dynamic characteristicsof the structure can be obtained byappropriately modeling the structure asdescribed in 4.3.4. Response spectrum of groundmotion can be obtained by appropriatelymodeling earthquakes or by averaging observedresponse spectra. By using these responsespectra, peak response accelerationcorresponding to natural periods offundamental and higher orders can be obtained.By superposing these peak accelerations,response acceleration can be obtained, fromwhich design seismic coefficient is determined.Observed acceleration response spectra fromstrong motion observations in Japanese portsare displayed in reference 3). For the purpose ofassessing dynamic characteristics of thestructure, not only the structure itself but alsosoil and water surrounding the structure haveto be appropriately modeled. Also thedissipation of vibration energy ha-ve to beconsidered appropriately.(2) The relation between seismic coefficient andpeak ground acceleration1),2) for gravity quay

walls is shown in FigAA.l, in which seismic

4-20

coefficient was obtained based on past quaywall damage and peak ground acceleration was

obtained from either observation or attenuationrelations. For sheet-pile quay walls, similarresult have been obtainedv". Application of theresults to other structures require prudentexaminations. By the way, peak groundacceleration in FigA.4.1 is a value obtainedwith SMAC-type accelerograph. Peak groundacceleration which is obtained with anothertype of accelerograph should be converted tothat of SMAC-type accelerograph beforecomparison.(3) Level-l ground motion for all port facilities(a) Regional seismic coefficient has been..J~..~.~~_~A .t'..~_ ...l..~ ...:l~~.....;'l......h~_ N{: _~_1~uC:;lIt:a,U1...U..J,C;U .LLV.lll VUC \Lli)lo.l.~JJu.w..U.1.L VJ. VCQ..fi.

ground acceleration with a return period of 75years2),5) . Here, return period is defined in a

probabilistic way and it does not necessarilyimply that the particular ground motion occursevery 75 years. For example, a structure with aduration of 50 years encounters a groundmotion with a return period of 75 year or longerwith a probability of approximately 50%. Forstructures with a shorter duration, it may bereasonable to reduce the return period of designground motion.(b) Table 4.4.4 shows peak ground accelerationwith a return period of 75 years. Regionalseismic coefficients have been obtained fromaveraged relation between seismic coefficientand peak g-round acceleration.

Reference

1) "Relation Between Seismic Coefficient andGround Acceleration for Gravity Quay Wall" S.Noda, T. Uwabe and T. Chiba, Report of theP01"t and Harbour Research Institute, Vol.14,

No.4, 1975 (in Japanese with English abstract).2) "Relation between Seismic Coefficient andPeak Ground Acceleration Estimated fromAttenuation Relations" by A Nozu, T. Uwabe, Y:Sato and T. Shinozawa, Technical Note of thePort and Harbour Research Institute, 1'10.893,1997 (In Japanese with English abstract),3) Annual Report on Strong-motion EarthquakeRecords In Japanese Ports (1995 & 1996) " by


Yukihiro Sato, Koji Ichii, Susumu Iai, YukoHoshino, Yoko Sato, Masafumi Miyata,Toshikazu Morita, Technical Note of tile Portand Harbour Research Institute, No.909, 1998(in Japanese with English abstract).4) "Analysis on Seismic Damage in AnchoredSheet-Piling Bulkheads" by S. Kitajima and T.Uwabe, Report of the Port and HarbourResearch Institute, Vol.18, No.1, 1979 (inJapanese with English abstract).5) "Expected values of Maximum Base RockAcceleration along Coasts of Japan" by S.Kitazawa, T. Uwabe and N. Higaki, TechnicalNote of the Port and Harbour ResearchInstitute, No.486, 1984 (in Japanese withEnglish abstract).

Table 4.4.4 Regional seismic coefficient and

.peak acceleration with a return

'period of 75 years

Peak groundRegional acceleration with

Area seismic return period of,

coefficient 75 years(Gal)

A 0.15 350B 0.13 250C 0.12 200D 0.11 ISOE 0.08 100

). 3 0 ";---~--- -- - ~--- --- -:-----~--.:.-----------r-----·-----------·-r---------- -------- ---1-------":'-----------1------- -------- -----7Ii Vertical bar indicates the estimated range ~f seismic coefficient : ,

.------- --- ---- -------7---- -------.-- ----i-------- ------:v--L-: TTL 'if: 6: ---------··-·~l~-H A;- )f.------:

0" '----------f~y ---:- 1_ ~t~--- f--~~-J------------------L------~i0" i---------,--t ·:-~~~:l)t-±+---T------------L----f

. . : . .: L V Seismic coefficient estimated for ports

0,00 . ----•._-----------.,--------- • ; __• •. --1 ..__..; --:

o 100 200 300

ASMAC (Gal)

400 500 GOD

Figure 4.4.1 Relation between peak. ground acceleration and seismic coefficient

4-21


4.5.1 General

In static conditions, earth pressures are calculated by

ordinary More-Coulomb earth pressure theory. On the

other hand, earthpressures during earthquakes are

calculated by the Mononobe/Okabe method (Mononobe

1957, Okabe 1924) with special treatment where beneath

the water table soil layers.

4.5 Lateral earthpressure

pressure during earthquake

and water In the case ·of using equation(4.5.2), it should be

included the dynamic water pressure during earthquake

when overall seismic stability calculation. The dynamic

water pressure is applied in the seaward direction.

Refernces:

Mononobe, N.,'Emthquake resistant design of civil

engineering structures',(Revised edition), 1957.

Okabe, S.,'GeneraI theory on earthpressure and seismic

stablility of retaining walls and dams',J.

JSCE, Vol.] O,No.6, 1924.

4.5.3 Dynamic water pressure during earthquake

In case of sea walls during an earthquake, external

forces can be summarized as shown in Figure(4.5.]).

4.5.2 Apparent seismic coefficient (Seismic

coefficient of submerged soil layer)

The concept of the apparentseismic coefficient k' is

indicated in following equation.

4.6 Liquefaction Prediction/Determination

Method

4.6.1 General

Saturated loose sandy deposits tend to liquefy during

earthquakes, causing damage to structures. Currently,

liquefaction phenomenon is a major keyword for seismic

design of port and harbor facilities. Past big earthquake

disaster reports show that liquefaction should be taken

into consideration in design and construction of

structures. Liquefaction potential should be assessed by

two step procedures as follow with considering the

condition of construction site, a degree of importance,

etc ..

(J ).Grain Size Distribution and SPT- N value

If the results obtained by (I) is not sufficient, following

procedure should be conducted.

(2).Undrained Cyclic Triaxial Test and seismic response

analysis

4.6.2 Grain Size Distribution and SPT-N value

A soil is classified according to the grain size

distribution by Fig.(4.6.1). The soil of which gr.ain size

distribution falls outside of the liquefaction possibility

zone in Fig.(4.6.1) is considered non liquefiable. For

the soil of which the distribution curve falls inside the

liquefaction possibility zone the following procedure is

conducted using standard penetration test blow

counts(SPT N value).

(l).Equivalent N Value

An equivalent N value is calculated by the following

equation.

(4.5.1)

DynamicEarth Press.

y,xk=(y-]O)xk'

DynamicWater Press.

A product of unit weight of a soil layer and seismic

coefficient over the water table equals a product of

submerged unit weight of a soil layer and the apparent

coefficient.

The apparent seismic coefficientofsubmerged soil layer

can be evaluated by equation (4.5.2).

Where: Y, =unit weight of a soil layer over the water

table (kl-l/rrr'), y =unit weight of saturated soil layer at

submerged area(kN/rri\ m=uniform external load at the

ground surfacefkl-l/rrl), h =thickness of arbitrary soil

layer(m), suffix i=over the water table and suffix

j=submerged area.

Fig. 4.5.1 Schematic diagram of external forces(N ) _ N - 0.0]9(d v -65)

65 - 0.0041(d v -65)+ 1.0(4.6.1)

4-22


Where: (N)65 =Equivalent N value, N =SPT N value of

a soil layer, (J' l' = effective overburden pressure of a

soil layer (kN/m") (The effective overburden pressure

should be calculated with respect to the ground surface

elevation at the time ofthe standard penetration test).

IV

/! III

/1/..V J

/;~V

~5

25

Fig. 4.6.2 Classification of soil layer for liquefaction

prediction based 011 equivalent acceleration and

eq uivalent N-values.

00 100 200 300 400 500 600

EQUIVALENT ACCELERATION (Gal)

30

UJ:::J:i 20>ZJ-Z I.,UJ-l«:2::::J 10C!UJ

gives a critical N value ofa soil under a given equivalent

acceleration.

Zone 1 has a very high possibility of liquefaction.

Zone n has a high possibility of liquefaction.

Zone ill has a low possibility of liquefaction.

Zone N has a very low possibility ofliquefaction.

10

10

GRAVEL

GRAVEL

2.0

2.0

I SAND

I SAND

0.1 1.0GRAIN SIZE (111m)

OJ 1.0GRAIN SIZE (rnrn)

0.Q75

0.075SILT

SILT

0.01

D.W5

0.005

CLAY I

CLAY I

~f- SAND WITH HIGH COEFFICIENTOF UNIFORMITY Uc>3.5:c I00,-----,-----=----,...----:--.---,,---oiii3 75>-'"B} 50zr;:lJJ 25o~15 o''-'----:-':-.,-----;!-;------:-'-:-------:!-;:---uc:<:uJc,

;jf- SAND WITH LOW COEFFICIENT OF UNIFORMITY Ucd.5:c100,----,-----,---,---:---.---::------:--,-oUi3 75>-'"ffi 50zr;:uJ 25c-cf-iii 0L-.---:"':-:---~-7:---___:_"::_------,0:_-

Uc:<:lJJc,

Fig. 4.6. J G radation of soil having the possibility of

liquefaction.

(2).Equivalent acceleration

An equivalent acceleration is estimated by the following

equation based on the maximum shear stress obtained

from earthquake response analysis.

(4.6.2)

(4).Correction of the equivalent N value (The fine

content «O.075mm) ofa soil is not less than 5%)

The equivalent N value of a soil of which the fine

content is not less than 5% is corrected as in the

following three cases:

Casel:The plasticity index ofa soil is less than 10 or

the fine content is less than 15%. An equivalent N value .

obtained from Eq.(4.6.1) is corrected by the following

equation.

where: (N)65corrected = a correctedequivalent N value, cN

= a correction factor obtained from, Fig.(4.6.3) based on

the fine content. The corrected equivalent N value is

plotted in Fig.(4.6.2) with an equivalent acceleration and

the zone to which a soil layer belongs is determined.

Case2: The plasticity index is not less than 10 and not

more than 20, and the fine content is not less than·15%.

Two corrected equivalent N values are calculated as

follows:

Where: (Xeq = equivalent acceleration, rroax = maximum

shear stress (kN/m"), (J'" = effectiveoverburden pressure

(kN/m2) (The effective overburden pressure should be

calculated with respect to the ground surfaceelevation at

the time of earthquake), g = acceleration of gravity

(980Gal).

(3).Check by the critical N value (The fine content of a

soil is less than 5%)

The zone in Fig.(4.6.2) to which a soil layer belongs is

determined from the equivalent N value and the

equivalent acceleration. The boundary line of the zones

(N)65corrected = (N)65 / cN (4.6.3)

4-23


Where: (N)65 =an equivalent N value obtained from

Equation(l), N =SPT N value of a soil layer, Ip = a

plasticity index ofa soil.

(N)65corrected = (N)65 /0.5

(N)65corrected = N + M

M = 8+ 0.4(1p - 10)

(4.6.4)

(4.6.5)

(4.6.6)

can be evaluated by the relationship between cyclic stress

ratio and number of cycles(=20) to the defined

liquefaction initiated state as shown in Fig.(4.6.4). The

in-situ liquefaction strength ratio Rrnax is given by the

following equation,

(4.6.7)

(4.6.8)

Undrained Cyclic Triaxial Test Results

Rrnaxtvibration type) '"

c

l2;:> 0.4of:::-c 0.3p::CI)CI)

~ 0.2f- ~------':"'_----6

CI)

0.1

O!:-:-----'-----:-'::-~--:-:':-::------:-:-:!0.1 10 20 100 1000

NUMBER OF CYCLES NI

Fig. 4.6.4 Correction of Rmax

0.5

In this equation, several corrections are included as listed

in followings.

(1).Stress condition correction: The stress conditions

between at site( Ko) and in the triaxial cell(isotropic

(2).Type of Input motion correction: The applied stress

condition between at a site high/low degree of

irregularity of input motion(impact type/vibration type)

and in case case ofcyclic triaxial test(harmonic).

Impact type input motion C, =0.55

Vibration type input motion C, =0.7

Applied stress ratio Lmax = T"max / (5' I' is calculated by

seismic response analysis.

The liquefaction potentiahsafetyfactorjf'., is given as,

,5 10 15 20

FINE CONTENT (BELOW 0.075I11m) ('To)

"'=>..J-c>zb:~1.0r----.....

-cu1=

'"u'"o~ 0.5

13;;:

~ 01 --'-- -'- -'--__---'-5 0"''"

Fig. 4.6.3 Reduction facto!" for critical STP-N value

based on the fine content.

The two corrected equivalent N values are plotted in

Fig.(4.6.2) with acceleration and the zone to which

a soil layer belongs is determined as follows. In

the case that the (N+ D N) is inside of the zone ] , the

soil layer belongs to the zone 1 . In the case that the

(N+D N) is inside ofthe zone II, the soil layer belongs

to the zone II . In the case that the (N+ D N) is inside

ofthe zone III or N, and the (N\5 /0.5 is outside ofthe

zone N, the soil layer belongs to the zone ill. In the

case that the (N+ 6.N) is inside of the zone ill or N, and

the (N)65 / 0.5 is inside of the zone N, the soil layer

belongs to the zone N.Case3: The plasticity index is not less than 20 and the

fine content is not less than 15%. A corrected equivalent

N value is calculated by Eqs.(4.6.5) and (4.6.6). The

corrected equivalent N value is plotted in Fig.(4.6.2)

with an equivalent acceleration and the zone to which a

soil layer belongs is determined.

4.6.3 Undrained Cyclic Triaxial Test and seismic

response analysis (Sensitive assess method)

When the liquefaction potential cannot be determined

from the grain size distribution and SPT N value,

liquefaction prediction is made by performing undrained

cyclic triaxial tests using undisturbed soil samples. The

index of a degree of liquefaction strength Rmax of a soil

In case ofFL< 1.0, the soil layer should liquefy.

Reference:(the text mentioned above is revised in 1998 )

POIi and Harbour Research Institute ed., 'Handbook on

Liquefaction Remediation of Reclamimed Land',

Balkema, 1997.

4-24


4.7 Seismic Design of High Seismic

Resistant Quaywalls

4.7.1 Evaluation of seismic performance ofhigh seismic resistant facilities.

(1) In the design process of high seismicresistant facilities, it is requested thattheir seismic performance should beevaluated for a level-2 ground motion toassure that their seismic resistance issatisfactory.

(2) Seismic performance should beevaluated by appropriately modelingthe soil and the structure of the facility,with a method which is appropriate forthe particular type of the structure.

Ground motion which is used for the evaluationof seismic performance should be determinedwith response analysis of the ground inprinciple.

Explanation

(1) Evaluation of the residual deformation ofhigh seismic resistant facilities, which is basedon a earthquake response analysis, is requiredfor the purpose of verifying that they willsustain their intended functions after a level-2ground motion. The reason is that, for theexamination of the stability of the structure orthe soil for a large ground motion such as alevel-2 ground motion, conventional pseudostatic method is not sufficient.(2) The judgement whether the high seismicresistant facilities will sustain their intendedfunctions based on the results of earthquakeresponse analysis should be based on thecombined considerations on the stability of thestructure after the earthquake, the functionsand the difficulty of restoration work. Althoughthe allowable residual deformation should bedefined for this judgement, it is not easy tospecify the allowable deformation at the presentstate of knowledge. Tables 4.7.1 and 4.7.2, inwhich the possibility of temporal use just after

4-25

the earthquake is presented, should be usefulfor the judgement. It should be noted, however,that these tables cannot be applied to aquaywall with cranes because the stability orthe function of the cranes is not addressed inTables 4.7.1 and 4.7.2. In the case of the 1995Hyogoken-Nanbu earthquake, some of thecaisson quaywalls with a normalizeddeformation (lateral residual displacement Iheight of the quaywall) of over 10-20% wastemporary repaired and offered for immediateuse just after the earthquake. .

Related information

forseismic resistant quaywalls.

4.7.2 Design Seismic coefficient ofhigh seismicresistance facilities

(1) When pseudo-static design is applied tohigh seismic resistant quaywalls, the designseismic coefficient should be determined by aglobal judgement base on the seismiccoefficient determined by EqAA.l withimportance factor 1.5, by following equationsfor which peak ground acceleration should becalculated with ground response analysis forlevel-2 ground motion, and by otherappropriate methods.1. If a is smaller than or equal to 200Gal,

Kh=a/g (4.7.1)2. Ifa is larger than 200Gal,

x,=(113) X ( a Ig)(lJ3) (4.7.2)Here, Kh is horizontal seismic coefficient, a ispeak ground acceleration at free surface and gLS the acceleration of gravity.

Explanations

(1) When the design seismic. coefficient can beaccurately determined by investigatingregional seismic activity, characteristics ofground motion, site response, ete., it ispreferable to use this design seismiccoefficient instead of the value designatedhere. For example, when the design ground


Table 4.7.1 Allowable residual deformation from the viewpoint of availability

i

I Amount of deformation

Type of I Gravity quay wanstructure. Sheet-pile quay wallI

Depth of More than 7.5m Less than 7.5m More than 1.5m Less than 7.5mwater

'Available: o-30cm o-20cm o-30cm o-20cmN at available 30-l00cm 20-S0cm 30-S0cm 20-30cm

Table 4.7.2 Allowable residual displacement from functional point ofview

Subsidence of whole apron 20-30cmMain structure Inclination 3 - 5 0

Irregularity of the horizontal 20-30cmdisplacement offace line

Irregularity of subsidence 3 -lOcmApron Gap between apron and backyard: 30-70cm

Inclination normal: 3-5% reverse: 0%

motion is determined based on the information

regarding regional seismic activities or based

on strong ground motion observations or when

seismic response analysis of the structure is

conducted, design seismic coefficient can be

determined based on these results.(2) In the design of high seismic resistant

facilities. target earthquake has to be selected

from earthquakes including hypothetical

earthquake in the disaster prevention plan set

by local government.

(3) One way of calculating peak ground

acceleration at free surface is to use multiple

reflection model for the response analysis of the

ground.

(4) Refer to the reference 1) and 2) for the

details ofEq.4.7.1 and Eq.4.7.2.

4-26

(5) From the experience of significant damage

at Kobe Port during the 1995 Hyogoken-Nanbu

earthquake, minimum design seismic

coefficient for high seismic resistant facilities

should be 0.25 if the site is ill a near-source

region.(6) When it is desired, seismic resistant qua

walls should be designed for level-2 ground

motion with a method other than pseudo-static

method such as earthquake response analysis.

In this case, it is necessary to make sure that

seismic resistant facilities will sustain their

structural stability for level-I ground motion.

Related information

(1) Level-2 ground motion for high seismic


Requirement of performance

IGround motion

I

Size of earthquake

Selection of target earthquake

, (Near-source or not?)

PGA at bedrock

Selection of waveform

Earthquake response analysis of ground

I

Regional seismic coefficient,XF'actor for subsoil condition>< Importance factor O. 5)

I

; Seismic coefficient

II

Type and parameters of structure,

soil improvement,I-

function of facility

I

Cross section of the facility -

IAssessment of liquefaction and mitigation l-

IExamination of residual,deformation for level-2 ground motion

r--

IDetailed design

Figure 4.7.1 Design process of high seismic resistant facilities

resistant facilities(a) Ifhypothetical earthquake is not designatedin the regional disaster prevention plan or if thehypothetical earthquake in the disasterprevention plan is not appropriate fordetermining level-2 ground motion, it isrecommended to select an earthquake whichbrings the largest ground motion to the siteamong earthquakes in the past andhypothetical earthquakes on active faults.Magnitude of hypothetical earthquake on activefaults can be estimated with following equation.

Log1oL=O.6M-2.9. (4.7.3)Here, L is the length of the fault (kID) and M isthe magnitude. Sometimes several active faultsare closely located to each other in the faultmap. In such cases, if one fault is within 51an

4-27

£rom another fault, these faults should beconsidered as one long fault m thedetermination of magnitude. If there isdifficulty in the application of EqA.7.3, themagnitude 7.2 can be used, which is the sameas the 1995 Hyogoken-Nanbu earthquake.(b) Following equation" can be used todetermine peak ground acceleration atengineering-oriented bedrock.

Log lOAsMAC=O.53M .-loglO(X+O.0062 x lOo.53~

- O.00169X+O.524. (4.7A)

Here, ASMAC is the peak ground accelerationmeasured with SMAC-type accelerograph (Gal),M is the magnitude, X is the closest distancefrom the fault to the site (km). The relation isshown in Fig.4.7.2. If the dip angle of the fault


0;Q.<t: 100C)

""'9

U<C 7;:;;UJ

"

10:3 ~ :s 6 3 04 5 G

'0 100

Fault distance X [km]

Figure 4.7.2 Attenuation relations for peak acceleration

for engineering-oriented bedrock

MAX. 170 Gel

2010

, .--------1---------~------·----__:_------·----

1 !

------------~---------------.1--------.------

; !

200

;;a I a) 5-252 NS a...

Q. 100

<::0 °+J

'"...i -100

o...: -200

0

Time [s]

201510

MAX, 161 Gal

---------r-------------.t----------,

-.- -----}-------------1--------------I •. :

:::;' 200

'" (b) 5-1210 E 041SQ.

<::0

+J °..j

-100

8...: -200

0

Time [s]

20I.10

---+----------------r---------------'---'----:'---------'------------

;;a 60O(c) Pi-7S us Bue

Q. "00

<:: 200

.S....... -200'"Qj

-4008< -600

°Time [s]

Figure 4.7.3 Time history of representative strong motion records

4-28


80r-------,-------,--------,-----,

,5

Ill} S-252. NS e-.

10

-------r-------r-----.-----: :i ;: :

Frequency [Hz]

'Or---~---,--------,-------,-----..,

i (b) 5-1210 e.1 S

:60 -------~--t--~-------r·-------;------

.. een-i--+-_==f-v Vf\A i. -+ ~':_-i~

'0Frequency [Hz]

i(e) Pf-7R N& a-

---j-'------+----:-J=-: : s.--UHzi------l-----:----

---~------+--_._ .._-+-_._--: :! :: :

~O ••-----.- -----~--.----!--~-----

! :

'ai'•en 200

Q.a'"....<)

'"0-so 100

.~....'"~

2010

oL__---'-__-=::::::::::::::::=h====dc

Frequency {Hz]

Figure 4.7.4 Fourier spectra of representative strong motion records

30

2S

a20c:.

$(

C.:l'" 1 S

'5

~'"r.:. /0

5

0

I, I

I, I

: : J ! ,/---r------l--------r-----:------T-l--------• I I III , I 1/

: : ; II, , ,: : : I-----;----------;-------;--------- T-----------, , I I

l : : IliB 1 ;--1---------;----..------~--- T-------: l I!! 1" ,

----t----------+-----~ -+ w_" ,i J IA: ~ I:: :

-r-~=~r -i-r--e

5 6 7 9

MagnitudeM

Figure 4.7.5 A diagram presented for determining whether near-source effects

should be considered

4-29


is unknown, the distance from the site to theupdip projection of the fault can be used inEqA. 7A. If the location of the fault cannot bedetermined, the distance from the site to asphere with a radius determined from followingequation can be used.

LoglOr=0.5M-2.25. (4.7.5)Here, r is the radius (km) and M is themagnitude. Engineering oriented bedrock isdefined as a soil layer with shear-wave velocityover 300m/s, a sandy soil with SPT-N valueover 50 or a cohesive soil with qu over650kN/m2

.

(c) Hypothetical earthquakes can be dividedinto intra-plate earthquaJre and inter-plateearthquake. Strong monon records at POItIsland during the 1995 Hyogoken-Nanbuearthquake can be used as a representativeground motion from intra-plate earthquake.Records at Hachinohe Port during 1968'Iokachi-oki earthquake or at Ofunato duringthe 1978 Miyagi-ken-oki earthquake can beused as a representative ground motion frominter-plate earthquake. It is recomended to usePort Island records when it is necessary to takeinto account near-source effects even when theearthquake is a inter-plate earthquake becausenear-source ground motion from inter-plateearthquake has not been obtained yet. Timehistory of these records are shown in Fig. 4.7.3and Fourier spectra of these records are shownin FigA.7A.(d) FigA.7.5 can be used to determine whetherthe site is in near-source region. If the site islocated in A of Fig. 4.7.5, then the site is innear-source region.(e) Sometimes design seismic coefficientobtained from EqA.7.1 and EqA.7.2 is smallerthan those obtained from regional seismiccoefficient, etc. This is partly because the factorof subsoil conditions are not necessarilyconsistent with the amplification factorobtained from response analysis. In the case ofweak soil, peak ground acceleration at thesurface is often smaller then expected formlinear theory due to nonlinear effects duringlarge earthquake. It should be noted thatground motion at weak soil site sometimes

4-:30

causes large deformation of structures evenwhen peak ground acceleration is small.

Reference

1) "Relation Between Seismic Coefficient andGround Acceleration for Gravity Quay Wall" S.Noda, T. Uwabe and T. Chiba, Report of thePort and Harbour Research Institute, Vol.14,No.4, 1975 (in Japanese with English abstract).2) "Relation between Seismic Coefficient andPeak Ground Acceleration Estimated fromAttenuation Relations" by A. Nozu, T. Uwabe, Y.Sato and T. Shinozawa, Technical Note of thePort and Harbour Research Institute, No.893,1997 (In Japanese w-ithEnglish abstract).


4.8 New Seismic Design of Open Piled Piers

4.8.1 General

The seismic design of open piled piers described in

this section is basically to verify if they possess the

required structural performance during earthquakes.

At first, the fundamental dimensions of structural

members to be verified are designed with the

allowable stress method against loads except seismic

forces. Then structural performance of the pre

determined section under seismic actions is checked

considering seismic energy absorption due to plastic

deformation of steel pipe piles. The required

structural performance will be determined m

consideration of the importance and the role of

structures, and will be expressed in terms of

horizontal displacement and the place and the timing.

of local damages.

Open piled piers were damaged by liquefaction of

the base ground or the backfilling soil of retaining

structures behind them due to the 1995 Hyogoken

Nambu earthquake. The liquefaction produced

Dynamic analysis of ground

Maximum acceleration of the base

Model for

Ground response anafsis (SHAKE, etc.)

Acceleration at 1/ below the seabed

unforeseen external forces and caused buckling of

steel pipe piles and cracks in concrete superstructures.

Since it should be rather difficult to estimate the

magnitude of such external forces, the design method

for open piled piers assumes that liquefaction does

not occur. Therefore, liquefaction should be

prevented. However, the effect of liquefaction

. should be considered for very important facilities.

4.8.2 Seismic performance requirements

_It should be examined appropriately whether open

piled piers will perform as required when earthquakes

happen. For common piers, structural performance

should be verified against Level 1 ground motions.

Both of Levels 1· and 2 ground motions should be

taken into consideration for high seismic resistant

piers. Seismic performance requirements depend on

the importance of piers and can be described with

indices such as extent of damage, maximum

displacement, and residual displacement after

earthquakes. The difficulty of· repair to damages

Natural period of pier

Frame model, etc.

Seismic coefficient

Fig. 4.8.1 Calculation flow of seismic coefficient

4-31


10

(4.8.2)

1. 0

Period (s)

Region AI. 00I-~~~~---"----------'

O. 01~ '---..L-_-,- ---'-,-.J

O. 1

Fig. 4.8.2 Standard seismic coefficient

......c(!)

'0:i=(!)oo O. 10 r---:----'---~+--~~-~~(.)

E(f)

'CD(f)

where Tg is the natural period of the ground, Hi is the

thickness of zth stratum, and Vsi is the shear wave

velocity. Vsi can be assumed using N value of the

ground.

The natural period of the wharf can be calculated

using Eq. 4.8.2:

should be considered well at the same time.

Resonance will make the response of a pier higher

than that expected when the natural periods of the

structure and the ground are close to each other. In

such a case, dimensions, geometry, etc. of the

original structure should be modified to achieve

different natural periods. Alternatively, the ground

should be improved so that the natural period of the

ground is changed.

For the structure of an access bridge to link a pier

to the land, the following should be taken into

account: (a) In case large seismic force acts on the

pier towards the retaining structure, possible contact

between the pier and the retaining structure, through

the access bridge, should be avoided by checking the

available clearance against the maximum

displacement. (b) In case the seismic force acts

towards the sea, the maximum displacement should

be checked to prevent the access bridge from falling

down.

If case that cargo handling machines such as

container cranes are equipped on piers, the interactive

vibrations between them should be well examined.

4.8.3 Design earthquake forces

Figure 4.8.1 shows the sequence for calculating the

seismic coefficient. Level 1 ground motions are

defined as those with a 50-percent probability of

exceedance in 75 years. Expected values of the

base ground acceleration depend on the region where

piers .are constructed. Acceleration response that

corresponds to the natural period of pier is obtained

as shown in Fig. 4.8.1. The design seismic

coefficient will be given by dividing the response

acceleration by the acceleration of gravity. Instead

of performing the dynamic analysis of the ground, the

seismic coefficient can be obtained using the standard

spectrum shown in Fig. 4.8.2 as an example.

The natural period of the ground can be

approximately calculated by Eq. 4.8.1:

(4.8.1)

where T, is the natural period of the wharf, W is the

weight of the wharf and surcharge, g is the gravity

acceleration, and Kh is the,horizontal spring constant

of each pile.

The seismic coefficient for Level 2 ground motions

is obtained using the same procedure shown in Fig.

4.8.1. Dynamic analysis of the ground is necessary

because charts like Fig. 4.8.2 are not available for

Level 2 ground motions at present.

4.8.4 Structural analysis procedures

Seismic performance of an open piled pier should

be examined with appropriate analytical. models to

simulate its non-linear behavior as close as possible.

Plastic deformation of steel piles provides excellent

seismic structural capacity particularly during Level

2 ground motions, but also initiates local minor

4-32


(4.8.6)

(4.8.7)

damages. Therefore, requirements regarding

locations and extent of these damages should be

specified at the verification.

For verification of the seismic capacity of a pier,

the ductility design methods recommended to be

applied are: (a) simplified analysis, (b) elasto-plastic

analysis, or (c) non-linear dynamic analysis.

(a) Simplified analysis

The superstructure of a pier is considered to be a

rigid body, and the capacity of the pier is evaluated as

the overall capacities of each pile. This method is

applicable to piers supported on vertical piles with

small variety of their rigidities.

(b) Elasto-plastic analysis

A pier and its surrounding ground are modeled by

a frame and springs, which represent their non-linear

properties. This method is suitable for complicated

structures whose capacities might be overestimated

by the simplified analysis. The seguence of local

failures (generation of plastic hinges, damage of

superstructures, etc.) and the maximum and residual

displacements will be verified.

(c) Non-linear dynamic analysis

The pier structural system is analyzed by the finite

element method considering non-linear and dynamic

properties. This method is applied when the whole

structural system is complicated or large deformation

of the ground is predicted.

In the ductility design method, the Newmark law

of constant energy is assumed. The basic equation

for the verification' is presented as Eg. 4.8.3:

(4.8.3)

where R; is the load carrying capacity during

earthguake, K; is the design seismic coefficient, and

W is the vertical loads including self-weight and

surcharge. Ra is given by Eg. 4.8.4:

s, = ~2/-1-a -1+8(/-I-a-1)2 Py (4.8.4)

where P-a is the allowable plasticity ratio, e is the

ratio of the secondary gradient to the primary

gradient in the load-displacement relationship, and P,

is the horizontal force corresponding to the elasticity

limit. At the elasticity limit, pile-head bending

4-33

moments can reach their fully plastic moments in

about a half of all piles, and after that, horizontal

displacement may rapidly increase.

P-a is the ratio of the allowable maximum

horizontal displacement to that at the elasticity limit.

On the basis of analytical and experimental results,

/-la is summarized in Table 4.8.1 for Level 1 ground

motions and presented in Eq. 4.8.5 for Level 2

ground motions.

Table 4.8.1 Standard values of u; for Levell

ground motions

Classification of pier P-aSpecial class 1.0

A class 1.3B class 1.6C class 23

!-La = 1.25+ 62.5(t /D) :s; 2.5 (4.8.5)

where P-a is the allowable horizontal displacement

ductility factor, t is the thickness of pipe pile, and D

is the diameter of pipe pile.

Instead of performing elasto-plastic analysis, the

elasticity limit P, can be given by Eqs. 4.8.6 and

4.8.7 based on parametric studies:

P, = O.82-p.all

Paall = L{2M pi /(hi +1/ f3i )}

where PI/all is the horizontal force when bending

moments at the pile head and the assumed fixed point

under the ground of all piles reach their fully plastic

moments, Mpi is the fully plastic moment of each pile,

and (hi+1/(3;) is the length between the head and the

assumed fixed point of each pile.

The residual horizontal displacement of a pier can

be calculated on the assumption that the

load-displacement relationship during unloading has

the same gradient as that during initial loading.

The structural behavior of high seismic resistant

piers should remain within elastic regions during

Level 1 ground motions, and it should be controlled

with horizontal displacement, as mentioned above,

during Level 2 ground motions. The load carrying

capacity R; in Eg. 4.8.4 can be calculated with eand


Fig. 4.8.3 f-L. and e

methods. The shear failure should not occur prior to

bending failure.

The properties of piles pushing into or pulling out

from the ground should be modeled taken into

consideration their non-linear behaviors. Lateral

resistance of the ground should also be modeled as

the same way above.

Buckling of piles often dominates the ultimate

state of a pier. Equation 4.8.14 gives the strain that

may cause local buckling in steel pipe piles:

C ma• = 0.44t / D (4.8.14)

where t is the thickness of the steel pipe and D is the

diameter of the steel pipe.

The moment at the strain of Cmax is not much

different from Mp in Eq. 4.8.8 for steel pipe piles with

commonly used dimensions.

"'0 Pp<tl0 P)'

CO.....c0N'C0I

0

f-La obtained by the result of elasto-plastic analysis as

shown in Fig. 4.8.3.

Either a bi-linear or a tri-linear model can be used

for the constitutive law of steel pipe piles. Fully

plastic moment Mp , yielding moment My, and their

respective curvature ¢>p and ¢y can be given by Eqs.

4.8.8 to 4.8.11.

M p = M pa cos(an /2) (4.8.8)

My = (Jy -N / A)Ze (4.8.9)

¢y = My / £1 (4.8.10)

¢p =(Mp-MJrf>y (4.8.11)

where Mpo is the fully plastic moment of the steel

pipe without axial force expressed as follows:

«; =Zpfy (4.8.12)

Zp is the plastic sectional modulus of the steel pipe,

which can be obtained as follows:

Zp =~{r3 -(r -tJ} (4.8.13)

where r is the radius of the steel pipe, t is the

thickness of the steel pipe, a is the ratio of applied

axial force N to yield axial force No (No = A h) when

bending moment is not applied, A is the area of the

cross section of the steel pipe, fy is the yield strength

of the steel pipe, Z, is the elastic sectional modulus of

the steel pipe, and EI is the flexural rigidity of the

steel pipe.

The concrete superstructure is represented by the

tri-linear bending moment-curvature relationship.

Cracking moment, rebar-yielding moment, ultimate

moment should be calculated with appropriate design

Allowable ultimate stagei

(5!' 0 p

Displacement

4-34

5. BASIC PRINCIPLES OF SEISMIC DESIGN AND CONSTRUCTION

FOR WATER SUPPLY FACILITIES

JAPAN WATERWORKS ASSOCIATION

5.1 Basic Concept of Anti-Seismic Measures for Water Supply Facilities 5- 1

5.1.1 General 5- 1

5.1.2 Planning, Designing and Implementation 5- 1

5.1.3 Geotechnical Survey of the Foundation 5- 2

5.1 A The Employment of Highly Earthquake Resistant Materials and Joints 5- 3

5.1.5 Anti-Seismic Design of Water Supply System 5- 3

5.1.6 Maintenance and Planned Improvements 5- 4

5.1.7 Assumption of Earthquake Damage 5- 5

5.1.8 The Order of Restoration Works and Relationship Between Emergency

Restoration Works and Permanent Restoration Works 5- 5

5.2 Basic Concept of Anti-Seismic Design 5- 5

5.2.1 General 5- 5

5.2.2 Seismic Motion Levels for Anti-Seismic Design 5- 6

5.23 Importance Ranking of Facilities 5- 6

5.2A Anti-Seismic Level Which Water Supply Facilities Must Maintain

During and Earthquake 5- 6

5.2.5 Earthquake Effects on Anti-Seismic Designs 5- 7

5.2.6 Sequence of Anti-Seismic Design Works 5- 7

5.2.7 Related Regulations 5- 8

53 Seismic Motion Input for Anti-Seismic Design 5- 10

53.1 Anti-Seismic Calculation Methods and Objective Structures 5- 10

53.2 Seismic Intensity Used in Anti-Seismic Design under the Seismic

Intensity Method for Ground Structures (Seismic Motion Levell) 5- 10

533 Seismic Motion Level Used in Anti-Seismic Design by the Seismic Intensity

Method for Ground Structures (Seismic Motion Level 2) 5- 11

5.3A Seismic Intensity Used in Anti-Seismic Design by the Seismic Intensity

Method for Buried Structures (Seismic Motion Levell) 5- 15

5.3.5 Design Seismic Intensity Used in Anti-Seismic Intensity Method for

Buried Structures (Seismic Motion Level 2) 5-16

5.3.6 Seismic Motion Level Used in the Response Displacement Method for

Buried Structures (Seismic Motion Levell) 5- 16

5.3.7 Seismic Intensity Used in Design of Buried Structures by the Response

Displacement Method (Seismic Motion Level 2) 5- 17

5.3.8 Seismic Motion Input Used in Design Using the Dynamic Analysis 5- 19

5.4 Geotechnical Surveys, Ground Displacement, and Ground Distortion 5- 19

5.4.1 Primary Subjects of Geotechnical Survey 5- 19

5.4.2 Methods of Geoetchnical Survey 5- 20

5.4.3 Soil Liquefaction and Lateral Spreading 5- 21

5.4.4 Ground Displacement and Ground Strain Caused by Liquefaction 5- 21

5.4.5 Ground Strain at the Incline of Artificially Altered Ground 5- 21

5.4.6 Reduction in Reaction Force and Ground Friction Force due to Soil Liquefaction 5- 22

5.5 Soil Pressure During an Earthquake 5- 23

5.5.1 General 5- 23

5.5.2 Calculation of Horizontal Soil Pressure During an Earthquake 5- 23

5.5.3 Calculation of Vertical Soil Pressure During an Earthquake 5- 23

5.5.4 External Pressure due to Lateral Spreading 5- 24

5.5.5 Buoyancy Generated by Soil Liquefaction 5- 24

5.6 Hydrodynamic Pressure During an Earthquake and the Water Sloshing 5- 24

5.6.1 Hydrodynamic Pressure During an Earthquake 5- 24

5.6.2 Water Sloshing 5- 25

5.7 Safety Check 5- 25

5.7.1 Combination of Loads 5- 25

5.7.2 Safety Check of the Structures Fabricated with Steel, Concrete, etc 5- 25.

5.7.3 Safety Check of Pipeline in their Anti-Seismic Calculations 5- 26

5.7.4 Safety Check ofthe Foundation Ground in its Anti-Seismic Calculations 5- 26

5.7.5 Safety Check of Foundation, Earthen Structures and Retaining Wall in

Anti-Seismic Calculations 5- 27

5.7.6 Safety Check in Anti-Seismic Calculations in Consideration of Critical State

under Seismic Motion Level 2 5- 27


5. BASIC PRINCIPLES OF SEISMIC DESIGN AND CONSTRUCTION FOR

WATER SUPPLY FACILITIESJAPAN WATER WORKS ASSOSIATION

5.1 Basic Concept of Anti-Seismic Measures for

Water Supply Facilities

5.1.1 General

For developing anti-seismic measures in water

supply, the following basic plans must be drawn

beforehand:

(1) Proper damage estimates before the

occurrence of an earthquake, and preventive

measures based on such estimates,

(2) Plans on emergency relief measures to be

undertaken immediately after an earthquake,

and disaster prevention measures including

effective emergency repair works and

(3) Detailed plans on the organization for the

implementation of permanent restoration

works in the period from temporary works

in above (2) to the completion of the

permanent works

The underlying goal of implementing anti

seismic measures for water supply systems is to

save human lives. To this end, a plan must be

established to provide well-balanced,

comprehensive measures to be implemented under

adequate mutual understanding with related

organizations, during: (1) the pre-earthquake

period; (2) the immediate post-earthquake period;

and (3) the reconstruction period.

In the pre-earthquake period, the potential scale

5-1

of the disaster must be properly assumed; the

reinforcement works must be implemented based

on rational anti-seismic designs; and everyday

facility maintenance must be carried out with

consideration of the anti-seismic measures.

In the immediate post-earthquake period, it is

essential to collect quick and accurate information,

and establish a communication network. A plan

must be drown. before-hand for calling out

personnel for their deployment for initial response

activities, which are considered the most

important.

During the reconstruction period, in

coordination with the police and fire departments

and under the rescue operations provided by other

water utilities, an emergency water service must

be implemented until restoration of regular water

supply; restoration works must efficiently be

implemented and manpower and materials and

equipment required for such activities must be

procured.

5.1.2 Planning, Designing and Implementation

For preparation of plans and designs of water

supply facilities and their implementation,

sufficient consideration should be given to

earthquakes in accordance with various conditions

in which the water utility is actually situated.

For important facilities, their structures must be


rationally designed with adequate consideration to

the effects from earthquakes.

The construction of water supply facilities must

be so implemented that the water supply system as

a whole retain as much capability to do water

service even though the system has sustained

certain damage.

Water supply facilities are fundamental to the

infrastructure supporting a city and to the lives of

the city's citizens. For earthquake disasters,

maximum effort to develop countermeasures must

be taken in order to insure that the water suppiy is

protected. This is true regardless of the size of

the facility.

Restoration of emergency water supplies is

crucial immediately after an earthquake disaster.

Implementation and execution of an effective

reconstruction plan must be applied in conjunction

with earthquake countermeasure upgrades.

In order to implement and execute adequate

plans for different distribution facilities,

examination of local earthquake records and

accurate predictions and .estimates of future

earthquakes based on changes in the earth's crust

(from geological surveys) must be completed.

5.1.3 Geotechnical Survey of the Foundation

It is desirable that water supply facilities are

founded on location where the foundation is firm

and the landscape is stable. Prior to the

construction of specially important facilities a

careful and detailed geotechnical survey must be

5-2

carried out.

For the construction of the water supply

facilities, it is essential to that a good ground site

be chosen. This is particularly true for the

construction of key facilities such as water intake

facilities, reservoirs, delivery facilities, treatment

facilities, service reservoirs, and main

sending/receiving lines.

The power of an earthquake's force on building

differs greatly depending on various ground

foundations, even ground foundations in the same

region. In addition, the scale of an earthquake's

motion may also differ, depending on the

topographical and geological differences of a

particular region.

For the construction of key facilities,

architectural designs for main buildings and their

foundations must be based on data gathered by a

detailed survey of ground conditions. These

detailed surveys of the construction site must

include an analysis of the site's dynamic behavior

during an earthquake.

When: the ground conditions are not the most

desirable, improving the foundation through

substructure work or additional slope stabilization

work must be applied. Preventative measures,

such as the use of flexible structures which

respond to ground 'floating' or displacement

during an earthquake, must also be used.

For ground foundations in areas with high

ground water levels, such as sandy soil (which

easily generate ground Fluidization), suitable

measures must be adopted. These methods


include flexible ground structures that absorb

ground displacement and reinforced ground

foundations.

5.1.4 The Employment of Highly Earthquake

Resistant Materials and Joints

For construction of main water supply facilities,

earthquake resistant materials should be employed

in structurally important locations.

For water containing facilities, structures,

which can absorb structural strain and abate stress,

must be designed with the provision of earthquake

resistant joints, which absorb expansion,

contraction and distortion must be provided

between interfacing structures which may move,

when an earthquake occurs, and leave relative

displacement.

Underground pipelines will bend as a result of

the ground displacement produced by an

earthquake. Such displacement tends to escalate

in areas where the geography or topography is

subject to sudden change. As a result,

connections between the structure and related

pipes are subject to great distortion. This

distortion results from the difference between the

rigidity of the structures and the related pipes. In

addition, the alteration of fluid ground also

produces irregular and uneven ground surfaces.

This results in movement and distortion of

structural bulkheads. On such ground, flexible,

anti-seismic joints capable of absorbing the

displacement generated during an earthquake must

5-3

be used. In addition, flexible materials that are

capable of absorbing earthquake displacement

may also be used to avoid structural damage.

5.1.5 Anti-Seismic Design of Water Supply

System

For anti-seismic design of water supply systems,

the followings are prerequisites:

1) Earthquake damage is localized as much as

possible.

2) The damage is easily repaired.

3) Measures, which will prevent secondhand

disasters as a result of an earthquake, must

be provided.

To meet these conditions, redundancy in

important facilities, interconnection of block

systems, grouping of such systems, separation of a

pipe network into blocks, and installation of

emergency cut-off valves must be implemented.

When water conveyance and distribution

pipelines receive earthquake damage, the water

supply in an entire distribution area may be cut off

ad severe conditions may result.

When the system of water conveyance pipelines

is interconnected, the transmission and

distribution of water may be cut off when one

portion of the system is cut off or out of service.

Therefore, it is desirable to have a system which is

capable delivering water even after sustaining

damage. Using an interconnected system with

different functioning lines is the most effective

method of supplying water when an earthquake


disaster occurs. This method is also effective for

responding to other common disasters and for

providing raw water for regular usage.

It is desirable to connect the main water supply

pipeline to other water works facilities in

neighboring vicinities. In order to minimize the

damage during an earthquake, construction of

pipeline networks must take into consideration the

following issues:

1) Minimizing the range of water delivery failure

after an earthquake by adequately spacing

gate valves in the pipeline network, making

the distance between them the shortest

possible.

2) Limiting the depth of underground pipes in

order to insure that they are not buried too

deeply. In addition, properly locating access

and work station doors in the facilities to

make restoration work swift and easy.

3) Using preventative measures, such as setting

gate valves both in back and in front of a pipe

when the pipe crosses over a railroad or a

large river and installing chlorine neutralizing

devices. These preventative measures must

be utilized because damage to a water work

facility may generate secondary damage to

important public and private facilities or to

neighboring residences.

5.1.6 Maintenance and Planned Improvements

Adequate inspection and maintenance of water

supply facilities must be undertaken at a basis to

insure their anti-seismic integrity.

5-4

Positive anti-seismic diagnostic inspections in

accordance with this manual must be conducted.

Facilities with low anti-seismic ratings must

undergo improvement works through planning.

Planning upgrades to existing water works

systems (in order to make them more earthquake

proof) must utilize competent anti-seismic

diagnosis. Such diagnosis in necessary for

existing facilities in order to execute

reinforcement or renovation. First, in order to

perform diagnosis, a water works system must be

broken down, with each facility being categorized

and listed in order of its importance. Second,

initial diagnostic inspections must be conducted

and the priority of work must be decided upon.

Third, improvements or reinforcement must be

proceeded with.

To create anti-seismic water work facilities,

design and execution must be carefully carried out.

After completing construction, constant inspection

and maintenance of the facility must be carried

out. To fulfill this purpose, listed inspections

and maintenance must be set and routinely

followed.

For the improvement of the existing facilities,

earthquake resisting measures and measures

aimed at the prevention of numerous, everyday

accidents must be taken. These improvements

must be carried out after a comprehensive and

integrated evaluation.


5.1.7 Assumption of Earthquake Damage

Based on assumptions on the type and

magnitude of damage to the water supply systems

as a result of an earthquake, plans for emergency

water service and repair works must be

established. To facilitate such works,

information networks, emergency manpower

mobilization plans, and mutual cooperation

system must be established; and comprehensive

preparation must be made for stockpiling

materials and equipment required for restoration

Each part of water supply facilities must be

designed to retain its respective capacity even

after an earthquake with the design intensity of

seismic tremor.

The more important the facility, the more the

need for such consideration.

For anti-seismic design of facilities, the design

method must be employed, which is suitable to the

characteristics of respective facilities and the

nature of their founding and surrounding soil.

works, preparation of complete sets of facility 5.2.2 Seismic Motion Levels

drawings and

management.

decentralization of their Design

5.1.8 The Order of Restoration Works and

Relationship Between Emergency

Restoration Works and Permanent

Restoration Works

As a general rule, restoration work after an

earthquake disaster should start with raw water

intake facilities, followed by, in sequence, water

treatment facilities, transmission and distribution

facilities, and finally water service connections.

To realize early resumption of water service,

sufficient consideration should be made on

relationship between emergency restoration works

and subsequent permanent restoration works.

5.2 Basic Principles for Anti-Seismic Design

5.2.1 General

5-5

For an anti-seismic design, two different

magnitudes of intensity must be employed:

Seismic Motion Levell, which has a return

probability of once or twice in the service lie of

the facility, and Level 2, which has a smaller

probability than the former but is greater in

magnitude.

Seismic Motion Levell (Ll) is equivalent to

the conventional seismic motion level set by many

civil engineering construction guidelines.

Seismic Motion Levell may be generated once or

twice during the in service period of a structure.

Seismic Motion Level 2 (L2) is the equivalent of

the seismic motion generated in areas with faults

or areas with big scale plates bordering. inland

areas, such as the earthquake which struck the

southern area of Hyogo Prefecture in 1995.

The probability of a water works system


experiencing Seismic Motion Level 2 is very low.

Nevertheless, the influence of a Seismic Motion

Level 2 is considered enormously great.

However, information on seismic motion

parameters for a fault may be used to search for an

appropriate location. If a seismic motion caused

by active faults is clearly understood using

preliminary surveys, a construction design can be

directly evaluated.

5.2.3 Importance Ranking of Facilities

In principle, for planning anti-seismic design of

water supply facilities, they must be categorized

into two:

(1) facilities at a high level of importance

(Rank A), and

(2) other facilities (Rank B).

Each water utility must sort the Rank A

facilities based on the actual position of their

system, and with consideration to the following

conditions:

1) Facilities which possess the potential to

generate serious secondary disasters.

2) Facilities located up stream of water supply

system.

3) Main facilities which do not have backup

facilities.

4) Feeder mains to important facilities.

5) Main facilities which are difficult to restore

if damaged

6) Facilities which will become the center for

gatheringinformation during a disaster.

5-6

It is not realistic to demand the highest level of

earthquake durability every component of a water

supply system. When implementing anti-seismic

planning for a facility, the facility's degree of

significance must be categorized into either Rank

A or Rank B. In addition, the degree of

importance must be combined with the two

Seismic Motion Levels, Level 1 and Level 2.

Through these combinations, it is possible to

create different designs with different anti-seismic

capabilities. Refer to 5.2.1 (general concept) and

5.2.4 (anti-seismic levels for water works facilities

during an earthquake).

The significant degrees are decided by

individual work groups, based on their own

judgment, experience, locallspecialized reasoning,

and consideration of local disaster prevention

programs. Factors effecting a facility's degree of

significance are grouped in two categories: those

factors which, during an earthquake, may

influence non-water works facilities and those

factors which may effect the conventional

functions of awter works facilities.

5.2.4 Anti-Seismic Level Which Water Supply

Facilities Must Maintain During an

Earthquake

Water supply facilities should maintain either

one of the following anti-seismic standards, which

are set by combining the Seismic Motion Level

(Ll and L2) and the importance ranking (Rank A

and Rank B) of the facilities.


\ Indhdd:u~rfa¢lllM".Q~Y

··r~;"t~H$p:t; ,Islil~~•••d~~:g#!!'l:rtd

rllil'llktl I$ayt\4l:b¢ahl~t\)$j;;nltJan,

'B~kt

r~t4rnHi:ln

?¢1I$ibl¢.

5.2.5 Earthquake Effects on Anti-Seismic

Designs

For anti-seismic design, the following effects of

earthquake must be taken into consideration:

1) Displacement and distortion of the

foundation soil during an earthquake,

2) Inertial force owing to the weight of

structures,

3) Soil pressure during an earthquake,

4) Dynamic water pressure during an

earthquake,

5) Water surface sloshing,

6) Lateral soil movements due to liquefaction

of the soil, and

7) Soil distortion on a slope of reclaimed land.

Facilities which are built on ground that is

clearly subject to rapid/dynamic change, such as

ground subject to horizontal, fluid displacement or

5-7

ground located above an active fault, must be

given sufficient consideration due to the potential

for disaster. In particular, new housing on slopes

which have been artificially altered are especially

subject to ground distortion and displacement

generated by a Seismic Motion Level 2.

This earth load stress can be evaluated using

methods based on the response displacement

method.

There are two types of hydraulic water force:

one which exerts inertial force directly against a

facility and one which exerts secondary,

osciliating force on the surface of free water.

Facilities which abut reservoir structures,

underground water storage tanks, dams, or water

intakes receive dynamic water pressure during an

earthquake. As a result, the design of such

facilities must take into consideration the

influence of this pressure.

The effects of surface oscillation in water on a

structure must be determined by analyzing the

oscillation characteristics of a structure and the

frequency of surface water.

5.2.6 Sequence of Anti-Seismic Design Works

As a general rule, anti-seismic designs of water

supply facilities must be carried out in the

following order:

1) Selection of the construction site,

2) Geotechnical survey at the site

3) Selection of the type of structure and the

study on geotechnical conditions of

foundation,


4) Anti -seismic calculation,

5) Examination of anti-seismic level

Figure 5.2.1 illustrates the method for anti

seismic calculation. This method may differ

according to the structural characteristics and

ground conditions involved. The appropriate

calculation. method must match the structure's

propose and condition.

A facility belonging to Significance Rank B

must designed by matching it to Seismic Motion

Level 1. Depending on the facility's necessity,

evaluation under Seismic Motion Level 2.must be

made.

5.2.7 Related Regulations

When designing water supply facilities, existing

laws and related regulations whichever applicable,

must be followed. In addition, it is desirable that

technical standards established by institutions or

associations are followed.

5-8


S TAR T

SELEcrION OFCONSTRUcrION SITE

DECISONOF SIGNIFICANTRANKING OF FACILITIES

SOILSURVEY

STRUcrURE DESIGNANDl'KAMiNG

STATIONARY LOADCALCULATION

ANTI-SEISMIC CALCULATIONSFORSEISMIC MOTIONLEVEL 1

No

ANTI-SEISMIC CALCULATIONSFORSEISMICMOTIONLEVEL2

Yes

No

No

E N D

fiGURE· 5.2.1ANTI-SEISMIC STRUcrURE DESIGNORDER

5-9


5.3 Seismic Motion Input for Anti-Seismic

Design

5.3.1 Anti-Seismic Calculation Methods and

Objective Structures

1. The following are standard anti-seismic

design methods to be applied for water supply

facilities. Their selection must be based on the

structural nature of the objective structures and

other factors.

Depending on the structural nature and special

subsoil conditions, the result of calculation by

means of either 1) or 2) must be cross-checked

with that obtained by mean of 3).

1) Seismic intensity method

2) Response displacement method

3) Reference to the results by dynamic

analysis

2. For ground structures, an anti-seismic design.

must be implemented using the seismic intensity

method. Because the effects of inertial force and

dynamic water pressure, in the case water levels

are full, cannot be neglected, verification of the

safety, using the dynamic analysis method, is

recommended after the seismic intensity method

is applied.

3. Buried structures must be designed using the

seismic intensity method or the response

displacement method. For the anti-seismic

design of a structure whose movements are

complex at the Seismic Motion Level 2, to verify

the results calculation using the seismic intensity

method or the response displacement method, the

5-10

dynamic analysis method must be applied when

required.

For an anti-seismic design of a massive,

partially buried structure (such as a settling basin),

the seismic intensity method may be used.

5.3.2 Seismic Intensity Used in Anti-Seismic

Design under the Seismic Intensity

Method for Ground Structures

(Seismic Motion Levell)

1. The horizontal seismic intensity to be used for

design or structures on the ground surface

shall be determined as follows:

Kh1

::::: C, .Kh01

(5.3.1)

Where:

Cz: Region-specific correction factor.

Values are 1.0-0.7.

Kh01: Standard horizontal seismic

intensity at the center of gravity of the

structure. Values are shown in Table

5.3.1 by the type of subsoil.

The value of KhOl shall be set at 0.16, 0.2,

and 0.24 for ground type I, IT, and ill

respectively.

2. Ground types for an anti-seismic design must

be classified based on proper period obtained

by the equation 5.3.2. If the base of the

foundation conforms to the ground surface,

the ground type must be Type I.

~.................................. (5 3 2)..

Where:


TAlltE

}'(n$T.rj,'"PE••qltOUNtJ··O'il",i!;2)·.ViHEEE··TaXS·THg·.NKl'LlRALPERJfJDOW THEGROU~l1(lt~

TG: Proper period of ground(s)

Hi: Thickness of the I th stratum (m)

Vsi: Average elastic wave velocity (rills)

3. Should the vertical seismic intensity (K.vl) be

taken into account, the following formula

shall be used.

1. The behavior of water works in reaction to a

seismic motion is dependent on factors such

as the earthquake's strength, its periodic

characteristics, its duration, the ground type,

the type of structure involved, the type of

foundation, etc.. Standard design for

horizontal seismic intensity takes these factors

into account.

2. Ground classification is used to determine the

horizontal seismic intensity value (Kh1) on a

construction design, using the seismic

intensity method.

As a "rule of thumb," Type I ground is made

up of ideal diluvial ground and a proper rock

bed. Type ill ground is considered poor

ground and is located at or around the alluvial

5-11

layer. Type II ground does not belong to

either Type I ground or Type II ground.

Instead, Type II belongs to either diluvial or

alluvial categories. The alluvial ground

mentioned here includes new sedimentary

layers created by landslides, landfills, and

other weak ground. The Diluvial layers

mentioned here include hardened sandy soil

layers and layers ofboulders.

5.3.3 Seismic Motion Level Used in Anti

Seismic Design by the Seismic Intensity

Method for Ground Structures

(Seismic Motion Level 2)

1. Horizontal seismic intensity (Kh2) used for

anti-seismic design based' on Seismic' Motion

Level 2 shall be determinedas follows:

KhZ = C, . K hoz···· (5.3.3)

Where:

Cz : Structure specific factor, which must

properly be determined based on the

magnitude of diminution due to the

response of the structure and the

capability of plastic deformation of the

structure.


lAn'lI:.5,;;l;:~ tJ:RPE~LIM:IT.YP:i;.I.JE.··O~'.THE ••STAN[DRD··:Eii€iE1Z0NTAL••SEIS'MIC••·rWTENSI':fTJRtiLDESIG1\!.OJi'~OY$GROtJ'l'fU STRUttt'tJ.R$t$ElSMlCrNTENSl'1'Y$;1B:T.HOP)--_.." .•.,.~~~----.,..

GROUNtlT'l"BE I J£ho2vPJ.;"tt"EAGP-,lNSTSTRl1C)",f'UP,;Jl:.J.. NNl'tJ1't.4.L 'f'BBIDT{Sl

··'1'YrE·"j·· ['ra~D;2J· ... ..w .... ,.... '1'<0,2 O,.z;G.'1'';;l.l) 1..0<:'1'

'H;,·:]'::f".....'O~·IT'l·G'":r'~'T'tl."r"·'1\.f:,,'\:·~t"~:;4-''f''': 1'n:.:~~'r:.=. ::_~:~.~t.q:_::~.-·.:0.'~S::l5

;~;~';~~;~~~:ir;~;'; ••.. ~C;;i¥E\ZB;;;~KkQ!l'~030 J£hoz==·l)} KM2=LOOO·f" lAM

T<O,2 "'··..'''"..,,,,.··'''1.•..•••• '··......'O;2$!1;i'~1 1..0<.'1'

1"Y?EIlKhM=~.lSi)1'Q,aC7 ...' . .. j[0;2&1'o<05J B:O\VEVER; .·KhC;;,~(Ulfl KhG2= 1.4 1{h(;~""lAQ(J1'__ lA(I~

T\?B ill[O,G~'l'GJ

""-·.."1....

ItAanL 5,3:$ E'OT1'M:tilM.l,£ViililJE{$'EI$M1CMOXION tJil'\tEt. 2) OF'STANtJAt1.D.HOR1Z()NT.~...LS:ElSMlt!lNmENSIt$tY,

tiSEnt0:R,GrtOiJNn··S11nU£"tUlte.DESrCN·.?'.a0M·.l'nB.SE~S~~n:c··rtiF:t'ENSln··M)o;'10D..... , ', .. --. .. " .. "': ... : ..... :-... _.,.:.. ~._- ... -.,,.: .... _--.- .... : .... : .. " .. ".:.,: ..... :,.:..:.,.:.: .. ,:': ..... ,'.".:': ......-, .. ,'-''':-···-··:----CC":··

TYP.E 1 Ctc<0.21Tt>JS THE 'NAT{JRlibGROUND PERIOD

'.t')9E'·Il·.'OEQVN'P···t(l,a;~Td<l:tbl

Khoz: Standard horizontal seismic

intensity at the center of gravity of the

structure, the values of which are

derived from Table 5.3.2, Table 5.3.3

depending on the importance of the

structure and the soil type.

However, KhZ shall not be less than 0.3.

The standard horizontal seismic intensity

(KhZ) at the ground surface must be: 0.7

(upper limit) - 0.6 (lower limit), 0.8 - 0.7, 0.6

0.4 respectively for Type I, Type II, and Type

ill subsoil classifications.

2. When taking vertical seismic intensity (K,Z),

the equation is:

5-12

3. If there is a possibility that the seismic motion

is largely amplified by such irregularities of

ground as the titled ground surface, the design

seismic intensity shall be increased by 1.2

times at maximum.

These guidelines were decided to be designed

horizontal seismic intensity and acceleration

response spectrum by the following methods.

CD Maximum acceleration on the ground

surface.

Using the seismic motion records

which can be accepted as based on

engineering standards (Kobe University

[NS EW]; East Kobe Ohashi [GL-33m,

N78E, N12W]; Port Island (GL-83m,


Nl2E), the ground surface response

acceleration was derived by the

equivalent linear method with

considering the flexibility dependence of

modulus of transverse flexibility

(modulus of rigidity) and dampening

coefficients based on total 150 points of

boring data from the Hanshin District.

Based on these results, divided into three

ground types: Type I ground (TG < 0.2s),

Type IT ground (0.2 < TG < 0.6s), and

Type III ground (0.6 < TG) according to

its natural period and asked the

generation frequency distribution of each

ground type. With this results, set non

over probability 90%, 70% and decided

the upper and lower limit values of the

surface acceleration. For example,

surface acceleration of Type I ground

600-700 gal is equivalent to non-over

probability of 70%-90%. Further, the

intensity is derived by dividing ground

surface acceleration by gravitational

acceleration and it is used to measure for

anti-seismic structural design by seismic

intensity method and judgment of the

liquidation. This design horizontal

seismic intensity must be set between

upper limit value and lower limit value of

the significance rank of the facilities.

CD Structural acceleration spectrum.

Before mentioned generation

frequency distribution of the structural

acceleration response spectrum

5-13

(acceleration response value) of the

ground surface seismic intensity which is

gathered from approximately 150

checking sites were surveyed at each I, IT,

and ill type grounds. Same as the

design horizontal seismic intensity

described before, the value equivalent to

the acceleration response value was

derived at each 90% and 70% of non

over probability and was shown in

Figure 5.3.1 x 1. As same with design

horizontal seismic intensities, the value

equivalent to the acceleration response

value was set within the limits of the

significant ranking of the facilities.

Also, concerning the area which periodic

zone is above LOs (Type ill ground must

be above 1.5s), on the logarithmic graph

the spectrum value is shown as a

'declining straight line. Also, 0.1 second

spectrum value has set to coincide to the

maximum ground surface acceleration

value as shown in (1). Also, the

response spectrum shows a result of

attenuation coefficient 5% and if the

structural attenuation coefficient may

differ from this result, the acceleration

response value will be assumed to

reverse proportion of the 2 root of

attenuation coefficient, and correct the

spectrum value. For example, in the

case of an acceleration response value of

10%, it may become 5% of the value:

$/.JW = 0.707 (times).


III

10.t 15T1WCTl:W:AL ~UmJI'.AL PROPEJ'SOr

(CtntmtiT,(1'r;"'Hf)u.l>ln:CGN\)'?<r!O~;llHi,.cTd

HGtlRfS~3+ .ACCELf.r",~Tl'o;~rRES?ONS£S?EC'ffiUI>;(,

($ElSMICMOT!ON LJ-:V'l::L2)

I

STRtiC1'UK41. NATr.7R.tiJ,.PROPF.l{TY(S)SECQNlJ1"'ll{,,: GROU1<iD FOUND,,:rl0N

(h) (O,2S;>:T".<O.6.s)

@ Standard Horizontal Seismic Intensity

(Khoz)·

The standard horizontal seismic

intensity (Khoz) is derived from dividing

the acceleration response value at the

center of gravity of the structure by

gravitational acceleration. Based on the

Figure 5.3.1 the result of formulating

upper and lower limit values of the

standard horizontal seismic intensity is

shown in Table 5.3.2 and 5.3.3. The

Upper and lower limit values are

equivalent to the bore mentioned at 90%

and 70% of non-over probability. This

means depending on the water supplier's

judgment on the significant degree of

water works facility may reflect choice.

@ Design Horizontal Seismic Intensity

(Khz)'

The design horizontal seismic intensity

(Khz) is derived by multiplying the

standard horizontal seismic intensity

(Khoz) with the structural characteristic

factor (Cs) ' This structural

characteristic factor (Cs) is derived by

multiplying the dampening characteristic

(D). Figure 5.3.1 shows the response

spectrum derived at with a structural

attenuation coefficient of 5%. If, the

structural attenuation coefficient differs

from this value, it can be corrected and

D '1 can be derived from this graph.

Further, the D'1 value is considered the

structural flexibility factor. Thus, it is

5-14


(K' hOI) and (KhOl) shall be derived using Table

intensity at the objective depth may be

derived by linear interpolation between KhOl

and K'hOI'

2. The design horizontal seismic intensity, when

applying Seismic Motion Level 1, shall be

determined as follows:

1) The design horizontal seismic intensity at

ground surface

s; = Cz ' K hOI

2) The standard horizontal seismic intensity

equivalent with factors which are used

for calculating equivalent horizontal

seismic intensity in "Road Bridge

Specifications. " This is defined as

follows:

D = J5hJh

D = 117 ~1+417

5.3.4. The standard horizontal seismic

't'lTL _VVUC>1C>

h =attenuation coefficient (%)

7] =durability ratio

The structural characteristic factors

(Cs) can only be used for seismic motion

Level 2. They cannot be applied to

seismic motion Level 1 anti-seismic

construction design.

5.3.4 Seismic Intensity Used in Anti-Seismic

Design by the Seismic Intensity Method

for Buried Structures


1. When anti-seismic design for buried

structures is carried out using the seismic

intensity method, the standard horizontal

seismic intensity shall be determined

employing the standard horizontal seismic

intensity (K' hOI) at the base ground level

assumed for the design and the ground surface

seismic intensity (KhOI)' The values of

5-'15

at the base ground surface

s ; = Cz ' K'hOi

Where:

C; Region-specific correction factor.

Values are 1.0-0.7.

3) When considering the vertical design

seismic intensity (KVI )

KY1 = Kh/2.

·.·rAJHES,3L4·STANDARlJHORlZ0f4i1;j\n.SlEtSMlCIN1""Et4Sfl')'{UEVEDllWH1£ff.jS·.TJSED•..FnR't$i$ .•$tJE~D$TRtle'TURA:EDEsIG1'j'.·.BiX· .1fJ3,$ ••$EISMlY.·.INTlU-iSIlT'Y·••METH0D

..... $TAh1)AiRril-to~6NS1AlmAlU)ffd~f1jtfGROIl'1>J"DTi:PB SmlSM1CIh'1'ENSlT'i': •.• .• .•....·N$l'm"· ...CLASSJElCATlom


When the buried structures are designed by the

seismic intensity method, the standard horizontal

seismic intensity which will act on the buried

structure can be considered as the standard

horizontal seismic intensity at the center of the

gravity of the structure.

Also for the underground standard horizontal

seismic intensity will be assumed that it will

change linearly between the base ground of the

anti-seismic design and ground surface.

Therefore, it will be obtained the value at the

center of the gravity of structure by the linear

3. If there is possible amplification of seismic

motion due to such irregularities of the ground

as tilted ground surface, the design seismic

intensity shall be increased by 1.2 times at

maximum.

Similar to seismic motion Level 1, the

design horizontal seismic intensity, which acts

on buried structures, may be acceptably

derived using linear interpolation at the

structure's center of the gravity. Here the

design horizontal seismic intensity Kh2 is not

interpolation. necessarj when considering the structural

characteristic factor.

1. For anti-seismic design of buried structures,

whose response characteristics during an

I

't'fPR dMJUND·t1C<C.2J ..

,VHE1'l:E'l'G isGRDlfNf.1.·NA'ttlH.i\L'?ERJOD(s)

TYPEll

[O.Z1iil'c<t't<i]

Tl'Ps.m

[t.?~5:,$:T~:J

tAaH15.35DESIGNfORHORlZONTAI..>SElSMIG ..rNTENSITY{S£lSMiCM01'lONLEVU2JWJ:liCHU$ED FORmJRISD$t~VC:nJltAL.hEsl(iN(SE1SMICtN'Tt~i$nY METHOD

5.3.6 Seismic Motion Level Used in the

Response Displacement Method for

Buried Structures


1. In the case anti-seismic design for an Buried

structure is carried out by the seismic intensity

method, the design horizontal seismic

intensity shall be determined by the design

horizontal seismic intensity (K 'h2) at the base

ground surface used for anti-seismic design

and the ground surface seismic intensity (Kh2) .

The values of (K 'h2) and (Kh2) shall be deri,:ed

from Table 5.3.5. The design horizontal

seismic intensity at the objective depth may

be determined by linear interpolation between

Kh2 and Kh2 '.

2. In the case vertical seismic intensity (Kv2) is

taken into account, shall be set as follows:

KV2 = Kh12.

5.3.5 Design Seismic Intensity Used in Anti

Seismic Intensity Method for Buried

Structures


5-16


earthquake are chiefly affected by movements

of surrounding ground, the response

displacement method shall be used.

Cross-sectional force, stress, strain, etc.

working on the structures shall be computed

based on the displacement or deformation of

the ground. The ground displacement

amplitude to be generated under Seismic

Motion Level 1 shall be derived by the

following formula at the distance x(m) from

the ground surface.

u,(x) =-;-SvTGK~l cos~ (5.3.4)Jr ~11

Where,

Uh (x): the horizontal displacement amplitude

(m) of the ground at the depth x from the

ground surface.

Sv: seismic motion velocity response spectrum

(cmls) of the ground per unit seismic

intensity.

TG: the natural period(s) for the surface layer

of the ground.

K 'h1: the design horizontal seismic intensity at

foundation ground surface where the

design is based (Refer to 5.3.4 Seismic

Intensity Used in Anti-Seismic Design by

the Seismic Intensity Method for Buried

Structures (Seismic Motion Levell))

H: the thickness of surface ground layer (m)

ill the case the vertical response

displacement amplitude is taken into account,

the following formula is used:

1U =-U

v 2 h

Velocity response(s) per unit seismic intensity

is derived using Figure 5.3.2 according to the

basic natural period for the ground surface layer.

Figure 5.3.2 shows the maximum relative

velocity generated by modeling a system with one

degree of freedom for the natural period TG'

Reduction of the constant hG (20%) for the surface

ground layer was accomplished. With this

derived value together with seismic records

expanded the maximum velocity to 1.0 g.

1 H)

Nt:rtJEAV·.r:'E1HOn.·(l'e)(sJOP·SUl'l.%'iGEe..·I'."~H".E•••• "...,. 2·i CRQtJND ROb NDA'1'HJNF:~·~..f>.".~-",,;j. .. ,",_. ,'" .': , _ ' -" , _.' _." .. " •. <

SPEED RE:'H>0NSESFECTRUMfORCONSTRUGT1QUI]E$fOW·i$El$~tlC M01'JON·LKVE:L2'l

5.3.7 Seismic Intensity Used in Design of

Buried Structures by the Response

Displacement Method


Similar to the case of Seismic Motion Levell,

like Buried structures, anti-seismic design of

structures whose response characteristics during

an earthquake are chiefly affected by displacement

of surrounding ground, the response displacement

5-17


method shallprincipally be used.

Cross-sectional force, stress and strain, etc.

working on the structures shall be computed

based on the displacement or deformation.

The ground displacement amplitude generated

under Seismic Motion Level 2 is derived by

the following formula at the distance x(m)

from the ground surface.

2 I 1lXU'; (x) = 7r 2 S; To cos 2H (5.3.5)

Where,

U; (x): the horizontal displacement amplitude

ground surface.

Sy: seismic motion velocity response spectrum

(cm/s) [See Figure 5.3.3]

To: the natural period(s) for the surface ground

layer.

H : the thickness of the surface ground layer (m)

When the ground vertical response

displacement amplitude Uv is considered, the

formula is:

1U =-U

y 2 h

If there is possible amplification of seismic

motion due to such irregularities of the ground

as tilted ground surface, the design seismic

intensity shall be increased by 1.2 times at

maximum.

These records were from the 1995 Hyogo

ken Nanbu earthquake. These records took into

account five wave forms obtained from ground

5-18

foundation and rock bed surface observations

within 20 km from the Hyogo fault. Figure

5.3.3 represents the velocity response spectrum

obtained from the acceleration response

spectrum of the ground surface. Engineering

judgment was added. Figure 5.3.3 shows two

different kinds of values - 200 cm/s (upper

limit) and 70 cm/s (lower limit) - as the

maximum response velocity. The system was

modeled with one degree of freedom for natural

periods above 0.7(s). Each of these values is

compatible to a probability not exceeding 90%

and 70%. The desigu value is increased or

decreased within the scope of the upper limit

and the lower limit, according to significance

rank of the structure.

r 10

FA'1'tiJLAiL l"B.RIOD(n:;){flFOR•.S.tJf\.f,....c·~:GR0lJNDrPUNDATI0N

EARTHQUAKE RESISTANT DESIGN CODES IN JAPAN January,2000

5.3.8 Seismic Motion Input Used in Design

Using the Dynamic Analysis

The seismic waves used for dynamic analysis

must fit the founding ground surface velocity

response spectrum is shown in Figure 5.3.3, the

ground surface acceleration response spectrum is

shown in Figure 5.3.1, or the seismic waves

observed in the vicinity of inland faults such as

ones caused by the 1995 Hyogo-ken Nanbu

earthquake.

When selecting seismic wave observation sites

for dynamic analysis against seismic motion Level

2, the ground types for the sites must be well

considered. In particular, whether or not the

observed seismic wave response spectrum is

similar to the design response spectrum in Figure

5.3.1 must be check. The maximum value of the

inputted seismic wave for dynamic analysis must

be for a ground surface that is 6,000 - 7,000 cm/s"

and 400 - 600 cm/s2 against the first ground type,

second ground type, and third ground type.

Similarly, the base ground must be 400-500 cm/s".

5.4 Geotechnical Surveys, Ground Displacement,

and Ground Distortion

5.4.1 Primary Subjects of Geotechnical Survey

For anti-seismic design of water supply

facilities, geotechnical survey at locations, where

construction works are situated, depending on the

importance of the facilities.

5-19

Soil surveys here include all surveys related to

topography, geology, ground, and soiL

Generally, less damage due to earthquakes is

found on good ground, that is firm and uniform

ground. Therefore, water works facilities must

be required to be built on such stable ground.

The following are not good ground conditions:

CD Sliding;

® Mountainous slope toes and slope shoulders;

@ Slopes;

@ Different soil layer interfaces;

@ Weak ground;

@ Reclaimed ground;

(J) Ground subject to fluidization or lateral

floating during an earthquake.

1. Survey using existing records

Rough soil conditions at the facility

construction site can be studied.

2. Common soil survey

Study of required items for construction

planning and earthquake resistance of

facilities will be conducted.

3. Survey of dynamic properties of soil

The physical properties of soil are

represented by the N value. Cohesion, C,

and the internal friction angle 1>, are for

static behaviors. However, the velocity

effect of stress to the constants of the

ground and the effects of stress during an

earthquake must be discussed. For these

studies, the following constants shall be

determined.

EARTHQUAKE RESISTANT DESIGN CODESIN JAPAN January, 2000

1) Modulus of dynamic distortion;

2) Attenuation coefficient;

3) Dynamic poison ratio;

4) Dynamic shear strength.

4. Survey of dynamic physical properties of

the ground

1) Velocity of elastic wave;

2) Ground predominant period; and

3) Other.

The geotechnical survey methods shall be based

on the following:

1. Follow the standard or criteria which are set

forth in the Japan Industrial Standard (JIS)

or the Japan Geology Society (JGS) for

various survey and laboratory test.

2. In principle, measurements shall be actually

conducted for dynamic soil constants and

dynamic physical properties of the ground.

When it is impossible to do so, they may be

obtained from the results of other surveys.

5.4.2 Methods of Geotechnical Survey Vfu-lOliS test-methods and soil Constants related

to ground and soil are shown in Table 5.4.1.

01

oo0 .:

~.

iOi fJ10.10o

o

oo

o

o

i

Jh7Klli/OR.BODY StAWIC'1'ESTS UNIJl...'t.J'AL.C(tMPRESSI0t'iSMfPtt"1C•.'l'E);.r·· ·.·.Tlo:;sT'l"~lA.."{lAX.

.CDM:?RESSrO.NTh:S't~•. DlRECtSflEAroNGT$tirt

..... ·lixiNA.lV!lC'l'Rb\XIAt

j. COMPR~S$!ON'T:m$T .··.·PYNi><MIC••stMPLl:l.·.. sHE~il:lNGT£b'"1)

.•.•..RE.so.N.• AJ.'<f.bE. M.z.nc•..•. H.OD....

.",.",_.."-__....;.. ~..,,.•;TgSTVIER;.4Trot>l·1'EST

O~.!stHE't!i:ST.:tl:ES't.r..;T •.PI~tC'fl,;';{REQ~rES<tED{)RUSEFORnlscussrON

O·M,A1U{.mlNPH~EGTW;'.REQ!1ES'!Eb··oa··tJSES··!N·.PtSCtlSS10N

'ON:l..;Y·':l'HEJ~ONG1'1'tml.N¥'.L ""~'l':.rE. S\?EEO\'tIl:;.L.BRDERrilED··{)NLY'THES.H!EAR!NG ELAS'I'ICCOEl1'FICIEN'T:\'IlLLJUlDERr\I1i:!)

5-20


5.4.3 Soil Liquefaction and Lateral Spreading

Liquefaction of the soil is a phenomenon

whereby sandy soil loses its strength and rigidity

rapidly and the whole body of soil behaves like

liquid.

Since the soil liquefaction causes damage to

water supply facilities such as flotation of buried

structures, and subsidence and/or tilting of other

structures, anti-seismic design with due

consideration to such aspects must be provided.

In the ground near the embankment bordering a

reclaimed land and slopil1.g ground, a phenomenon

of lateral movements, of liquefied soil may occur

and may damage foundation of structures and

water mains. For the examination of anti

seismic safety of such lateral soil movements shall

be taken into account. Judgment on the

possibility of soil liquefaction shall be made if the

soil possesses all the following conditions:

(1) Saturated soil layer thinner than 25m from

the ground surface.

(2) Average grain size Dso is less than lOmm.

(3) Content by weight of small grain particles

(soil grain size of less than 0.075mm) is less

than 30%.

5.4.4 Ground Displacement and Ground Strain

Caused by Liquefaction

In the ground near the embankment or sloping

ground, there is a possibility of lateral spreading

due to liquefaction. For anti-seismic design

buried pipelines for water supply, ground

5-21

displacement and strain due to such lateral

spreading must be taken into account.

Large-scale lateral movement of a revetment,

caused by an earthquake, is possible in reclaimed

areas. This occurs when the tensile strain of the

ground, in a direction perpendicular to revetment

line, is in the range of 1.2 - 2.0%. Figure 5.4.1

shows the frequency of ground strain occurrence

100m from the revetments in the Hyogo-ken

Nanbu and Niigata earthquakes. Based on this

distribution, the probability was calculated and

tabulated in the Figure. ~~ ground strain value of

1.2 - 2.0% was obtained from the 70% and 90%

non-exceeding probability. For anti-seismic

design of underground pipelines, an appropriate

ground strain may be selected within this range,

depending on the pipeline's degree of importance

and difficulty in restoring.

5.4.5 Ground Strain at the Incline of

Artificially Altered Ground

In the case, the surface of artificially altered

ground (such as in a housing estate) is inclined,

displacements downwards along the slope may

occur during a severe earthquake with such

seismic intensity as the Seismic Motion Level 2.

The effects of such ground displacement must

be taken into account for anti-seismic design of

buried pipeline.

Ground strain for inclined ground (non

fluidized) during the Seismic Motion Level 2 is


·0$1%Q)7Q$4

according to the degree of fluidization.

and foundation structures is reduced when ground

is fluidized. Reducing ground reaction greatly

effects the behavior of structures during

earthquakes. The ground reaction coefficient

and ground friction force must be reduced as

shown in Table 5.4.2. This must be done

account. The types of ground subject to

investigation are: valleys filled with ground,

ponds, and embankments with more than 10%

average slope.

filt'5lJRE ·,s.4L..lC~0~*bitg#Stka. brST(n~!T1QN F'RtQttt>CY bl$TRlIHPTlQNNBAR BFL1;l4B;Att:<EJCHBORHOQDARE:A

within a range of 1.0 - 1.7%.

Anti-seismic design for buried pipelines for the

Seismic Motion Level 2 must be taken into

5.4.6 Reduction in Reaction Force and Ground

Friction Force due to Soil Liquefaction

If there is a possibility of soil liquefaction, the

ground reaction force coefficient for the design of

buried pipeline and foundation structures must be

reduced according to the degree of liquefaction.

Ground reactions which act on buried pipelines

5-22


SCOP2bF$AJ:'WtYRAT!t'}j"'LACAU1Sl' ••LIQJJlW1>H;nONGEMEHAtllJN

ntlrt:H1I'B0MGThOlJ:ND S'l;JE'f'AC~m}

'~=-,,~---'---;""'-+~ ,.;..,.:.........=-"f-..---~-+=-"=-"~.-.-+-- ..........~........., .........._=-"~

5.5 Soil Pressure During an Earthquake

5.5.1 General

For anti-seismic design of structures attached to

the earth, the soil pressure during an earthquake

shall be determined according to the following:

1. The horizontal soil pressure during an

earthquake must be derived by the Mononobe

Okabe soil pressure formula.

2. In case vertical seismic intensity for the

surcharge load during an earthquake, the

surcharge load must be multiplied by (1 +Kv).

5.5.2 Calculation of Horizontal Soil Pressure

During on Earthquake

For calculation of the horizontal soil pressure 5.5.3 Calculation of Vertical Soil Pressure

during an earthquake, the cohesiveness of soil, if During an Earthquake

any, shall be taken into account.

The vertical soil pressure on buried pipeline

1. Soil classification for earth pressure must be calculated taking into account, the

calculation. For soil classification and for influence of lateral friction, if any.

various numerical soil values of earth pressure,

refer to Table 5.5.1.

5-23


5.5.4 External Pressure due to Lateral

Spreading

On the ground, which may be subject to lateral

spreading due to liquefaction, anti -seismic design

of foundation structures must be carried out with

consideration to the external force caused by such

spreading. In this case, the influence of inertia

force from the super-structure and the base

structure don't have to be considered.

Great concern about the external pressure

created by lateral ground flow exists, especially,

with regards to water works facilities built on

suspect ground. Anti -seismic structural design

must consider earth and flow pressure.

It is shown in the experiments that fluidization

flow pressure (which acts on the buried structure)

in the liquefied ground layer is below 30% of the

total load pressure.

The lateral flow of the external pressure is

stated in Figure 5.5.1.

5-24

5.5.5 Buoyancy Generated by Soil Liquefaction

ill case the liquefaction resistance coefficient,

FL , refer to Explanation of 5.4.3 (Soil Liquefaction

and Lateral Spreading) of soil surrounding such

buried structures as pipeline is smaller than 1.0,

the safety of the structure in regard to buoyancy

shall be examined.

Specific gravity of fluidized soil is 18 - 20

kN/m3 (1.8 - 2.0 X 10-3 kgf/cnr'). If the actual

specific gravity includes the content volume or

it will become smaller than this value and the

buried structure will have a tendency to balloon.

The upper portion of the non-fluidization layer,

the weight of the road surface pavement materials,

and the shearing resistance will usually block out

the floating up. However, past examples

(Niigata earthquake, etc.) illustrate that floating up

bad broken pipelines or manholes. Careful

examination is necessary.

5.6 Hydrodynamic Pressure During an

Earthquake and the Water Sloshing

5.6.1 Hydrodynamic Pressure During an

Earthquake

For anti-seismic construction design of

structures that come into contact with water,

dynamic water pressure during an earthquake

must be considered.


Structures which contact water (such as a dams,

water tanks, etc.) and are subject to an earthquake

must be considered. These structures receive

dynamic water pressure during an earthquake.

The action of dynamic water pressure during an

earthquake must take into account two factors: (1)

whether free surface water is present and (2)

whether the complacability of the water can be

ignored.

Dynamic water pressure action created during

an earthquake can be dived into two factors: (1)

inertial action which interacts proportionality with

secondary dynamic water pressure generated by

free surface water oscillation. Generally, the

inertial force of dynamic water pressure

interaction is more significant and, therefore, will

be taken into account by the design. The action

of surface water oscillation is a supplemental issue

for dynamic analysis.

The complacability of water, with regards to

structures like water tanks and water intake towers

in water works facilities, can be ignored without

creating problems. However, for pipeline

structures, the complacability of water must be

considered. It is not, an excessive load for the

design may result.

5.6.2 Water Sloshing

For anti-seismic design of water tanks, water

sloshing must be considered when necessary.

For water tanks with free surface water,

5-25

sloshing is induced during an earthquake. The

effects of sloshing bring about overflow or impact

pressure against the roof.

Whether such sloshing cause damage, or not, it

depend on the close relationship between the

natural period of water sloshing in the tank and

the periodic characteristic of the seismic motion.

The sloshing of water inside of the tank shall be

checked by following methods.

a: Response spectrum method based on the

potential theory.

b: n wave response method.

c: Response spectrum method based on the

potential theory.

However, when the competent seismic wave

has inputted, dynamic response analysis is

acceptable.

5.7 Safety Check

5.7.1 Combination of Loads

Structure safety in anti-seismic calculations

must be checked by combining the normal load

(dead weight and live load at ordinary times) and

seismic effects.

5.7.2 Safety Check of the Structures Fabricated

with Steel, Concrete, etc.

For safety checks of structures fabricated with

concrete, steel bars, structural steel pre-stressed

concrete(pC) etc., the following related standards

must be used.


Specifications for Highway Bridges (Japan

Road Association);

Concrete Standard Specifications (Japan

Society of Civil Engineers);

Iron Sluice Valve Technology Standard (Iron

Sluice Valve & Pipe Society).

5.7.3 Safety Check of Pipeline in their Anti

Seismic Calculations

As a general rule, safety of pipeline during an

earthquake must be checked with consideration to

I the strength and flexibility of the pipeline.

A pipeline structure for a water works facility

varies in types. If roughly categorized, the

following two types would emerge:

1. Jointed pipeline structures - Here, most of

the flexibility is dependant on the joint.

2. Continuous pipeline structure - Here, most

of the flexibility is dependent on material the pipe

is made of. The anti-seismic calculation method

for the direction of principal buried pipelines is

described in this edition of the guidelines.

Anti-seismic ability is checked using the

response displacement method. This method is

based on the behavior of the pipeline. This

behavior is generated through the relative

displacement of pipeline and the ground.

The pipelines, which possess the characteristics

of (1), are represented by ductile iron pipe. The

pipeline which possess the characteristics of (2)

are represented by steel pipe. The basic concept

of the safety check on pipelines with these

characteristics is summarized in Table 5.7.1. For

either the seismic motion Level 1 or seismic

motion Level 2, the pipeline component stress will

not exceed the allowable stress of the pipe

materials. With jointed pipeline structures under

live loads and under ordinary conditions, the

jointed component expansion capacity will not

exceed the maximum expansion capacity of the

design. This is the main point for anti-seismic

checking.

With safety checks against seismic motion

under live loads, must basically be below yield

point stress for the pipe component material.

Distortion, which corresponds with the yield point

stress, is:

E = (J IE =2,400/2,100,000 =0.11 %

After field condition are completely considered,

appearances seem better, since distortion of the

pipe component is below 23t/D (%) (about 0.15

0.20)% and the anti-seismic capability can be

checked. Here, t is the pipe thickness and D is

the diameter of the pipe. With seismic motion

Level 2, the distortion of the component, even

considering the stationary free load, is below

46tID (%) (about 0.3 - 0.4)%. The anti-seismic

capability can be checked.

5-26


5.7.4 Safety Check of the Foundation Ground

in its Anti-Seismic Calculations

As a general rule, safety of the foundation

ground in anti-seismic calculations must be

checked in accordance with "Supporting Ground

and Allowable Bearing Force".

5.7.5 Safety Check of Foundation, Earthen and

Retaining wall in Anti-Seismic

Calculations

As a general lule, safety check of foundation,

earthen structures, and retaining wall in anti

seismic calculation must be checked in

accordance with "Anti-Seismic Calculation

Methods for Foundations" and "Anti-Seismic

Calculation Methods for Earthen Structures and

Retaining Wall".

5.7.6 Safety Check in Anti-Seismic Calculations

in Consideration of Critical State under

Seismic Motion Level 2

Safety check in anti-seismic calculations in

consideration of critical state must be carried out

using the following rules:

1. Based on the results of proper analyses or

testing the anti-seismic safety of structures

must be checked with reference to the

critical state found in such analysis and

testing.

2. In anti-seismic design based on the critical

state, tenacity of structures must be secured

5-27

so that no plastic yield shall occur until the

structures have reached te critical state.

3. For the anti-seismic design based on the

critical state, an appropriate safety factor

must be employed with reference to the

critical displacement.

6. RECOMMENDED PRACTICES FOR EARTHQUAKE RESISTANT

DESIGN OF GAS PIPELINES (DRAFT)

JAPAN GAS ASSOCIATION

6.1 Introduction

6.2 High-Pressure Gas Pipelines

6.2.1 Basic Policy on Earthquake-Resistant Design

6.2.2 Earthquake-Resistant Design against Seismic Motions of Level 1

6.2.3 Earthquake-Resistant Design against Seismic Motions of Level 2

6.3 Medium- and Low-Pressure Gas Pipeilnes

6.3.1 Basic Policy on Earthquake-Resistant Design

6.3.2 Earthquake-Resistant Design Procedure

6.3.3 Design Ground Displacement

6.3.4 Ground Condition

6.3.5 Pipeline Capability to Absorb Ground Displacement

6.3.6 Allowable Strain and Allowable Displacement

6.4 Appendix

6.4.1 Earthquake-Resistant Design ofHigh-Pressure Gas Pipeline

6.4.2 Improvement of Earthquake Resistance of Pipelines

6.4.3 Block System of Pipeline Networks

6- 1

6- 1

6- 1

6- 3

6- 4

6-17

6-17

6-17

6-17

6-19

6-20

6-22

6-24

6-24

6-29

6-29


6.RECOMMENDEDPRACTICESFDREARrHQUAKE-RESfSTANTDESIGNOFGASPIPELINES(DRAFT)

JAPAN GAS ASSOCIATION6.1 Introduction

The presently used "Recommended Practices

for Earthquake-Resistant Design of Gas Pipe

lines" was established as the recommended

practices for earthquake-resistant design of

high-pressure gas pipelines (See Appendix

6.4.1.) and medium- and low-pressure gas

pipelines in March 1982, after the Miyagiken

Oki Earthquake (June 1978),

The Hyogoken-Nanbu Earthquake occurred

in January 1995. Since the earthquake far

exceeded conventional theory, the Central Dis

aster Prevention Council reviewed its Basic

Plan for Disaster Prevention and the Japan

Society of Civil Engineers presented a proposal.

These actions showed the necessity for and

concept of containing the recommended prac-

.tices for the earthquake-resistant design of

important structures in methods of design for

seismic motions of a higher level, level 2 seis

mic motions, which correspond to the shocks

generated by the Hyogoken-Nanbu Earth

quake in the Kobe District.

The gas utilities are also now revising the

Recommended Practices for Earthquake

Resistant Design of Gas Pipelines in the high

pressure gas pipelines section, mainly for the

purpose of improving the resistance of high

pressure gas pipelines to seismic motions of

level 2, especially in the concept of design in

put seismic motions. This revision is aimed at

achieving a more carefully-formulated respon

se to advanced seismic needs worldwide in the

light of technological findings since the pre

sently used Recommended Practices were es

tablished 17 years ago. Regarding this re

vised edition of Recommended Practice for

6-1

Design of High Pressure Gas Pipelines, be

cause its official issue may be after the publi

cation of the English version, it is hoped to

recognize it as based on a "Draft" of the revised

recommended practice.

The presently used Recommended Practices

for Earthquake-Resistant Design of Gas Pipe

lines has not been revised in the medium- and

low-pressure gas pipelines section, since it has

been confirmed that the recommendations

therein are reasonable for earthquake

resistant design, judging from the results of

investigation of the Hyogoken-Nanbu Earth

quake.

6.2 High-Pressure Gas Pipelines6.2.1 Basic Policy on Earthquake

Resistant Design

(1) Basic Concept of Earthquake-Resistant

Design

For the earthquake-resistant design, two

levels of seismic motions are assumed to se

cure the earthquake-resistant performance

specified for the respective levels of seismic

motions in principle.

(Description)

(a) The Basic Plan for Disaster Prevention of

the Central Disaster Prevention Council

was reviewed based on the Hyogoken

Nanbu Earthquake which occurred on

January 17, 1995, and it now stipulates

that the earthquake-resistant design of

structures, facilities, etc. to be constructed

in the future shall not suffer any serious

loss of function even should general seismic

motions with a probability of occurring once


or twice within the service life of the pipe

line occur, and shall not have any serious

influence on human life even should a

higher level of seismic motions of low prob

ability occur, due to an inland type earth

quake or trench type huge earthquake.

(b) For the earthquake-resistant design of gas

equipment, two levels of seismic motions

are assumed, and considering the influ

ence of structures, facilities, etc. on -human

life, the influence on relief activities and on

the prevention of secondary disasters, and

the influence on economic activities, gas

equipment must have earthquake-resistant

performance suitable for its respective

kinds and degree of importance.

(c) Based on the above basic concept; earth

quake-resistant design is performed to se

cure the earthquake-resistant performance

required for the two levels of seismic mo

tions, as described in the following chapter.

(2) Seismic Motions to be Assumed for

Design, and Earthquake-Resistant

Performance

The seismic motions to be assumed for de

sign, and the earthquake-resistant perfor

mance required of them are shown in Table

6.2.1.

Table 6.2.1 Seismic Motions and EarthquakeResistant Performance

Seismic Motions to be Assumed Earthquake-Resistan tfor Design Performance

SeismicGeneral seismic motions Operation can be

motionswith a probability of resumed immediately

ofoccurring once or twice without any repair.

level 1during the service life ofgas pipeline are assumed.

Very strong seismic mo- The pipeline does nottions due to an inland leak. though de-

Seismic type earthquake or formed.motions trench type earthquake

of likely to occur at a lowleve12 probability rate during

the service life of gas

pipeline are assumed.

6-2

(Description)

(a) Seismic Motions of Level 1, and Earth

quake-Resistant Performance against

Them

[Seismic Motions]

Seismic motions specified in the previous

Recommended Practices for Earthquake

resistant design of High Pressure Gas Pipe

lines (March 1982).

[Earthquake-Resistant Performance]

The earthquake-resistant performance re

quired for the seismic motions of level 1 is

such that "Operation can be resumed imD:1e-

diately without any repair." based on the Re

port of the Committee for Preventing Seismi

cally Caused Gas Disasters.

(b) Seismic Motions of Level 2, and Earth

quake-Resistant Performance against

Them

[Seismic Motions]

A proposal concerning the seismic standard,

etc. of the Japan Society of Civil Engineers

presents concrete images as "seismic motion

near the hypocenter fault of an earthquake

caused by any internal strain of a plate of

magnitude 7 class (hereinafter called an in

land type earthquake)" and "seismic motion

in the hypocenter region by a large-scale in

ter-plate earthquake occurring near land

(hereinafter called a trench type earth

quake)", and the present "Recommended

Practices" assumes the seismic motions of

these two earthquake types; inland type

earthquake and trench type earthquake.

Further, even if there -is no active fault

found in the existing documents, there is a

possibility that an inland type earthquake

may occur. Thus, it was decided to adopt a


concept that a lower limit level is set when

seismic motions are assumed.

[Earthquake-Resistant Performance]

The earthquake-resistant performance re

quired for the seismic motions of level 2 is

such that "the pipeline does not leak, though

deformed." based on the Report of the Com

mittee for Preventing Seismically Caused

Gas Disasters.

(3) Evaluation of Earthquake-Resistance

Since seismic motions repetitively forcibly

displace the pipeline, the fatigue damage at

a very low frequency caused by them is

evaluated for earthquake-resistant design,

When the ground of the planned pipeline

is likely to be greatly deformed by liquefac

tion, etc., it must be examined adequately.

(Description)

The method for evaluating earthquake

resistance was decided, considering that seis

mic motions have the following characteristics:

a) the loads are short-term ones, and

b) since the strains (or relative displacements)

caused in the ground by seismic motions are

repetitively applied to the pipeline, the loads

are periodically displacement-controlled, and

also in reference to the concepts of existing

standards(ASME Sec. III, etc.) which specify

these loads.

6.2.2 Earthquake-Resistant Design

against Seismic Motions of Levell

The earthquake-resistant design against

seismic motions of level 1 is performed ac

cording to the Recommended Practices for

Earthquake-resistant design of High Pres

sure Gas Pipelines (Japan Gas Association,

6-3

March 1982)*. However, for the "apparent

propagationvelocity of seismic motion", the

value stated in "Apparent wavelength of

seismic motion" is used, and for the "ground

spring constants in the axial direction ofthe

pipe and in the transverse direction of the

pipe", the values stated in "Confiningforce

of ground" are used.

* See Appendix 6.4.1.

(Description)

For earthquake-resistant design against

seismic motions of levell, Recommended Prac

tices for Earthquake-Resistant Design of High

Pressure Gas Pipelines* (Japan Gas Associa

tion, March 1982) is applied.

However, the following portions among the

latest results of research concerning the

earthquake-resistant design, especially among

the findings obtained after the 1995 Hyogo

ken-Nanbu Earthquake inclusive should also

be applied, in view of their nature, to the.

earthquake-resistant design against seismic

motions of level 1. So, for the following val

ues stated in the 1982 Recommended Practices,

those stated in the present Recommended

Practices are used.

(1) "Apparent propagation velocity of seismic

motion" in "Design seismic motion"

(2) "Ground spring constants in the axial di

rection of the pipe and in the transverse di

rection of the pipe" in "Earthquake

resistant design of straight pipe in uniform

ground", "Earthquake-resistant design of

straight pipe in roughly varying Ground"

and "Earthquake-resistant design for bend

and tee".



against Seismic Motions of Level 2

(1) Entire Flow ofEarthquake-Resistant

Design

(a) The procedure for setting the design seis

mic motion is shown in Fig. 6.2.1.

(b) The earthquake-resistant design flow

based on the set design seismic motion is

shown in Fig. 6.2.2.

(2) Setting of Design Seismic Motion

[A] Procedure and Method for Setting

Design Seismic Motion I, II and III

The design seismic motion is set as fol

lows based on "[B] Investigation of active

fault" and "[C] Judgment as to existence of

active fault".

1) When it has been concluded that the ex

istence of any active fault is positive:

· The seismic motion obtained by multi

plying the design seismic motion I stated

in "[D] Design seismic motion I" by the

seISmIC zone coefficient stated in "[G]

Seismic zone coefficient" is used as the

design seismic motion.

· Alternatively if fault analysis can be per

formed, the seismic motion calculated ac

cording to the fault analysis stated in "[F]

Design seismic motion III" is used as the

design seismic motion. However, if the

calculated design seismic motion is smal

ler than the seismic motion obtained ac

cording to the procedure of 2), the seismic

motion of 2) is used as the design seismic

motion.

2) When it has been concluded that the exis

tence of any active fault is negative:

· The seismic motion obtained by multi-

6-4

plying the design seismic motion II stated

in "[E] Design seismic motion II" by the

seismic zone coefficient stated in "[G]



3) When it has been concluded that the exis

tence of any active fault is unknown:

. The seismic motion obtained by multi

plying the design seismic motion I stated

in "[D] Design seismic motion I" by the

seismic zone coefficient stated in "[G]



(Description)

(1) The seismic motion of level 2 to be applied

for design is set using any of the three kinds

of seismic motion described below based on

the conclusion as to whether the existence

of any active fault is positive or negative.

Design seismic motion I: Seismic motion

decided for the inland type earthquake

based on the observation records of

Hyogoken-Nanbu Earthquake

Design seismic motion II: Seismic motion

decided for the trench type earthquake

based on past earthquake observation

records

Design seismic motion III: Seismic motion

based on analytical decision for the in

land type earthquake by modeling the

hypocenter fault and using the hypocen

ter parameter and the information on

the ground and physical properties of

propagation routes

(2) If it is concluded that the existence of any

active fault likely to greatly affect the

planned pipeline is positive, it can be con-


sidered to analytically calculate the seismic

motion by modeling the hypocenter fault

and using the. fault parameter and the in

formation onthe ground and physical prop

erties of propagation routes (this method is

called fault analysis). However, presently

the data necessary for analysis and the

analytical method are not sufficiently es

tablished. Therefore, the design seismic

motion is set by using the design seismic

motion I decided based on the observation

records of Hyogoken-Nanbu Earthquake,

one of the recent largest inland type earth-

quakes, or by fault analysis.

(3) When it has been concluded that the exis

tence of any active fault is negative, it is re

quired to take only the trench type earth

quake into consideration, and the design

seismic motion is set using the design seis

mic motion II for the trench type earth

quake.

(4) When it has been concluded that the exis

tence of any active fault is unknown, the

design seismic motion is set using the

above-mentioned design seismic motion I,

from the viewpoint of obtaining conserva-

tive results for design, since it cannot be

concluded that there is no active fault.

Investigation of active fault nearthe design site (B)

Negative

Design seismicmotion II (E)

Selection of seismiczone coefficient (G)

Corrected designseismic motion II

Design seismicmotion I (D)

Selection of seismiczone coefficient (G)

Corrected designseismic motion I

Decision of designseismic motion

No

Positive

Yes

Design seismicmotion ill (F)

* 1) If the design seismic motion III is smaller than the corrected design seismic motion II, the correcteddesign seismic motion II is used as the design seismic motion.

Fig. 6.2.1 Design Seismic Motion Setting Flow

6-5


Fig. 6.2.2 Earthquake-Resistant Design Flow for High Pressure Gas Pipelines againstSeismic Motions of Level 2

Design Seismic Motion I or II(set based on earthquakeobservation records)

Design Seismic Motion III(set by fault analysis)

.Apparent Wavelength of Seismic Motion:L=V·TV;Apparent propagation velocity of seismic motion

.Natural Period of Ground of Surface Layer

4.H :EVsjOH j:T=-=-, V s =--'---

v, H

H ; Thickness of ground of surface layer (m)

Vs ; Shear wave velocity in the ground of surface layer (m/s)

Ir Elastic wave survey xC""", Sand 0.7 E . 0.6L Clay 0.7 E • 0.85Estimate from N value -.:::::::::: Sand 0.7 E • 6NO.2!

Clay 07 E12 • NO·078

V(rn/s)

(2.5,800)

(0.15, 100)

T (s)

.Ma:cimum Velocity in the Ground of

Surface Layer at Design Site (at bur

ied depth of gas pipeline): v

Maximum ground displacement: Uh

.Apparent Horizontal Propagation Velocity of Wave: V

a. Apparent propagation hodograph

b. Calculation of simple phase velocity

c. Detailed analysis (Haske] matrix method, etc.)

To calculate according to any of a, band c.

.Ground Displacement of Surface Layer

1tZ. T· Sv : cos-

aa.Ground Strain

V(cm/s)

(0.1,8.0)

(0.7, 100)f ~7,50)

v ; Seismic zonecoefficient

z ; Buried depth ofpipeline (m)

Sv; Standard responsevelocity (cm/s)

(0.1, 4.0) T (s) .

.Ground Strain of Uniform Ground

: E Gl=2 1t X UhlL

• Ground Strain of Irregular Shallow

Ground: EG2= IE G12+ EG/

E G3: Ground strain caused by irregular

shallow ground

(* *)

6-6

No


.Strain Transfer Coefficient

~ ; Ground springconstant in axialdirection of pipe

Design of Straight Pipe

Ground strain due to (*) or (**) E G

q. Coefficient considering sliding between

pipe and ground

Al = ~ KIE'A

• Strain of Pipe caused by earthquake

E p = a • E G (a • E G < E y)

: E p = E G (ex • E G ~ e y)

E y; Yield strain of pipe material

Design of Bend and Tee

Ground displacement due to (*) or (**) Uh

.Displacement Transfer Coefficient

a * = q* • aa

q* ; Coefficient considering sliding

between pipe and ground

Relative displacement between pipe

and ground: 6. = (1- a *) . Uh

In the case of irregular shallow ground,

the value at or near the place where the

bend or tee is installed is used.

.Strain of Bend or Tee during

Earthquake

E B,T= f3 B,T • 6. (f3 B,T~ 1.27 Ey)

E B,T= C· f3B,T ·6.(f3 B ,T > L 2 7 E y)

f3 B,T ; Coefficient of conversion

C ; Plastic state correction factor

.Allowable Strain: Allowable strainof straight pipe,bend and tee 3%

6-7

No

Examination of Design Modification


of any active fault is "positive", "negative" or

"unknown" can be made in reference to Ta-

(Description)

(1) The conclusion as to whether the existence

....Conclusion Criterion"Positive" · It is judged that "The existence of

any active fault likely to producelarge seismic motions is positive."Fig. 6.2.3 shows the relation be-tween the distance from an activefault and the magnitude of anearthquake.

"Negative" · It is judged that "The existence ofany active fault likely to producelarge seismic motions is negative."Fig. 6.2.3 shows the relation be-tween the distance from an activefault and the magnitude of anearthauake.

"Unknown" · It is not confirmed that there is noactive fault in a plain covered with athick sedimentary layer.

·A complicated earth structure isformed with boundaries of threeplates gathering underground, as inthe metrooolitan area.

ble 6.2.

Table 6.2.2 Criterion for concluding that the

existence of any active fault is "positive",

"nezative" or "unknown"

(2) The boundary line of Fig. 6.2.3 is obtained

by calculating the weak ground conditions

with a ground surface velocity of 64 cmls as

the boundary on the conservative side. If

the shortest distance from the active fault

concerned to the planned pipeline and the

magnitude of the earthquake likely to be

caused by the active fault exist on the left

side of the boundary line, the ground sur

face velocity caused at the planned pipeline

when the active fault aets is larger than 64

cm/s. If they exist on the right, the ground

surface velocity is smaller than 64 em/s.

The surface ground velocity of 64 cm/s was

obtained by converting 50 cm/s, which is the

For investigation of any active fault, the

information concerning the position, prob

ability, activity; etc. of any inland active fault

likely to produce large seismic motions to the

planned pipeline is collected from existing

documents.

[B] Investigation ofActive Fault

(Description)

(1) For any inland active fault, basically, the

active faults belonging to probabilities I and

II of "Active Faults in Japan (New Edition)"

are investigated for comprehensive evalua

tion also in reference to the active fault list

stated in "Investigation and Observation

Plan for Foundations Relating to Earth

quakes", the earthquakes assumed in the

regional disaster prevention plan and other

findings in the latest investigation and re

search results.

(2) If any active fault found as a result of ac

tive fault investigation is found not to be

imminent in activity and not to act during

the service life of the pipeline, it can be ex

cluded from the investigation.

[C] Judgment as to the Existence of Active

Fault

Whether the existence of any active fault

likely to give large seismic motions to the

planned pipeline is "positive", "negative" or

"unknown" is concluded by taking the fol

lowing into consideration:

(1) Distance of the planned pipeline from the

active fault

(2) Magnitude of earthquake estimated from

the length ofthe active fault

6-8


The design seismic motion I is shown in

'.00.:5 J.O 2.D

.7-;50)0

V

(O.l.U30.1 0.2

_.Th~ _d~Sign seismic motion II is shown inIl!'lg. (j.~.5.

300

(Description)

(1) The design seismic motion I was decided by

obtaining the velocity response spectrum on

the seismic base rock (engineering under

ground base rock) based on 16 observed

waves of two horizontal components at the

hypocenter region and nearby (within 10 km

from the active fault) eight sites of the

Hyogoken-Nanbu Earthquake, considering

the non-excess probability.

[E] Design Seismic Motion II

Fig. 6.2.5 Velocity response spectrum of as

sumed trench type earthquake

Natural period of ground of surface layer T (5)

(Description)

(1) The design seismic motion II was set at one

half of the design seismic motion I, in refer

ence to the two spectra.

V Earthquake-resistant design Course,

Highway Bridge Specifications and Descrip

tion (December 1996). Earthquake-resistant design (draft), Design

Standard and Description of Railway Struc

tures, Etc. (November 1998)

[F] Design Seismic Motion III

II II

v

{O.1.I.Q

3Or--+-74--+++f+H--+-+-+-I

I I I I I II II I I I IL I

"Positive' I II

, II I I I

, I I

I 1.-1" 'Negative" .I II I 11111 11111I I 11111 1111IVI I I I

c-,

-.; a>c -o B> ~

"'r~'ec _Co >~er;~

6

~I . .

~s5 I I Io J0 20 30 40

Fig. 6.2.4.

8

response velocity of design seismic motion II

caused by the trench type earthquake speci

fied in "Design seismic motion II", into the

ground surface velocity (50 x 4J 7[ =64, 4J 7[:

coefficient for converting the response of

single-degree-of-freedom system into the re

sponse of continuum).

The shortest distance from an active fault, d (km)

Fig. 6.2.3 Criterion for concluding whether

the existence of any active fault likely to pro

duce large seismic motions is positive or

negative

(3) As an example of the methods for estimating

the magnitude of an earthquake, Matsuda

proposes the following formula:

LoglOL =a.6M - 2.9

L: Length ofthe active fault

M : Magnitude of an earthquake specified

by Meteorological Agency

[D] Design Seismic Motion I

I ,

I I I0..3 1.0 1.0 5.0

Natural period of ground of surface layer T (5)

Fig. 6.2.4 Velocity response spectrum of

assumed inland type earthquake

The design seismic motion III is calculated

by fault analysis.

(Description)

(1) If the seismic motion calculated by fault

6-9


Vs : Shear wave velocity in the ground of sur

face layer (rn/s)

n Vs; Shear wave velocity ofJ"" Vs - H j-th layer (mJs)f;:: j j ~: Thickness of j-th layer

H (m)

analysis is smaller than the corrected design

seismic motion II caused by the trench type

earthquake at the planned pipeline, the cor

rected design seismic motion II is used as

the design seismic motion.

[G] Seismic Zone Coefficient

(1) The zone classification is the same as the

classification specified in the Recom

mended Practices for Earthquake

Resistant Design of High Pressure Gas

Pipelines (Japan Gas Association, March

1982).

(2) The seismic zone coefficient is the value

stated in Table 6.2.3 for each zone.

Table 6 2 3 Seismic Zone Coefficient

Zone Classification Seismic Zone Coefficient

Special A Zone 1.0

A Zone 0.8

B or C Zone 0.7

(Description)

Fig. 6.2.6 shows the zone classification map

for the seismic zone coefficient.

(3) Ground Displacement and Ground Strain

of Surface Layer

[A] Natural Period of Ground of Surface Layer

The basic natural period of ground of surface

layer is obtained from the following formula:

4- HT=~ where

Vs

T: Natural period of ground of surface layer(s)

H: Thickness of ground of surface layern~ -_. .

(=LH j) (m)j=l

[

_ Special A Zone

m AZone

§ BZone

o .CZone

Fig. 6.2.6 Zone Classification for Seismic Zone Coefficient

6-10


[B] Apparent Wavelength of Seismic Motion

The apparent wavelength of seismic mo

tion in the direction along the ground sur

face is obtained from the following formula:

L= V' T

where L : Apparent wavelength of seismic

motion in the direction along the

ground surface (m)

V: Apparent propagation velocity of seismic

motion (m/s)

T: Natural period of ground of surface layer(s)

The apparent propagation velocity of

seismic motion is obtained from Fig. 6.2.7.

of Section 6.2.3 (2) [D]

T: Natural period of ground of surface

layer(s)

z : Buried depth of pipeline (m)

H: Thickness of ground of surface layer (m)

(2) When design seismic motion II is used, it

is obtained from the following formula:

U =~. T' v· S (T)' cos ( JrZ )h Jr 2 VII 2H

where SVII(1): Response velocity of design

seismic motion II (cm/s), according to

Fig. 6.2.5 of Section 6.2.3 (2) [E]

'rnp ntnpr i'lvmhnli'l Hrf~ HI'l snecified for (1)- --- - ----- -oJ ----- --- --- - -- - - ...- - - " ,.

Natural period of ground of surface layer, T (s)

[C] Ground Displacement of Surface Layer

Fig. 6.2.7 Apparent propagation velocity of

seismic motion

where E G1 : Ground strain of surface layer

in the case of uniform ground

v: Seismic zone coefficient, according to

where E GIlD: Ground strain of surface

(3) When design seismic motion III is used,

the ground displacement of the surface

layer at the buried position of the pipeline

is directly calculated.

The ground strain of surface layer in the

case of uniform ground is obtained as fol

lows:

(1) When design seismic motion I is used, it


( 1rZ)E G1 =V • E GIO • cos 2H

6.2.3 (2) [G]

E GIO : Ground strain of surface layer of

design seismic motion I in the case of

uniform ground, according to Fig. 6.2.8

(2) When design seismic motion II is used, it


( 1rZ)E Gl =V • E GIlD • cos 2H

[D] Ground Strain of Surface Layer in the Case

of Uniform Ground

5.02.01.00.50.2

I ! i II III I I !I,I i

I 1111

. i II I I u,I

! i. :I I I , II / ! !

! I ! i I!! I ./ I I ii i I ! 10 I ! II

I Viii' I I ii ' ; I ! I

/11 I

III1i

I II I: !,:

1=(0.15,1~

I ,! I , I , :50

0.1

200

500

100

1000

3000

2000

The ground displacement of surface layer

is obtained as follows:

(1) When design seismic motion I is used, it is

obtained from the following formula:

U =~. T' v· S. (T)' cos ( JrZ )h Jr 2 r I 2H

where U;,: Ground displacement of surface

layer (em)

v: Seismic zone coefficient, according to Sec

tion 6.2.3 (2) [G]

SVI(T): Response velocity of design seismic

motion I (cm/s), according to Fig. 6.2.4


Natural period of ground of surface layer, T (5)

adopted.

(c) The ground strain of irregular shallow

ground is taken into account when the angle

of inclined base rock is 5 or more.

In the case of irregular shallow ground, a

ground strain larger than that in the uniform

ground can happen, and this must be taken

into account for earthquake-resistant design.

(4) Ground Strain ofIrregular Shallow

Ground

[Description 1J Ground Strain of Irregular

Shallow Ground

(a) The ground strain caused in irregular shal

low ground is calculated by superimposing

the ground strain of uniform ground on the

ground strain caused by inclined seismic

base rock.

cG2 =.Jg2 G1 + c2G3

cG3 = n -0.3 (%)

where E G1: Ground strain of uniform ground,

according to 6.2.3 (3) [D] "Ground strain

of surface layer in the case of uniform

ground"

E G2: Ground strain caused in irregular shal

low ground

E G3: Ground strain caused by inclined seismic

base rock

n: In the case of corrected design seismic

motion I: v (seismic zone coefficient, ac

cording to Section 6.2.3 (2) [G] )

In the case of corrected design seismic mo

tion II: 0.5 x v

(b) As the ground strain of uniform ground E ci.

the ground strain of uniform ground at the

position where the surface layer thickness

becomes maximum at the irregular shallow

ground portion or that at the position where

it becomes minimum, whichever is larger, is

5.0

I I5.0

I2.0

2.01.0

0.5 1.0

0.5

! ! ! ! !110.2

0.2

, , ,, , , , ,0

, !\Qj.o:~i

, ,!

i , ,i I L..-ri !'N..,!

lA ! i i i i !!~ ! i i' , i i !. ,

nVCQr-°.l1i I i

,

Ii I .....i i i i , i

I I (2..S.1i.16jI ! i ! ! ! Il i

(0.1.0.102 , , , ,, ! , , ,5 i I i i ! ! !

! i i I ! i i i ! i

I j 1 i ! I i i i ! i! I i

I i I i ! ! !!! I ! !2 i ,

! tI i !i

J i iI ! i i i ! 1

,, , , ,.50

, , !

i j !j i i i i ! ! ! i

i.(O.7,0~!

,j !i ! ! i

20'~. I*" i i

,I I : I I i !

(0.15.0.~ ! i i ! I i I

1n , l l ] i-, , , -~, ,

! ~.o.~)--+-05 i i '.J i

(0.1.0.05 11 i i i ! !,

!! i

, , ! i I j ,, i ,, , , I i· i ! I ! ! ii i ,

i ! ii, I

02 I,

i i II!i Ii i! i I II

II

i ! i i i i01 ! i

0.20

0.01.-1-1---'---'---'---'---'-.............__'----'-....o.........I0.1

".§0; __ 1.0U)~:::~

.~] 0.5" :lr" 0::: ,.~o ~

~ E» ....~ ,~

- cg ~ 0.1~~- OJ- '"rJJ '" 0.0'00:: ~.~ -=-= .5rJJ _ 0.0

~ g.:l .o ~

o:E

Fig. 6.2.8 Ground strain of surface layer

of design seismic motion I in the case of

uniform ground

layer of design seismic motion II in the

case of uniform ground, according to Fig.

6.2.9

.~"'g 00; :l<:: 0

'""' .....;: CJo s:::.... - 0~ ~ .c: .-- s:::

: ~ 0o ......~ a

.... <:::l '"~ <:3 D.o OJs:::..l:.- ~

~ .5rJJ:::: D.-:: c;; .!:!e "'0 o.

CJ :E 0.1

Natural period of ground of surface layer, T (s)

Fig. 6.2.9 Ground strain of surface layer of

design seismic motion II in the case of

uniform ground

(3) When design seismic motion III is used,

the ground strain of the surface layer at

the position of the buried pipeline is di

rectly calculated, including the influence

of irregular shallow ground.

6-12


The confining force a of ground in the

[B] Confining Force of Ground in the

Transverse Direction of Pipe

prpe 18

/AY/

direction of a

IHo=LSm~~o

transverse

(O.15kgflcm~

Ground spring constant: k, =6.0N/cm3

(O.6kgficmS]

The confining force of ground in the axial

direction of a pipe is approximated by bi

linear expression using the critical shear

stress 'C cr per unit surface area of the pipe

and the ground spring constant k 1, or ob

tained by measurement.

Critical shear stress: 'C cr =1.5N/cm2

(5) Confining Force of Ground

[AJ Confining Force of Ground in the Axial

Direction of Pipe

r crapproximated by the bilinear expression or

the straight line using the initial gradient of

the bilinear expression, using the maximum

confining force (J cr of ground per unit pro

jected area and the yield displacement i5 cr,

or obtained by measurement.

Relative Displacement 0

Table 6.2.4 Confining Force of Ground in the Transverse Direction of Pipe byDiameter

Maximum Confining Force of Ground Yield Displacement k2 = a crt 6 crDiameter (mm)

a cr N'cm" (kgf/crrr') 6 cr cm Nzcm" (kgf/cm'')

(Typical example) 750 32 (3.2) 3.0 11 (1.1)

c

C::lo...""c.e a crcco...

<.2

""c

Straight line

Bilinear expression

: 0' crD (Outer diameter)

Relative displacement O.

6-13


(6) Earthquake-Resistant Design of Straight

Pipe

[A] Strain of Straight Pipe Caused by

Earthquake

The strain of a straight pipe caused by an

earthquake is obtained from either of the

following formulae:

(1) If the strain of the straight pipe is in the

elastic range, that is, if a . E G':::: E s» then

(2) If the strain of the straight pipe is in the

plastic range, that is, if a • E G > E y' then

where E p: Strain of the straight pipe caused

by earthquake

0:: Strain transfer coefficient of the straight

pipe, according to (B) of the following

section.

E G: Ground strain, according to and

E y : Yield strain of the pipe material

[Description]

(1) If the strain of the straight pipe exceeds

the buckling limit, the strain of the pipe af

ter buckling is calculated, for example, using

FEM analysis with buckling behavior taken

into account. The buckling limit is the

buckling initiation strain E buckle (%) speci

fied in the Recommended Practices for

Earthquake-resistant design of Gas Pipe

lines.

where t: Wall thickness of the pipe (em)

Dm : Average diameter of the pipe (em)

[B] Strain Transfer Coefficient

The strain transfer coefficient of a straight

pipe is obtained from the following formula:

6-14

a = q. ao

where a: Strain transfer coefficient of the

straight pipe (for the strain transfer coeffi

cient a of the straight pipe, the same for

mula as used in the Recommended Practices

for Earthquake-Resistant Design of Gas

Pipelines is used.)

a 0: Strain transfer coefficient of the straight

pipe without sliding taken into account

~: Ground spring constant in the axial di

rection of the pipe per unit length of pipe

line [N/cm2 (kg£'cm~], according to Section

5.4.1.

L: Apparent wavelength of seismic

motion(cm)

E : Elastic modulus of the pipe [N/cm2

(kgficmZ)], E =2100000 kgf/crrr'

A: Sectional area of the pipe (em")

t : Wall thickness ofthe pipe (em)

r G: Sear stress acting on the pipe surface

[Nzcm" (kg£'cmZ)]

t cr: Sliding initiation critical shear stress

when sliding occurs between the pipe

and the surrounding ground [N/cm2

(kgficmZ)]

q: Sliding reduction coefficient

q ::: 1- cos ~ + Q. -(; - ~) sin ; ,

q = arcsin ( :: J' q s 1

rG'::::r cr q=l

Q: Correction factor for evaluating q on the

conservative side, 1.5


(7) Earthquake-Resistant Design ofBend

and Tee

A bend or tee may be greatly strained de

pending on the piping conditions, and this

must be taken into account for earthquake

resistant design.

[Description 1] Strain of Bend Caused by

Earthquake

The strain of a bend caused by an earthquake

is obtained from either of the following formu

lae or by FEM analysis,

(1) If the strain of the bend is in the elastic

range or partially plastic range, that is, ifflB6- ~ 1.27&y' then

&B = flB6-

(2) If the strain of the bend is in the full plasticrange, that is, if flB6- > 1.27&y' then

&B =CBflB6-

where EB: Strain of the bend caused by earth

quake

{3 B: Coefficient of conversion of the bend

(Vern), according to Description 4.

zl: relative displacement (em), according to

Description 3

Ey : Yield strain of the pipe material

GB : Correction factor for the strain ofthe bend

in the full plastic range

GB =2 (below 600A)

GB =1 (over 600A including 600A)

[Description 2] Strain of Tee Caused by

Earthquake

The strain of a tee caused by an earthquake

is obtained from either of the following formu

lae or by FEM analysis.

(1) If the strain of the tee caused by an earth

quake is in the elastic or partially plasticrange, that is, if flr6- ~ 1.27&y' then

6-15

Gr = fJr6-

(2) If the strain of the tee caused by an earth

quake is in the full plastic range, that is, if

flr6- > 1.27&y' then

&r = 2flr6-

where E r: Strain ofthe tee caused by

earthquake

{3r: Coefficient of conversion of the tee (Vern),

according to Description 5.

zl: Relative displacement (em), according to

Description 3

Ey: Yield strain of the branch pipe adjacent to

the tee

[Description 3] Relative Displacement between

Pipe and Ground

The relative displacement between a pipe

and ground is obtained from the following for

mula: Ll =(1- a*)o Uh

where Ll: Relative displacement (em)

Uh: Ground displacement of the surface layer

(em)

ex * : Coefficient concerning relative displace

ment between pipe and ground

a* =q * 0 a o

ex 0: Strain transfer coefficient of the straight

pipe without sliding taken into account

q*: Sliding reduction coefficient concerning

relative displacement

( 2 ~2)q*=sin~ 0 1+~ -2 -~ ocos~, q*::;l

rG~rcr q=l

where ~ ~ arCSin( ::) Furthermore,

r G: Shear stress acting on the pipe surface

[N/cm2 (kgf/cm2) ]


r or: Sliding initiation critical shear stress

when sliding occurs between the pipe and

the surrounding ground [N/cm2 (kg.fi'cm~]

[Description 4] Coefficient of Conversion of

Bend

The coefficient of conversion of a bend is ob

tained from the following formula:

2iB A12

D/(5 + R1~11 + 413115(1 + b2 ) - bIIfiB = -3

10A+5L12 (1+b 2)+10Ab3

1+ 2R1 + (Jr - 2}nR 2 12

• _ ~? -;;-2 / • "\ ~ -.. -;;-3b = l-LnlCA -~4-Tr)nlCA

2 {1 + RI~ + TrnRI + (4 -Tr)nR 2I2}

3-3{Tr Tr] ( ])b3 = nR A - + 2 + 1- --2 b,2 2nAR nAR

{+( 2_+!!.-+ Tr] 2)b,}RA 2 2nAR

where [3B: Coefficient of conversion ofthe bend

(l/cm)

iB: Stress index for the bending load of the

bend, obtained from the following formula:

i - 1.95 or 1.5, whichever is larger

B- (~~r/3

n : Flexibility factor of the bend, obtained from

the following formula:1.65

n = (~~)

A: Sectional area of the pipe (crrr)

R: Radius of curvature (em)

I: Moment of inertia (em")

D: Outside diameter of the pipe (cm)

L: Apparent wavelength ofseismic

motion(cm)

-,{If;A: V4t-

11;.: Ground spring constant in the transverse

direction of the pipe per unit length [Nzcm"

(kg.fi'cm~]

E: Elastic modulus [N/cm2 (kgflcm~]

[Description 5] Coefficient of Conversion of Tee

The coefficient of conversion of a tee is ob

tained from the following formula:

fJT = 42;2D]A2~: 1)

4A 2 + LI]A] C

where the subscripts for D, A, land 1. express

the following:

Subscript 1: Branch pipe side

Subscript 2: Main pipe side

[3 T: Coefficient of conversion of the tee (l/cm)

D: Outside diameter (em)

A: Sectional area (crrr')

I: Moment of inertia (ern")

L: Apparent wavelength of seismic

motion(cm)

I: V~l11;.: Ground spring constant in the transverse

direction of the pipe per unit length

[N/cm2 (kg:flcm~]

E: Elastic modulus [Nzcm" (kgflcm~]

(8) Allowable Strain

The allowable strain of a straight pipe,

bend or tee is 3%.

[Description]

(A) Allowable strain on the seismic motion of

Level 2 was determined based on the damage

caused by the cyclic ground displacement of

the extremely low cycle. Regarding the

number of the cyclic ground displacements,

6-16


by taking the same concept as the Recom

mended Practice for Earthquake-Resistant

Design of High Pressure Gas Pipeline (1982.

3), the repetition number of times of the

maximum strain was determined to be"

equivalent to the fatigue damage which one

seismic motion of Level 2 gives to the pipe

line. As it is enough if one seismic motion of

Level 2 occurs during the design lifetime of

the pipeline, the number of the cyclic ground

displacements to be considered on the seis

mic motion of Level 2 are approximately 3 to

5 times.

Setting the allowable strain in the light of

the fatigue design curve of ASME, the allow

able strain of the base metal is 3% if assumed

the repeated times of 3 to 5. It can be con

sidered in general that the strain of 3%

doesn't impede the operation and it has

enough safety margin from the viewpoint of

the experimental data and the performance

of the steel pipe.

(B) Buckling is allowed because it doesn't lead

to leakage directly. But in the case that

there is possibility of strain occurrence to

cause buckling on a straight pipe, in other

words, the case that the occurred strain ex

ceeds the initial buckling strain specified on

the seismic motion of Levell, 35 . tJDm (t :

pipe thickness (em), Dm: average diameter of

the pipe (cm), the strain which occurs on the

pipeline after buckling should be calculated

correctly by the method such as the finite

element method (FEJ\.1).

6.3 MEDIUM-AND LOW-PRESSURE GAS

PIPELINES

6.3.1 Basic Policy on Earthquake-Resistant

Design

(1) General Principles

Earthquake-resistant design for medium

and low - pressure pipelines is aimed at

achieving greater pipeline flexibility and there

by reducing gas pipe leakage or breakage.

(2) Quantitative Flexibility Evaluation

Method for Pipelines

Aseismic strength is judged by calculating

the capability of the pipeline to absorb the

stipulated ground displacement. If the value

exceeds the design ground displacement de

termined by ground and other conditions, the

pipeline is judged to be earthquake-resistant.


Procedure

The procedure is shown in Fig. 6.3.1.

Evaluation of earthquake resistance is based

on the following items.

CD Selection of burying conditions

@ Calculation of design ground displace-

ment

@ Calculation of pipeline ground displace-

ment absorption

@ Selection of ground displacement input

@. Selection of standard strain and standard

displacement

@ Evaluation of earthquake resistance

6.3.3 Design Ground Displacement

The design ground displacement for

evaluating pipeline flexibility is determined by

the following formula.

. 1) Horizontal displacement (in axial direction

of pipe) : U = a 1a2UO

2) Vertical displacement (perpendicular to

6-17


Desig-n Ground Displacements Canabilitv to Absorb Ground Displacement

~ Allowable Limits I

Horizontal U = a 1a 2 Uo

Vertical V = 1/ 2U

in which

a 1= Seismic zone factor

a 2 = Factor according to the

combination of pipeline

type and ground

conditions

Input Ground

Displacement Models

. Horizontal

Displacement

. Vertical

Displacement

Designing

Pipings

. Straight

Pipings

. Pipings with

bends, branche

etc.

I

Uo=Standard design ground

displacement

. Allowable strain

(e0)

. Allowable

displacement

(00' eo)

Evaluation of Capability to

Absorb Ground Displacement

Simple formulas

[. Nume~cal calCulatiOn]

. Expenment

,

I L\u and L\v I

- Evaluation of Flexibilitv

Su » U

L\v> V

I

Fig. 6.3.1 : Flow Diagram of Earthquake-Resistant Design

of Medium - and Low - Pressure Pipelines

6-18


pipe axis): V =1/ 2UIn the formula, a] is determined by the

seismic zone factor in Table 6.3.1 of which the

division of area is the same as that shown in

Fig. 6.3.1 in 6.4 Appenclix 6.4.1.

Table 6.3.2 shows that a 2 is a factor repre-

senting the combination of pipeline type and

ground condition.

Uo is determined as 5.0 (em) in standard

design ground displacement.

The ground condition type in Table 6.3.2 is

based on "6.3.4 Definition of Ground Condi-

rion."

6.3.4 Ground Condition

Ground conditions are determined the state

of the ground in the general area where piping

is installed and by the piping installation's

geographic location.

I . Area formed by any of the following ground

types or areas where the three are found in

combination

(1) Soil layer dating back to the Triassic E;a.

or earlier (hereinafter called "rock layer")

(2) Diluvium layer

(3) Alluvium layer less than 10m thick or

layer in which soft layer is less than 5m

thick.

* Provided there exists a rock layer or firm

diluvium layer (N) 50, seismic wave veloc

ity of more than 300m/sec.)

II. Area formed chiefly by alluvium layer of

more than 10m or soft layer of more than

illa. Mixture of soil layer equivalent to Condi

tion I and a layer equivalent to Condition

IT, or are in which the two types are mixed

illb. Border are between soil layer and sturdy

structure built upon foundation equivalent

to Condition IT and other locations where

displacement is evidently discontinuous

Table 6.3.1: Seismic Zone Factors (a l )

SA1.0

A

0.8B

0.6C

0.4

Table 6.3.2: Factors according to the combination ofthekind of pipeline and ground conditions (a2)

~Classification I II illof Pipeline

Medium pressure A0.9 1.3 1.8(3 ~ P< 10kgflcmZ)

Medium pressure B0.7 1.0 1..4(1 ~ P< 3kgflcmZ)

Low pressure (main)0.5 0.7 to(P< lkgflcmZ)

Low pressure (service)0.7 1.0 1.0(P< Lkgf/crn")

6-19


6.3.5 Pipeline Capability to Absorb Ground

Displacement

(1) Capability of Straight Piping to Absorb

Ground Displacement in Axial Direction

(1w)

The capability of a straight pipe to absorb

ground displacement in the axial direc

tion(!::..u) under ground conditions I, IT, and

Illa , as shown in Fig. 6.3.2 is a ground dis

placement that can be absorbed by the pipe at

a displacement input that focuses on one

point on the ground surface.

i) A pipeline with continuous restraint force

from projection in axial direction

[a] Reduced elastic modulus model (for poly

ethylene pipe, etc.)- 2

!::..U = AE£o [mm]wr

where, A: Area of cross-section (mnr')

D : Pipe diameter

E : Reduced elastic modulus (Nzmnr)

r : Restraint force of ground per unit sur

face of pipe (Nzmm")

e 0 : Allowable strain (specified in Section

6.3.6)

[b] Elastoplastic calculation model (welded

steel pipe)

AE {£} +.11.(£02

- e v2

) }

Su =----'--------:....!....

trDre v : Yield strain of pipe

e0 : Allowable strain of pipe

E : Elastic modulus (N/mmZ)

}.E : Tangent modulus of pipe

ii) Piping showing localized reduction in ten

sile stress on cross-section (such as steel

pipe with screwed joint)

6-20

F2

!1u=--JiDrAE

where, Fa: Allowable tensile strength

of screwed joint portion

iii) Piping with mechanical joint

Su = 00 +2 (01 +02+·····+0n)

Where, °0 is the maximum displacement

ofjoint in the center of ground displacement,

at which leakage or serious damage ofjoint is

expected. 51' °2 , ••• , On represents allow-

able displacement (slipout) in joints adjoining

the joint in the center, calculated taking into

account the reduction in load due to the

ground restraint force between the joints.

The capability of a straight pipeline fixed at

one end in Ground Condition lib to absorb

axial ground displacement is ground dis

placement that can be absorbed when the

input of ground displacement that concen

trates at the border of a structure and ground

is added, as shown in Fig. 6.3.4.

(2) Capability of a Straight Piping to Ab

sorb Ground Displacement in Direction

Transverse to Axis

The capability of straight piping to absorb

ground displacement in the direction trans

verse to its axis (!1v) in Ground Condition I ,

IT, or IIJa is ground displacement that the

piping can absorb when transverse displace- .

ment concentrates on one point on the ground,

as shown in Fig. 6.3.5.

i) A pipeline with homogeneous rigidity along

its axis (steel pipe with welded joint

or polyethylene pipe)

2.fie'" ~4El!1v= --£D kD 0

Where, E : Reduced elastic modulus

(N/mm~


f:,u Ground Displacement~ ......lJ>

I I

Restriction of Soil--.. --.. --.. --.. -+ --..

Pipe

Fig. 6.3.2 : Ground Displacement Input for Ground Conditions I, II, and IDa

1~__- ~----=-_...... AE

O"v

Fig. 6.3.3: Bilinear Elastoplastic Model of Steel Material

Ground Displacement

--v>c....

Ground Restraint Force--..-+-+-+--..-+

I>:?i

~

~

'-:

~~r---;:;U-----:=-----r-------------~

.3:r.;

Fig. 6.3.4 : Ground Displacement Input on Piping Fixed at One End in Ground Condition IDb

~~-------------r"" -~~-.--.~--- .. -.... -..

--_ ....... - .......... -_ ..... - .. --- .... -_.,I,

.............................................. .. ..'I

Fig. 6.3.5: Ground Displacement Input in Transverse Direction Under

Ground Condition I, II, or ID a

6-21


1 : Moment of inertia of cross-section (mm4)

k : Reduced coefficient of subgrade reaction

(Nzmm")

ii) Piping with localized drop in strength

against bending moment (steel pipe with

screwed joint)

.fieJr

/

4~4EI

Llv= --MEI kD 0

Where, Mo : Mome.it atthe location of

localized drop in strength (N . mm)

E : Elastic modulus (N/mm~

The capability to absorb ground displace

ment when the pipe is fixed to structure un

der Ground Condition J:Ifu, as in Fig. 6.3.6, is

displacement that the pipe can absorb when

displacement concentrates at the border of

the structure and ground.

(3) Capability of 3-D piping to Absorb

Ground Displacement (Llu)

The capability of 3-D piping system com

prised oflow - pressure service and internal

pipes under Ground Condition I, Il , orma

is ground displacement that the piping can

absorb at the displacement shown in Fig.

6.3.7.

The absorption capability of a 3-D piping

system buried under Ground Condition Illb

and fixed at one end to a structure is ground

displacement that can be absorbed when the

ground displacement shown in Fig. 6.3.4 is

applied.

6.3.6 Allowable Strain and Allowable

Displacement

(1) Allowable Strain in Pipe Material (£0)

and Elastic Modulus (E)

The Allowable strain (£0) that is set over

6-22

the plastic limit and the reduced elastic

modulus (E) applied when calculating the

material's ability to absorb ground displace

ment, which depends upon the material, are

shown below.

1) Steel pipe: Allowable strain .... £0=3 [%]

Reduced elastic modulus

.... E =3.0X 104 [N/mm2J

2) Ductile cast-iron pipe

: Allowable strain.... £0=2 [%]


.... E =3.0 X 104 [Nzmm'']

3) Polyethylene pipe

: Allowable strain .... £0=20 [%]


.... E =3.0 X Hf [N/mmZ]

When, however, reduced elastic modulus is

inapplicable for steel or ductile cast-iron pipe,

Young's modulus that is within the range of

elasticity is applied.

Steel pipe: 2.1 x lOS [Nzmm']

Ductile cast-iron pipe: 1.6 X lOS [N'mnr']

Coefficient A used to determine the tan-

gent modulus (AE ) used to calculate elastic

ity of steel pipe is founded upon the following:

--1 =7.1 X 10-3

(2) Allowable Displacement for Mechanical

Joints and Expansion Fittings

Standard displacement for expansionjoints

such as mechanical and flexible joints for

connecting pipes in ways other than welding

is the official value specified under JIS or

other equivalent standards. If no nominal

value is found, it is determined as the dis

placement that removes airtightness or inflic

ts serious damage or deformation upon a ma

jor part of the joint.


f.................. ".. .6.v V

Fig. 6.3.6 : Ground Displacement Input in Transverse Direction for Piping

Fixed at One End Under Ground Condition ill b

Location of Ground Displacement Input

<;Pwad ~ ~l

Residential

LandI I

Gas Meter

Main or Service Pipe Service Pipe

Crank Pipe

Internal Pipe

a) Location of Ground Displacement Input

Service Pipe Element Internal Pipe Element

b) Division of Service Pipe and Internal Pipe Elements and

Displacement of Each Element

Fig. 6.3.7: Ground Displacement Input for Service and Internal Pipe System and

Calculation of Ground Displacement Absorption Capability (Sample)

6-23


6.4 Appendix

6.4.1 Earthquake-Resistant Design of

High-Pressure Gas Pipeline

(1) Basic Concept of Earthquake-Resistant

Design

A Recommended Practice for Earthquake

Resistant Design of High-Pressure Gas Pipe

line is based on greatly improved concepts with

regard. to the evaluation of seismic motions

and interaction (slippage) between the ground

and the gas pipeline. Features of the Recom

mended Practice (Standards) are as follows.

(1) The design method consists of strain de

sign. Strains during. an earthquake are

allowed to be in excess of the elastic limit

by evaluating fatigue damage m plastic

range.

(2) The design method for bends and tees is

very important because seismic forces

concentrate in them, while smaller strains

in a straight pipelines are due to the slip

page between the pipe and the ground.

(3) The standard consider the seismic waves

apparently propagating along the ground

surface and the strain in ground with in

clined base rock.

Table 6.4.1 shows the flow diagram of the

earthquake resistant design based on the

above concept.

Fig. 6.4.1 : Seismic Zone Coefficient

6-24


Table 6.4.1 Flow Diagram. of Earthquake-Resistant Design

)£ DMtelSIDlC a ions or esrgn(2) Natural Period of Surface Layer (3) Apparent Wavelength of

(1) Horizontal Seismic Intensity at 4 'H :rr~ . HI Seismic Motion

Base Rock T=--,r;= L== V' TI'; H

& == 0.15' u i : U,H : Thickness of Surface Layer (m)

V: Apparent Propagation

U 1 : Coefficient of - Velocity of Seismic Motion

Importance V s : Shear Wave Velocity in Surface Layer

Pipeline Buried~

~(~') V (1.0, 800)

under Public Road Others '. Elastic Wave X C< Sand 0.6 f--. (m/s)

~in UrbanArea Survey Clay 0.85

VI LO 0.8 Estimate from< Sand 62N 0.021

u, : Seismic Zone Coefficient N,Value Clay 122N 0.078 T (s)

I~

(4) Displacement Amplitude of the Surface Layer (5) Strain in Ground with Uniform Surface Layer

2 JrZ 2w' U.U =-T'Sv 'K cos- E 0 1 .=

Lh JrZ oJ< 2H3 • S•• K•• w'

(0.6, 150) &al = • cos-.-

Sv fr' V 2H

(cm/s)

/,:.1' 25)

,j.(6) Strain in Ground with Inclined Base Rock

T (s) cG2 =~CG/ +C G/

51 : Velocity Response Spectrum per Unit SeismicK wz

c G3 =X'~' tan(J • cos-Intensity at Base Rock (cm/s) V, 2H

z : Depth of Pipeline (m) Z : T<0.3s Z==405'T

Z : T~0.3s X == 122

. (J : Inclination of Base Rock (deg.)

( (7) Design of Straight Pipe )

(7)' Strain Transfer Coefficient[ (8) Design of Bend and Tee )

Ial

; q

l+(~rs : a Q

(8)' Displacement Transfer CoefficientAt·r.

q : Coefficient Considering Slippage between a* == q* • a o

Pipe and Ground q' : Coefficient Considering Slippage between

J. = ~ K, K 1Ground Spring

Pipe and Ground

, E'A Constant in Axial Relative Displacement of Pipe and Ground

Direction Do == (1- a*) • Un1 J.

(7)" L Strain in Straight Pipe -

Uniform Ground c"1 =a • cGJ (8)" Strain in Bend and Tee

Inclined Base Rock Cd =a . cG2 Bend 5 B=PB• Do

2. Strain in Joint (welded) of Straight PipeTee 5 r == Pr' • Do

Uniform Ground 5"2 =i, . a' 5 Gl

Inclined Base Rock c V4 -t. . fJ : Coefficient of Conventiona' 5 G2

i; :Stress Index

II( J(9)Allowable Pipe Strain 1

(9)' Allowable Strain in Straight Pipe (9)" Allowable Strain in Joint of Straight Pipe,(i) 1.0% or Bend and Tee

(ii) 35t/Dm ("10) (Buckling Strain Obtained by Actual 1.0%Measurement with Safety Factor of 1.25 taken intoConsideration), Whichever is Smaller

6-25


In Equation CD, 0.15 is the basic input at

base rock. Here, VI is a coefficient of impor

tance and v2 is the seismic zone coefficient

show in Figure 6.4.1.

(b) Natural Period of Surface Layer

Equation (2) gives the natural period of the

surface layer.

(2) Seismic Motion for Design

(a) Horizontal Seismic Intensity at Base Rock

The horizontal seismic intensity to be con

sidered for design is given by Equation CD

. .....@

H in Equation ® represents the thickness

of the surface layer. "Vs- shows the shear

wave velocity in the surface layer. Determi

nation of the base rock face depends on an N

value not less than 50 or a measured shear

wave velocity of 300m/sec or more.

(c) Apparent Wavelength of Seismic Motion

Apparent wavelength of seismic motion is

given by

L= V' T

V in Equation @ is the apparent propaga

tion velocity of seismic motion. Figure 6.4.2

shows the relationship between the natural

period and the apparent propagation velocity.

······CD

......@.

(0.25, 100)

(1.0,800)

5.00.1 1.0

Natural Period (8)

Fig. 6.4.2: Apparent Propagation Velocity of Seismic Motion

5.0

(0.6, 150)

(0.1, 25)

50

1Q '--_---J_--'-_---'-_.LJ....L..1-.l....L__-'-_...l.-_-'--~

0.1

100

0.5 1.0

Natural Period (8)

Fig. 6.4.3 : Velocity Response Spectrum per Unit Seismic Intensity

6-26


where

(b) Strain in a Pipe Welded Joint

The strain in a pipe welded joint is given by

Equation @.

"I • .... _ n _ -:;- . f _ "\'\ n2 _ "'121 -t- L. - .I'( - A. -t- v~ -.t..) - n - 1'\ - A

i \' : stress index (i v = 2.0)

(4) Design of Bend and Tee

(a) Strain in a Bend

The strains in bends (eB) are given by

Equation ®.cB = flB • .1 ®

where

fJ B : Coefficient of convention of bend

iB : Stress index for bending load on bend

n : Flexibility factor of bend

A : Sectional area of pipe

R : Radius of curvature of bend

I : Moment of inertia

D : Outside diameter of pipe

L : Apparent wavelength of seismic motions

~'4~. V4E0

,(_2+.::+ ;r'] ) . b}\.R.1 2 2'n'A'R2

2

where

.1 : relative displacement between the

pipe and the 'ground

fJB : coefficient of convention for bend

2 'iB 'A .12

'D ·1(5+R' 2) 'bl l+ 4 ' 2 3 <I : ~. (1+b2)-b\!PB=. 3

10 'A+5'L']' l . (l+b:z)+1O·,A·~

bl = }(l+R 'l)' {2+;r'n'R' l+(4-;r)'n ·R2• A.

1-2' n . R2 • 12'_ (4 _ it') . n •R3

• 23

b2

= (1+ R . 2){2 +;r' n •R • l +(4 -;r) . n . R 2 .,A.'}

~=n'R3.23.{'::+ ;r'] 2+(1 ] 2)''12 2'n'A'R n'A'R

K 2 : Ground spring constant in the trans

verse direction to the axis per unit pipe

length

E : Young's modulus of pipe

The relative displacement between the pipe

and the ground is given by Equation @,

.1 = (1- a*) . U;

......(J)

......@

1lZ }, ••••••@COS-

2H

KC G3 =k : ~ tan e.

r.

CPl =a' cGI

(1) Strain in Ground with Inclined Base Rock

The strain in the ground with inclined base

rock is given by Equation @.2 2

cG2 = cGl + cG3

where

S; : velocity response spectrum per unit

seismic intensity

z: depth of pipeline

S, is given by Figure 6.4.3.

(e) Strain in Ground with Uniform Surface

Layer

The strain in the ground with a uniform

surface layer is given by Equation @.

cGl =21r' Uh

/ L @

(d) Displacement Amplitude of Surface Layer

Equation @) gives the displacement ampli

tude of the surface layer.

2 JCUh

=-T . S; • Koh

• cOS-· .•....@1C 2H

where

E G3 : strain in ground occurring by differ

ence in displacements of two points

e : inclination of base rock

k : coefficient related to the natural pe

riod of ground surface

(3) Design for a Straight Pipe

(a) Strain in a Straight Pipe

The strain in a straight pipe is given by

Equation (J).

6-27


......@ ,

......@

where

a * : displacement transfer coefficient

a* = q * . a o

q" =Sine: .1*) -2; l*co{2; .1*) + ~2 {l-(4~*r} ;J(Adopt q* = 1, when slipping judging value

SJ<l)

(b) Strain in a Tee

The strains in tees (cT) are given by Equa-

tion @ and@.

cn = f3T! • L1 2

cn = f32 • L1 j

branch

f3- ,4' I/ ° D1T • A2 ° (C-l)

T1 - IT A 34 ° 2 + L • II ° Al ° C

c = 1 + 4 (Y1 I Y2 ) 3 ( D: / D 1 )

1 + 2(A.! 1.A.2)3(D2

/ D 1 )

where

iT : Stress index

fJ' . Y22oD2ToAI (1 2 )

T2 = IT ° , '2.-Al+2oLo12°A.21:T

Note: Subscripts with sectional area A, sec

ond moment of area I, outside diameter D, and

...1. are:

Subscription 1

Straight pipe in branch part

(c) Stress Index and Flexibility Factor

Stress index and flexibility factor of bends

and tees are shown in the below table.

~j : relative displacement between the main

pipe and the ground

L1 2 : relative displacement between the

branch pipe and the ground

f3n : coefficient of convention for branch

when seismic waves input in parallel to

main pipe

f3 T2 : coefficient of convention for main pipe

when seismic waves input in parallel to

Subscription 2

Subscription IT

Subscription 2T

Straight pipe in main part

Tee in branch part

Tee in main part

Type Stress Index Flexibility Illustration

Factor Unit: em

The Larger one of 1.65

n~Bend IB1.95

or 1.5. e;2R

)(Butt weld elbow) ('/)'" . :-1> -

-The larger one of

+~"( )2/3Tee Ir0.67'7 or 2.0

(Butt weld tee)

where

t: Wall thickness

R: Radius of curvature

r: Mean radius of pipe

D: Outside diameter of pipe

6-28


......@

(5) Allowable Strain

Seismic strains in straight pipeline are uni

form tensile or compressive strains in the en

tire area. The allowable strain in a straight

pipeline is smaller value of 1% or the allowable

strain due to buckling given by Equation @.

4 tE=- • --n

3 Dm

where

E : buckling strain

n: 0.11

t: pipe wall thickness (em)

Dm: mean diameter of pipe (=D-t) (em)

The allowable buckling strain is given by 35

(tlDm) (%) using Equation @ with a safety

factor of 1.25.

Earthquake Countermeasures for Gas

Distribution Systems - the Status Quo

6.4.2. Improvement of Earthquake

Resistance of Pipelines

Improving the earthquake resistance of pipe

lines is essential to : (1) prevent disaster

caused by gas leakage; (2) minimize the sus

pension of supply of gas; and (3) minimize the

restoration works thus enabling fast restora

tion of supply of gas to the customers.

The Recommended Practices for Earth

quake-Resistant Design of Gas Pipelines de

scribed in the preceding Chapters are aimed at

the improvement of the earthquake resistance

of newly constructed pipelines.

Retrofitting techniques have been developed

and are being applied to the old pipelines.

There are several kinds of retrofitting tech

niques which are recommended by the Japan

Gas Association and officially approved by the

government authorities as safe and reliable

techniques for use in the gas industries. Lin

ing the pipelines from inside with polymer

tubes is typical of these techniques.

6.4.3. Block System of Pipeline Networks

Damage susceptibility of pipelines depends

on : (1) the distance from the origin of earth

quake (the shorter the distance is, the more

intense the earthquake ground motion is, in

general); and (2) the ground conditions

(damage is apt to be concentrated to the areas

with very soft grounds, in general). Therefore,

the degree of concentration of damage varies

greatly from an area to another.

To isolate heavily damaged areas from less

damaged areas, the block system of pipeline

networks are in effect in major gas industries

in Japan. This system is aimed at minimizing

the number of suspended customers, as a re

sult, maximizing the efficiency of reetoration

activities.

The block system takes a hierarchical struc

ture; large blocks cover wide areas and the

blocking valves are remotely operated at the

control center; these blocks are divided into

medium size networks which are not connected

mutually; the medium sized blocks are

equipped with block-valves by which the blocks

can be divided further into small blocks (valves

are rperated manually).

6-29

* THE 1995 HYOGOKEN-NANBU EARTHQUAKE

* Ei'&W~B!E~!J-;q 1994~J-A!Jy~±lliEiEi'&~~,*1!r

EARTHQUAKE RESISTANT DESIGN FOR CIVIL ENGINEERINGSTRUCTURES IN JAPAN 1984EARTHQUAKE RESISTANT DESIGN FOR CIVIL ENGINEERINGSTRUCTURES IN JAPAN 1988* EARTHQUAKE RESISTANT DESIGN FOR CIVIL ENGINEERING

• STRUCTURES IN JAPAN 1992* EARTHQUAKE RESISTANT DESIGN CODES IN JAPAN• January,2000

B?jfIJ61~10~ B5: 933

ijZJJX:8~6~ A4: 306

ijZJJX:5~5~ B5: 254

ijZJJX:6~12~ B5:407

ijZJJX:9~2~ B5: 499

BBfIJ59~7~ B5:265

BBf1J63~7~ B5: 259

ijZJJX:4~10~ B5:259

ijZJJX:12~1~ A4: 171

25,000

971

4,854

45,714

6,796

5,825

9,500

6,000

7,767

2,700

EARTHQUAKE RESISTANT DESIGN CODES IN JAPANJanuary, 2000

Published by

Earthquake Engineering Committee

Japan Society of Civil Engineers

Yotsuya Lrchome Shinjukuku Tokyo, 160-0004 Japan

FAX +81-3-5379-2769 E-mail jsce-pubescivil.or.jp

Distributors

Maruzen Co.,Ltd. International Division

P.O.Box 5050 Tokyo International 100-3191 Japan

TEL +81-3-3273-3234 FAX +81-3-3278-9256

Copyright © JSCE 2000 EARTHQUAKE RESISTANT DESIGN

ISBN4-8106-0266-4

Printed in Japan, Waco Co.,Ltd.

earthquake resistant design codes in japan 2000

Documents

part icommon

reduced elastic

design earthquake

earthquake

ground displacement

ductility

ductility

japan road