United States Department of Agriculture

Forest Service

Engineering Staff

Washington, DC

Slope Stability Reference Guide for National Forests in the United States

Volume I11 August 1994

While reasonable efforts have been made to assure the accuracy of this publication, in no event will the authors, the editors, or the USDA Forest Service be liable for direct, indirect, incidental, or consequential damages resulting from any defect in, or the use or misuse of, this publication.

Cover Photo

CAMP 5 SLIDE, Willamette National Forest, Region 6, Eugene, OR

This photo shows the toe of a 250,000 cubic yard landslide that was initiated by a 15-foot horizontal road realignment. The mechanics of failure were largely controlled by excess pore-water pressures at the base of a clay soil having only residual shear strength. The slide was stabilized by installing 7,000 linear feet of horizontal drains which lowered the piezometric surface 14 feet. This photo point is located at milepost 0.1 on Forest Service Road 1926, approximately 13 miles northeast of Oakridge, Oregon.

Photo by Michael Long, Willamette National Forest, Eugene, OR.

The United States Department of Agriculture (USDA) prohibits discrimination in its programs on the basis of race, color, national origin, sex, religion, age, disability, political beliefs and marital or familial status. (Not all prohibited bases apply to all programs.) Persons with disabilities who require alternative means for communication of program infona- tion (Braille, large print, audiotape, etc.) should contact the USDA Office of Communications at 202-720-5881 (voice) or 202-720-7808 (TDD).

To file a complaint, write the Secretaly of Agriiulture, US. Depattment of Agriculture. Washington, D.C. 20250 or call 202-720-7327 (voice) or 202-720-1127 (TDD). USDA is an equal employment opporlunity employer.

United States Department of Agriculture

Slope Stability Reference Guide for National Forests

Forest Sewice

Engineering Staff in the United States

Washington, DC

EM-71 70-1 3 Volume I11 August 1994

Coordinators:

Rodney W. Prellwitz Thomas E. Koler John E. Steward

Editors:

David E. Hall Michael T. Long Michael D. Rernboldt

SECTION 6

EVALUATION OF STABILIZATION ALTERNATIVES

Principal contributors:

Mike Burke, Geotechnical Engineer (Section Leader)

USDA Forest Service San Juan National Forest 701 Camino Del Rio Durango, CO 81301

Cliff Denning, Geotechnical Engineer USDA Forest Service Mt. Hood National Forest 2955 NW Division Street Gresham, OR 97030

Ken Inouye, Geotechnical Engineer USDA Forest Service Regional Office Engineering 2245 Morello Avenue Pleasant Hill, CA 94523

Gordon Keller, Geotechnical Engineer USDA Forest Semce Plumas National Forest 159 Lawrence Street Quincy, CA 95971

Mike Long, Engineering Geologist USDA Forest Service Willamette National Forest P.O. Box 10607 Eugene, OR 97440

Jim McKean, Engineering Geologist USDA Forest Service Regional Office Engineering 2245 Morello Avenue Pleasant Hill, CA 94523

Mike Rembolt, Assistant Forest Engineer USDA Forest Service Payette National Forest Supervisor's Office P.O. Box 1026 McCall, ID 83638

R e d Renteria, Geotechnical Engineer USDA Forest Semce Regional Office Engineering 324 25th Street Ogden, UT 84401

Ed Rose, Geotechnical Engineer USDA Forest Service Klamath National Forest 1312 Fairlane Road Yreka, CA 96097

John Steward, Geotechnical Engineer USDA Forest Service Washington Omce Engineering 201 14th Street, S.W. Washington, DC 20250

Ted Stuart, Regional Geotechnical Engineer USDA Forest Service Regional Office Engineering 2245 Morello Avenue Pleasant Hill, CA 94523

Richard VanDyke, Geotechnical Engineer USDA Forest Service Siskiyou National Forest Westside Engineering Zone 93976 Ocean Way Gold Beach. OR 97444

Rod Prellwitz, Geotechnical Engineer USDA Forest Service Intermountain Research Station 1221 S. Main Moscow, ID 83843

Page

. . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 6A Stabilization Considerations 733 . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 6A.I Introduction 733

6A.2 Prestabilization Analysis Considerations . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 733 6A.3 Selection of Feasible Stabilization Techniques . . . . . . . . . . . . . . . . . . . . . . . . . . 734 6A.4 Analysis of Stabilization Alternatives . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 734

. . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 6A.5 When Does "Safe" Become "Unsafe"? 736 . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 6A.5.1 Litigation 736

6A.5.2 Review of Designs by Others . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 736 . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 6B Modification of Geometry 737 . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 6B.1 General Information 737

6B.2 Necessary Redesign Information . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 737 . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 6B.3 The Stability of Design 738

6B.3.1 Modification of Road Cross-Section . . . . . . . . . . . . . . . . . . . . . . . . . . . 738 . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 6B.3.1.1 Flattened Slopes 738

6B.3.1.2 Steepened Slopes . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 739 . . . . . . . . . . . . . . . . . . . . . . . . . . 6B.3.1.3 Lightweight Embankments 740

. . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 6B.3.2 Shift in Horizontal Alignment 741 6B.3.3 Modification of Vertical Alignment . . . . . . . . . . . . . . . . . . . . . . . . . . . 741

. . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 6B.3.3.1 Sag Vertical Curve 742 . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 6B.3.3.2 Crest Vertical Curve 742

6B.4 Construction and Preconstruction Considerations . . . . . . . . . . . . . . . . . . . . . . . . 743 6B.5 Relative Costs . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 744 6B.6 Sample Problems . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 744

. . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 6C Surface and Subsurface Drainage 767 . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 6C.1 General Information 767

. . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 6C.2 Surface Drainage 767 . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 6C.3 Subsurface Drainage 768

. . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 6C.3.1 Cutoff Trenches 768 . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 6C.3.2 Horizontal Drains 781

. . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 6D Horizontal Drains 783 . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 6D.1 General Information 783

6D.2 Predesign Information-Investigation Techniques . . . . . . . . . . . . . . . . . . . . . . . . 784 6D.2.1 Area Reconnaissance . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 784 6D.2.2 Ground Control Survey . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 784 6D.2.3 Subsurface Interpretation . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 785 6D.2.4 Drive Probe Exploration . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 785 6D.2.5 Electrical Resistivity Profiling . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 785 6D.2.6 Drilling Exploration . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 785

. . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 6D.2.7 Permeability Testing 786 . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 6D.2.8 Ground Water Tracing 786

6D.2.9 Water Surface Contours . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 786 6D.2.10 Test Drain Installation . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 787

6D.3 Drainage System Design . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 788 6D.3.1 Ground Water Recharge Capacity . . . . . . . . . . . . . . . . . . . . . . . . . . . . 788 6D.3.2 Number of Drains Needed . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 789 6D.3.3 Slot Width and Spacing . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 790

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6D.3.4 Effective End Spacing and Drawdown . . . . . . . . . . . . . . . . . . . . . . . . . . 790 6D.3.5 Collector System . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 797

6D.4 Construction Considerations . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 797 6D.4.1 Suggested Construction Practices . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 797 6D.4.2 Inspector Duties . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 798 6D.4.3 Alternative Construction Method . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 799

6D.5 Post-Construction Monitoring . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 799 6D.6 Case History Summaries . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 800 6D.7 Design Example-< amp Five Slide . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 802

6E Buttresses . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 803 6E.1 General Information . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 803 6E.2 Rock Buttresses . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 803

6E.2.1 General . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 803 6E.2.2 Foundation Bearing Capacity . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 804 6E.2.3 External Stability . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 804 6E.2.4 Drainage . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 805 6E.2.5 Construction . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 805

6E.3 Earth Buttresses . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 805 6E.3.1 General . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 805 6E.3.2 Internal Stability . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 806 6E.3.3 Drainage . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 806 6E.3.4 Construction . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 806

6E.4 Retaining Walls . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 807 6E.4.1 General . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 807 6E.4.2 Foundation Bearing Capacity . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 807 6E.4.3 External Stability . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 807 6E.4.4 Internal Stability . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 807 6E.4.5 Drainage . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 807 6E.4.6 Construction . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 807

6E.5 Buttress Sample Problem . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 808 6E.5.1 Description of Area . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 808 6E.5.2 Background . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 808 6E.5.3 Field Survey . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 809 6E.5.4 Field Investigation . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 810 6E.5.5 Laboratory Testing . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 810 6E.5.6 Analysis Procedures . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 810 6E.5.7 Construction . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 811

6F Soil Slope Stabilization-Reinforced Fills . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 817 6F.1 General Information . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 817

6F.l.l Introduction . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 817 6F.1.2 Advantages and Disadvantages . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 817 6F . 1.3 Applications . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 819

6F.2 Necessary Predesign Information . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 819 6F.3 The Stability Design . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 820 6F.4 Unique Construction and Preconstruction Considerations . . . . . . . . . . . . . . . . . . . 821

6F.4.1 Internal Reinforcement Materials . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 821 6F.4.2 Standard and Marginal Backfill Material . . . . . . . . . . . . . . . . . . . . . . . . 822 6F.4.3 Fill Face Slope and Facing Needs . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 823

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6F.4.4 Drainage Requirements . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 824 6F.4.5 Typical Construction Sequence . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 825

6F.5 Relative Costs . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 825 6F.6 Sample Problems . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 826

6F.6.1 Willow Slide Reinforced Fill . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 826 6F.6.2 Elk River Buttress vs . Reinforced Fill . . . . . . . . . . . . . . . . . . . . . . . . . 837

6G Shear Trenches . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 841 6G.1 General . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 841 6G.2 Preconstruction . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 841 6G.3 Shear Trench Width . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 843 6G.4 Shear Trench Depth . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 845 6G.5 Analysis with Passive Wedge Exiting Through Shear Trench . . . . . . . . . . . . . . . . 846 6G.6 Analysis with Passive Wedge Exiting Behind Shear Trench . . . . . . . . . . . . . . . . . 847 6G.7 Analysis with Active Wedge Exiting Shear Trench . . . . . . . . . . . . . . . . . . . . . . . 847 6G.8 Counterbalance Fill . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 848 6G.9 Construction . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 849 6G.10 Post-Construction . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 850 6G.11 Sample Problem . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 850

6H Rock Slope Stabilization . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 851 6H.1 General Introduction . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 851 6H.2 Advantages and Disadvantages . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 851 6H.3 Necessary Design Information . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 851 6H.4 Stabilization Components and Systems . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 853

6H.4.1 Reduce Draining Forces . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 853 6H.4.2 Increase Resisting Forces . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 853 6H.4.3 Rockfall Mitigation . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 857

. . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 6H.5 Example Problems 859 61 Comparison of Alternatives and Decision Analysis . . . . . . . . . . . . . . . . . . . . . . . . . . . . 861

61.1 Introduction . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 861 61.2 Methods for Equivalent Comparison of Alternatives . . . . . . . . . . . . . . . . . . . . . . 861

61.2.1 Defining Mutually Exclusive Alternatives . . . . . . . . . . . . . . . . . . . . . . . 861 61.2.2 Setting Project Life for Comparison . . . . . . . . . . . . . . . . . . . . . . . . . . . 862 61.2.3 Determining Equivalence Using the Time Value of Money . . . . . . . . . . . 863

61.2.3.1 Present Worth . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 866 61.2.3.2 Annual Worth . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 866 61.2.3.3 Future Worth . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 866

61.3 Comparison Under Conditions of Uncertainty and Risk . . . . . . . . . . . . . . . . . . . . 868 61.3.1 Break-Even Analysis . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 868 61.3.2 Sensitivity Analysis . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 869 61.3.3 Hazard Assessment and Risk Analysis . . . . . . . . . . . . . . . . . . . . . . . . . 872

61.3.3.1 Types of Uncertainty . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 873 61.3.3.2 Quantifying Uncertainty . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 874 61.3.3.3 Simulation Techniques . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 874

61.4 Decision Analysis-How to Select the Preferred Alternative . . . . . . . . . . . . . . . . 881 61.4.1 Why Do a Decision Analysis? . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 881 61.4.2 Decisions Using Assumed Certainty . . . . . . . . . . . . . . . . . . . . . . . . . . . 882 61.4.3 Decisions Considering Risk . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 882

61.4.3.1 Dominance . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 882

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61.4.3.2 Expected Value . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 882 61.4.3.3 Most Probable Future Principle . . . . . . . . . . . . . . . . . . . . . . . . 882

61.4.4 Decisions Using Uncertain Outcomes . . . . . . . . . . . . . . . . . . . . . . . . . . 883 61.4.4.1 The LaPlace Principle . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 883 61.4.4.2 The Optimist Principle . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 884

. . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 61.4.4.3 The Pessimist Principle 884 61.4.5 The Use of Decision Trees for EMV Analysis . . . . . . . . . . . . . . . . . . . . 885

61.4.5.1 What Is EMV? . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 885 61.4.5.2 How to Use a Decision Tree . . . . . . . . . . . . . . . . . . . . . . . . . . 886 61.4.5.3 EMV Analysis Using Imperfect Information . . . . . . . . . . . . . . . 888

6L4.6 Decisions Using Preference Theory . . . . . . . . . . . . . . . . . . . . . . . . . . . . 894 65 Construction Control . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 897

. . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 6J.1 Acknowledgement 897 . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 6J.2 General Information 897

65.3 Process . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 898 6J.4 Organization . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 898

65.4.1 Forest Service Construction Staff . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 898 65.4.2 Geotechnical Specialist . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 898

. . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 65.5 Plans and Specifications 898 6J.6 Inspection . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 899

. . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 65.7 Contract Elements 899 . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 6J.7.1 Erosion Control 899

. . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 65.7.2 Earthwork 900 . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 65.7.3 Clearing and Grubbing 900

. . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 6J.7.4 Slope Ditches 900 6J.7.5 Excavation . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 901 61.7.6 Rock Excavation . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 901 6J.7.7 Embankment Foundations . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 901 65.7.8 Compaction . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 901

6J.7.8.1 Compaction-Moisture Control . . . . . . . . . . . . . . . . . . . . . . . . 902 6J.7.8.2 Compaction-Weaving and Pumping . . . . . . . . . . . . . . . . . . . . 902 65.7.8.3 Rutting . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 903 65.7.8.4 Compaction Equipment . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 903 6J.7.8.5 Compaction in Confined Areas . . . . . . . . . . . . . . . . . . . . . . . . 903 6J.7.8.6 Compaction Inspection . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 904

6J.8 Construction Control-Mechanically Stabilized Embankments (MSE's) and Walls . 904 65.8.1 Inspection Elements . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 905 6J.8.2 Plans and Specifications . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 905 6J.8.3 Review of Site Conditions and Foundation Requirements . . . . . . . . . . . . . 906 65.8.4 Material Requirements . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 906

61.8.4.1 Reinforcing Elements . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 906 65.8.4.2 Precast Elements . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 907 6J.8.4.3 Facing Joint Materials . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 907 6J.8.4.4 Placement of Reinforcing Material . . . . . . . . . . . . . . . . . . . . . . 907

6J.8.5 Backfill . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 907 65.8.5.1 Placement . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 907 6J.8.5.2 Compaction Equipment . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 908

6J.9 Construction Control-Underdrains . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 908

Table of Contents

Page

6J.10 Excavations and Trenches-Safety . Stability. and Design . . . . . . . . . . . . . . . . . . 909 . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 6J.10.1 General 909

65.10.2 Collapse Prevention . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 909 . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 65.10.2.1 Sloped Sides 910

. . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 65.10.2.2 Tabulated Criteria 910 . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 61.10.2.3 Trench Box 910

. . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 6J.10.3 Analysis and Design 910 6K Post-Construction Monitoring of the Technical Structures and Projects . . . . . . . . . . . . . . 913

. . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 6K.1 Acknowledgement 913 . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 6K.2 Introduction 913

. . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 6K.3 Limited Monitoring Program 913 . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 6K.4 Comprehensive Monitoring Program 913

. . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 6K.5 Planning Monitoring Programs 914 6K.5.1 Purpose of the Monitoring Program . . . . . . . . . . . . . . . . . . . . . . . . . . . 914 6K.5.2 Define the Project Conditions . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 914 6K.5.3 Predict Mechanisms That Control Behavior . . . . . . . . . . . . . . . . . . . . . . 914 6K.5.4 Select the Parameters to be Monitored . . . . . . . . . . . . . . . . . . . . . . . . . 914 6K.5.5 Predict Magnitudes of Change . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 915

. . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 6K.5.6 Devise Remedial Action 915 6K.5.7 Assign Monitoring Tasks for Design, Construction, and Operation Phases 915

. . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 6K.5.8 Select Instruments 915 . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 6K.5.9 Select Instrument Locations 918

6K.5.10 Plan Recording of Factors That May Influence Measured Data . . . . . . . . 920 6K.5.11 Establish Procedures for Ensuring Reading Correctness . . . . . . . . . . . . . 921

. . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 6K.5.12 Prepare Budget 921 6K.5.13 Write Instrument Procurement Specifications . . . . . . . . . . . . . . . . . . . . 921

. . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 6K.5.14 Plan Installation 921 6K.5.15 Plan Regular Calibration and Maintenance . . . . . . . . . . . . . . . . . . . . . . 922 6K.5.16 Plan Data Collection, Processing, Presentation, Interpretation.

. . . . . . . . . . . . . . . . . . . . . . . . . . . . . . Reporting, and Implementation 923 6K.5.17 Write Contractual Arrangements for Field Instrumentation Services . . . . . 924

. . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 6K.5.18 Update Budget 924 . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 6K.6 Executing Monitoring Program 924

6K.7 Discussion on the Use and Misuse of Instrumentation . . . . . . . . . . . . . . . . . . . . . 924 References . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 927

Appendices 6.1 Horizontal Drains Design Example-Camp Five Slide . . . . . . . . . . . . . . . . . . . . . . . . . . 935

. . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 6.2 Shear Trench Sample Problem 991 6.3 Rock Slope StabilizationSystem Specifications Examples . . . . . . . . . . . . . . . . . . . . . 1015 6.4 Powder Creek SSI Demonstration Problem . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 1039 6.5 Probability and Statistics Refresher . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 1047 6.6 Field and Laboratory Evaluation of Geocomposite Drain Systems

for Use on Low-Volume Roads . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 1053 . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 6.7 Soil Nailing 1065

6A. Stabilization Considerations

Rodney W. Prellwitz, Geotechnical Engineer, Intermountain Research Station

6A.1 Introduction

6A.2 Prestabilization Analysis Considerations

Section 6 discusses methods of stabilizing a slope against mass failure. The focus is on techniques to prevent or correct mass wasting (landsliding), which can be analyzed by methods discussed in section 5, and not surface erosion. Although biostabilization with vegetation can be thought of as preventing shallow mass erosion, here it is considered to be a technique for the prevention of surface erosion. Section 6F contains a discussion of biostabilization in conjunction with reinforced fill slopes to aid in stabilization of the over-steepened slope, but the reader is referred to the cited references for more complete coverage of this technique. The purpose of this section is to point out the basic steps of stabilization analysis and to discuss several stabilization techniques commonly applied to slope failures on Forest Service roads. Figure 6A.1 illustrates where the topics covered in section 6 fall within the three-level slope stability analysis process.

I LEV,ELl (c( I I A N D Y D E I ANALYSIS DATA BASE INVENTORY

RESOURCE 1 9 \

LEVEL ll LEVEL ll I ANALYSIS JC( DATABASE I*\ \ -I- -

PROJECT

Figure 6A.l.- Section 6 is concerned with level I11 analysis within the three-level system

Fundamental to the solution of any problem is the definition of what that problem really is. In this case, the selection of the appropriate slope stabilization technique must be preceded by an adequate determination of the site conditions likely to cause instability. The techniques covered in section 3 for preparing a detailed field- developed cross-section on which to develop a hypothesis of subsurface conditions,

and subsequent subsurface investigation to verify those conditions, are essential prerequisites of the stabilization analysis. Following the site investigation, a thorough stability analysis of the unstabilized slope is necessary to identify the extent of the stability problem and to provide a standard of comparison for all possible stabilization alternatives. All levels of stability analysis discussed in section 5 and demonstrated in the sample problem in section 5G may be useful in this unstabilized slope analysis for back-calculation analysis and parametric value evaluation.

6A.3 Selection An evaluation of the results of the unstabilized slope analysis and the existing site

of Feasible conditions will usually indicate the most practical stabilization efforts. This may

Stabilization appear to be fundamental logic, but this initial scoping of all possible stabilization techniques can save analysis and design time and perhaps prevent oversight of the

Techniques obvious; as Wooten's Third Law states, "The acquisition of uncommon knowledge inhibits the application of common sense" (Wooten, 1971). Some examples: if ground water appears to be the dominant cause of instability, then a drainage system should be the most effective stabilization measure; if the unstable slope is already excessively steep, then a reinforced fill or retaining wall should function better than a rock buttress; if a minor realignment of the road location can avoid the problem completely, then maybe that is the best alternative. The point is that all possible stabilization techniques should be identified, based on the unstabilized slope stability analysis, before selecting alternatives to evaluate.

6A.4 Analysis Once the stabilization alternatives applicable to the site conditions are identified, a

of Stabilization stability analysis of each is necessary to determine the extent, dimensions, etc., of

Alternatives that technique required to bring the unstabilized slope to an acceptable level of stability. What is an acceptable level of stability? A probabilistic method in a risk analysis format-the preferred method+onsiders the probability of the stabilization measure failing, the cost of construction of the measure, and the consequences of failure, objectively comparing possible alternatives. Section 61 introduces risk analysis for the selection of stabilization alternatives. However, the current state of the art is based on deterministic slope stability analysis. This methodology develops alternatives which have reasonably secure safety factors against failure and then compares them subjectively on the basis of construction cost, potential for failure, expected life span, consequences of failure, maintenance cost, and aesthetics. In this deterministic context, selection of the design factor of safety determines the acceptable level of stability. Vandre (1980) gives the following perspective on the design factor of safety (FOS):

The design factor of safety allows for a margin of error between the parameters and assumptions used in design and those that actually exist in the field. The magnitude of the desired safety factor considers:

(1) The consequences of instability [environmental impacts, property damage, replacement cost, closed roads].

(2) The adequacy of investigations [confidence in the failure model, definition of the site geology and soillrock definition].

The reliability of the design assumptions [natural variation in soil parameters, temporal distribution of ground water, type of ground water (unconfined or confined)].

The ability to predict adverse conditions [ground water extremes, storm events, earthquakes, etc. "Design on the basis of the most unfavorable assumptions is inevitably uneconomical, but no other procedure provides the designer in advance of construction with the assurance that the soil- supported structure will not develop unanticipated defects" (Terzaghi and Peck, 1948)l.

The possibility of construction deviations from the design [inadequate compaction, inadequate trench depth, unsuitable embankment material, incorrect location].

Judgment based on past experience [acceptable choices based on historical success].

There are techniques available to the designer that reduce uncertainty and increase confidence; these include better definition of the site conditions and failure model through site investigation, ground water monitoring, laboratory testing, and back- calculation analysis. These techniques tend to reduce large errors in Vandre's components (2). (3), and (4). This leaves components (I), (5). and (6) with some degree of uncertainty, which is inevitable in the selection of the design FOS. Even with the most intensive investigation and lab testing program, and using the most sophisticated methods of back-calculation analysis, to define the conditions where the FOS equals 1.00 for the unstabilized failure model, most experienced engineers would not be comfortable with a design FOS = 1.00 for the stabilization alternative. This discomfort is the result of lingering uncertainty in components (3), (4), and (5); the concern for the consequences in component (1); and, often, the lack of historical precedence (experience) in component (6).

Ideally, a universally applicable and accepted design FOS-such as 1.25, 1.50, or 3 . ~ 0 u l d be used as a minimum safe level to account for all of these uncertainties and variations. In reality, no design FOS can substitute for an adequate site investigation, parameter evaluation, or problem definition program. It can only provide for an additional margin of error. Sections 5A.12 through 5A.14 contain a discussion of the manner in which stability in terms of FOS can be perceived differently in deterministic and probabilistic analyses. As applied to this section, the experienced deterministic analyst will compensate for this uncertainty and natural variability in the evaluation of the unstabilized slope to arrive at the standard of comparison for the stabilization measures. This must be factored into the selection of the design FOS to avoid the danger of compounding conservatism, The experienced practitioner will not use a fured design FOS (e.g., 1.25 or 1.50). largely because of the additional cost which might be necessary to reach that number but which might not be justified in the overall consideration of the components stated above. In the sample problems of sections 6C through 6H you will note the range of design FOS which these experienced practitioners have determined to be "safe" for their particular problems. Even for the same site conditions, different design FOS's may be deemed necessary for different stabilization alternatives as a result of how components (3). ( 3 , or (6) relate to that specific alternative.

6A.5 When Does "Safe" Become "Unsafe"?

6A.5.1 Litigation

6A.5.2 Review of Designs by Others

There are many answers to this question when all of the components are considered in selecting a design FOS as discussed in section 6A.4. The following discussion concerns what the experienced practitioner perceives to be a "safe" design FOS for a given situation when he or she is completing the design, compared with what the public might perceive. This is primarily the result of two situations:

(1) During litigation in a liability suit for slope failure.

(2) In review of designs by others for slopes to be constructed on national forest lands.

Each of these cases is unique and will be discussed and illustrated separately,

A good example of this is a liability suit in which the agency is sued for death or injuries resulting from a landslide or slope failure on a Forest Service road. The design engineer designs a fill slope to be "stable" for traffic in the traveled way but accidents still occur as a result of travel on the extreme shoulder (outside 2 to 3 feet). The design and construction engineers are aware that the extreme edge of a fill slope cannot be adequately compacted. They allow "fill widening" to account for this with the assumption that the prudent driver will travel only within the traveled way. This assumption has cost the agency several million dollars in more than one lawsuit. In the litigation arena, when a vehicle's wheel sinks into the soft shoulder and the vehicle goes off of the fill slope as a result, it is a "slope failure." The obvious preventive measure is not to increase the design FOS for the traveled way but to restrict or prevent travel on the shoulder.

This is a different arena for dealing with the public. The agency design engineer is charged with the responsibility to review the design prepared by an outside consulting firm for a slope to be constructed on national forest lands. This may force the agency design engineer to establish a minimum design FOS for the consulting engineer. In establishing guidelines for the design of mine tailings dams to be constructed on national forest lands in the Intermountain Region, Vandre (1992) gives the following definition:

A minimum safety factor is a threshold value for safety assurance. . . The fact that the slopes designed to satisfy this criterion generally have been stable, provides the assurance that future slopes that satisfy the same minimum will also be stable . . . The safety factors in our review handbook are used as guidelines. A guideline is distinguished from a requirement in that it implies allowable exceptions, flexibility, and avoids technicalities (1.29 versus 1.30). To prudently justify a lower safety factor, uncertainties would need to be reduced by controlling variability by making more conservative assumptions or designing for worst-case conditions.

The establishment of a minimum design FOS helps in the review process as a communication tool between the review agency engineer and the consulting engineer. It is not foolproof, however, and may intensify the review process, but it ensures that an acceptable level of investigation and realistic parametric values are being used as the basis for the design.

6B. Modification of Geometry

Rent! Renteria, Geotechnical Engineer, Intermountain Regional Ofjice

6B.1 General Modification of geometry is a mitigation measure that closely mimics nature. The Information force of gravity is constantly reshaping the landscape through natural slope building

processes. Unfortunately, nature's forces also affect areas with human-built structures. Sometimes these human-built structures give nature a little "push."

An advantage of modifying the geometry of a slide area is that for many situations, the repair works with instead of against nature. Some examples are: flattening the slope angle, using a sag vertical curve, using a steepened embankment, and using a lightweight fill that would reduce the weight load on a slope. Most of these options are substantially less expensive than buttresses or retaining structures.

A disadvantage of modifying the geometry is the possible increased effect on land resources. Flattening a cut slope could result in a large exposed area, possibly creating sediment production through erosion. In addition to creating a "bend in the road," a shift in horizontal alignment creates additional excavated material that must be disposed of. A sag vertical curve may concentrate water runoff into a sensitive area.

The geotechnical specialist must carefully determine the advantages and disadvantages for each application. All significant benefits and impacts should be relayed to the decision maker.

68.2 Any road design project will consist of meeting the road management objectives

Necessary (RMO's) decided upon in an interdisciplinary forum. When considering modification

Predesign of geometry as a slope stability mitigation, some important design considerations are:

lnformation Desired road width

Horizontal alignment impacts

Vertical alignment impacts

Design speed

Sight distance

Traffic during construction

Type of traffic

Seasonal use of road

Materials availability

Site drainage

Economic

68.3. The Stability Design

68.3.1 As demonstrated in section 5, stability is a function of material strength, weight, and Modification of slope angleheight. Thus, mitigation can be accomplished by increasing the material Road strength, reducing the driving weight, decreasing the slope angle, reducing the slope Cross-Section height, or a combination of these methods. This section explores modification of

road geometry and how it uses these mitigation techniques.

6B.3.1.1 Flattened Slopes

Flattened slopes can be applied to either cut slopes or fill slopes. A flattened slope is the typical natural response of most slope failures. A cut slope failure is typically followed by the maintenance response of material removal from the road. However, a fill slope failure often results in lost road width and downslope resource damage from the failed material. In either case a proactive design is encouraged.

An example of flattening a fill slope is sidecast fill removal. Sidecast fill is typically "drifted over the groundslope during road const~uction. The result is a fill slope at the angle of repose and possibly a foundation of organic material. It is not uncommon to find logs at the base of these fills acting as artificial retaining support (figure 6B.1). As the logs and other organics decay, downhill movement and cracking or loss of roadway result. The movement will continue until a more stable, flatter slope is achieved. Removal of the sidecast material will return the slope to the flatter angle of the original ground. Of course, this alternative requires reducing the road width, possibly obtaining additional road width from the opposite side, or rebuilding the outside edge (figure 6B.2).

POTENTIAL WATER FLOW P A M

A-

ORGANIC LAYER /-&'

Figure 6B.I.-Typical view of a sidecastfill.

*+, ORGANIC LAYER

I

Figure 6B.2.-Typical section for "move into the hill."

Reducing the slope angle of a cut may increase the FOS as a result of increased resisting weight and the lengthening of failure surfaces. An additional effect of a reduced slope angle is a greater drawdown of the phreatic surface. A disadvantage of flattening a cut slope is the increased excavation created and associated resource disturbance. This can be a problem economically as well as for the slope stability of the generated material. If adequate disposal sites are not readily available within economical haul distances, the geotechnical specialist may need to consider and analyze less stable locations for an engineered waste embankment.

6B.3.1.2 Steepened Slopes

The use of steepened slopes as mitigation is generally applied to fill slopes rather than cut slopes. The use of a steepened slope generally requires a design based on materials chosen. Some examples of steepened slopes are: rock buttresses (see

section 6E), reinforced embankments (see section 6F). and retaining walls (see section 6E).

A typical situation for steepened fill slopes is sidecast fill removal. As was described for flattened fill slopes, sidecast material caught on logs or built on organics will settle and move downhill as the organics decay. If the road template cannot be modified, it may be economical to remove some or all of the sidecast material and rebuild a steepened slope. The greatest stability is achieved when the rebuilt section is placed on a firm foundation, generally entirely on the foundation material. This can be quite critical for a retaining structure, and the external stability of the foundation should not be overlooked (see section 6E.4). An advantage of building a steepened fill slope is that suitable embankment material can be conserved and borrow minimized. This results in less impact on the right-of-way and possible impact on a soil borrow or rock material source. A case history is presented in problem 1.

A steepened cut slope alternative is possible with soils exhibiting some cohesion and generally results in a benched cut. As was demonstrated in section 5C.3, height is one of the factors controlling stability for soils with cohesion. However, for cohesionless materials with no ground water influence, it was shown that for any slope angle, the stability is controlled by the friction angle of the material and is independent of height (section 5C.5). Therefore, for soils with cohesion the idea of a steepened slope is to limit the height. By placing a bench in the slope it is possible to achieve a desired FOS with economical excavation quantities compared to a flattened slope. The use of a benched cut for cohesionless materials will generally reduce erosion rather than increase mass slope stability.

6B.3.1.3 Lightweight Embankments

Lightweight embankments have been used for both bearing capacity and slope stability problems. Because of the expense, lightweight embankments are generally considered when alignment cannot be modified or the additional weight of other alternatives does not provide an acceptable increase in stability (see problem 2).

Lightweight embankments have been built using various materials and at various slope angles. Some of the materials used, such as wood chips and shredded tires, are waste byproducts. Other materials, such as polystyrene foam and plastics, are produced for other uses. In all cases, the geotechnical specialist must assign material properties either through laboratory testing, research of existing test data, or engineering judgment. The parameters are then used in a stability analysis, along with site constraints and economics, to determine the embankment slope angle and the possible need for internal andlor external reinforcement.

Some typical parameters used for wood material embankments are as follows:

y (moist)

40-50 pcf I y (saturated)

I 50-80 pcf

C ' 4' 0 psf 30-36'

68.3.2 Shift in An alternative that shifts horizontal alignment is typically referred to as a "move into Horizontal the hill" since it is almost always applied to a cut slope side of the road template. Alignment This alternative is typically combined with the flattened fill slope alternative. The

idea is to gain additional road width by shifting the roadway off the slide feature- either the head of a slide that results in a drop in grade or the lost outside edge of the roadway. The general result is a full-bench road template where the main stability concern is the stability of the cut slope (figure 6B.3). Because a larger base width is placed in natural ground, the usual result in forest terrain is a higher cut slope. The geotechnical designer may need to consider a flatter cut slope than the previously existing cut or possibly a benched cut. The benched cut may also be required for equipment access. A case history for a shift in alignment is presented in problem 3.

S e c t R-R' Shift Alignment (benched)

1625 , 10 most crttlcal surfaces, MINIMUM BISHOP FOS = 1.131

900 930 960 990 1020 1050 1080 1110 1140 X-AXIS (feet)

Figure 6B.3.-XSTABL cross-section for a cur slope stability analysis

6B.3.3 The modification of vertical alignment is a method that combines road geometrics Modification of with stability mitigation. The geotechnical designer should be acquainted with Vertical Alignment standard geometric design to understand the terminology and design constraints of

vertical curves. The following sections will present the use of sag and crest vertical curves as a method of slope stability mitigation.

6B.3.3.1 Sag Vertical Curve

A sag vertical curve (figure 6B.4) results when a descending grade meets an ascending grade. A sag profile is the typical result of a road that crosses an active slide. Just as a flattened slope can result in stability closer to a natural state, so can a sag vertical curve. The use of a sag vertical curve is most appropriate at the head of a slide and where road design speed and minimum curvature allow for it. One potential negative aspect of a sag vertical curve is drainage. The low point of the sag is typically within the slide mass and ditch water may drain into the slide, or a ditch relief culvert may place water onto the slide mass. The geotechnical designer is encouraged to consider drainage methods that limit ditch water entering the slide area, such as placing relief culverts before water enters the slide boundaries and using pipe extensions (such as corrugated plastic pipe) within the slide mass. The use of sag vertical curves can be a very inexpensive short- or long-term stability mitigation. A case history is presented in problem 4.

Figure 6B.4.-lllusfration of a sag vertical curve.

6B.3.3.2 Crest Vertical Curve

A crest vertical curve (figure 6B.5) results when an ascending grade meets a descending grade. A crest vertical curve may be used for a buttressing effect on a cut slope failure or as a method to reduce cut slope height. A crest vertical curve also results in a positive means to divert water drainage from the crest point. The geotechnical designer decides on the appropriate fill slope based on stability and economics previously discussed. A typical "fill slope" used with crest vertical curves is the natural ground slope for a full-bench cut, or a mechanically stabilized embankment (MSE). A case history for the use of a crest vertical curve is presented in problem 5.

Figure 6B.5.-illustration of a crest vertical curve

6B.4 Construction and Preconstruction Considerations

Some general issues to consider when analyzing the use of modification of geometry:

Will a change in geometry repair any ongoing resource damage, or will it just restore the road template? What are the future risks to the road if resource damage continues?

Will construction interfere with traffic flow?

Can the road be closed for short periods?

What type of equipment is available?

What is the reach (typically 2&30 feet upslope or 15-25 feet downslope) for an available tracked excavator?

How much removed soil material might be reusable?

Is there a suitable location within economical haul distance (generally 1&20 miles) for waste material?

Will an access ramp be needed to reach the extent of the excavation? Can the access ramp be constructed within the designed excavation limits, or will it need to be constructed in the roadway? From where would the material come? To where will it be removed?

After construction of new cut slopes or fill slopes, will they need surface stabilization for erosion control (i.e., erosion mats, straw, grass)?

For lightweight embankments, what materials are available economically? Are there possible side effects (leaching, toxicity, etc.)?

For a shift in alignment, is the shift large enough to accommodate construction equipment (dozer or excavator)? Will the new horizontal alignment allow passage of long vehicles (i.e., log trucks and yarders) such that tires track within the roadway?

For sag vertical curves, can water drainage be removed adequately (water concentration at low point, which is generally the center of the slide)? Will vehicles with trailers have adequate clearance when the vehicle is on the "upside" and the trailer is on the "downside"?

For crest vertical curves, will vehicles have adequate sight distance? Will vehicles with trailers have adequate clearance when the vehicle is on the "downside" and the trailer is on the "upside"?

6B.5 Relative Costs

Construction of geometric changes can generally be accomplished using common road construction equipment. Unless reinforcement or a retaining structure is to be employed, typical unit costs for road construction can be used for estimates. The following table lists some common unit costs that may be used to estimate alternative costs. Because these values average the cost of mobilization, unit costs should be increased 1&50 percent for small projects and reduced 5 to 20 percent for large ones. These unit costs are given for comparison of alternatives and should not be used for contract preparation (for this, use time and equipment methods). All costs are per cubic yard.

( Excavation I Blasting I Embankment I Haul I Borrow

6B.6 Sample The following five sample problems are case histories that demonstrate embankment

Problems failure mitigation, use of a lightweight fill, shift in horizontal alignment, use of a sag vertical curve, and use of a crest vertical curve.

soil: $1.00-4.00 rip: $ 3 . W . 0 0

Problem 1: Embankment Failure Mitigation

Case History: Blackhawk Campground Road, USDA Forest Service, Uinta National Forest. Original investigation by Rene' A. Renteria, Geotechnical Engineer, September 2, 1992.

$8.00-15.00

Observations

The road begins from a ridge top alignment at a junction with the Mt. Nebo Loop Road, #015. The road descends along a ridge top alignment, then crosses the ridge and continues at the base of the slope just above a large bench area. After about a mile, the road descends slightly before ascending back to a ridge line and saddle.

$3.00-7.00 (in-place)

$0.4@0.60 per mile (loose)

$1.50-5.00 (loose) + haul

At approximately milepost 1, approximately 240 feet of the outside lane of the two-lane paved road has settled. The road had been patched this summer, and already new cracking of the pavement has occurred. The cracking varies from 1-2 feet in from the edge to a width of 18 feet from the outside edge. There are no visible cracks that cross the entire asphalt pavement.

The slope below the road has benched terrain, with arid-type vegetation. Below the small bench is a large benched area with meadow-like characteristics. At the toe of the embankment there is no evidence of water flow exiting at the embankment1 natural ground contact. The fill slope appears to have a convex shape (bowed outward). The fill slope is well vegetated with grass.

From observations of the cut slope, two soil units were identified and field classified. The upper layer, labeled soil unit A (SU-A), is a silty sand with rock fragments (field classified by the USCS as SM). The probable origin is glacial. At the east end of the settlement area, a second soil layer is exposed at ditch elevation. This soil, labeled soil unit B (SU-B), is considered a sandy silt (field classified by the USCS as ML); probable origin is sedimentary (decomposed rock?). This soil unit is visible in the cut slope just above ditch elevation from the east to west ends of the shoulder settlement area. During this visit, no water was visible either in the ditch or on the surrounding slopes. Discussion with Carol Johnson confirms evidence that water does flow into the ditch in late springtearly summer. Toward the west end of the settlement area, an old slope failure was noted above the road. There was some indication that the cut slope had been displaced toward the road, but it does not seem to be a recent event (older than 10 years). Minor cracks were noted just above ditch elevation on this displaced mass, but it is unclear whether these are cracks from creep movement of the cut slope or just shrinkage cracks after the ditch moisture dried up. The probability that the old cut slope movement and the road movement are related is extremely low. This is based on the observation that no cracks cross the width of the pavement. It was also noted that at the area of greatest crack width, there is a low spot in the ditch line that would collect water with no outlet for drainage.

Assessment

Based on past experience, the model of failure for this site is probably movement of the embankment as a result of moisture within the filVnatural ground contact zone. There may be organic material remaining along this contact that is decomposing, and the reduced strength produces continued movement. The water is probably the result of seepage into the fill from the ditch line. The sources of water include normal snowmelt, normal rain runoff, and the likely interception of some ground water that flows along the SU-AISU-B contact exposed in the ditch. This hypothesis is supported by the observation of no movement or water flow at the toe of the embankment, but a bowed fill slope that indicates internal movement, with the resulting road grade settlement. A preliminary interpretation of a cross-section through the wide-width crack area indicates the majority of the road width is embankment material. The lesser-width crack areas should have a corresponding smaller embankment.

Potential Mitieation Alternatives

The following alternatives are possible mitigation actions based on the data collected on the initial assessment visit. Additional investigation is necessary to better quantify the risks and provide a more accurate assessment on the likelihood of success. Estimated costs are given for comparison and are not to be considered an accurate project estimate. Further investigation and an accurate survey would provide more accurate estimates and possible refinements in mitigation alternatives.

In all alternatives, consideration should be given to providing positive drainage of the low spot in the ditch by installing a culvert (at an estimated cost of $1,000).

Interceptor trench. An interceptor trench in the ditchline would reduce the subsurface water flow under the roadway. This alternative by itself is considered to have limited success potential. The trench would reduce the subsurface water, but water could still penetrate under the pavement section along the ditch slope, especially during snowrnelt. Another condition not addressed is that no positive action is taken inside the slope if decomposing organics exist. The estimated cost of a interceptor trench installation is $2,500.

Retaining wall. A Hilfiker-style welded wire retaining wall could be placed at the outside edge of road. The excavation would be into natural ground, thus removing any organics that may be in the embankment. Positive drainage could be provided in the backfill. This alternative may require finding a site for waste material, as well as quality backfill. Some pavement would be removed and replaced. The estimated cost of installing the wall and pavement is $57,000.

Excavate and rebuild reinforced embankment. This alternative would involve excavating through the existing embankment to natural ground and removing unsuitable material. The new embankment could be built at 1:l using geotextile or geogrid reinforcement (figure 6B.6). This method would re-use the suitable existing embankment material, thus limiting the need for a new material source. Positive drainage could be provided if needed. The pavement would have to be removed and replaced. The fill slope should be seeded to limit erosion. The estimated cost of this alternative is $26,000.

Figure 6B.6,-Typical section for excavate-and-rebuild reinforced embankment.

Recommendation

The recommendation at this time is to install a culvert at the low spot along the ditch line and investigate further the alternative of excavating and rebuilding a reinforced embankment. Further investigation would consist of: (1) a survey of the area from below the road to the upper area of the old cut slope area and slightly east and west outside of the settled roadway section; (2) geotechnical field confirmation of soil layers and possible subsurface water; and (3) geotechnical confirmation of the stability of the old cut slope failure area. At this time it is anticipated that the additional field investigation would not require a drill machine.

Problem 2: Use of a Lightweight Fill

Case History: Slide #4-Horseshoe Bend Hill (Idaho), Federal Highway Administration, p. Contact Ron Chassie, Portland, Oregon.

Slide #4 on Horseshoe Bend Hill was a perennial maintenance problem. The slide was occurring through an approximately 400-foot length of sidehill fill constructed over a weak bentonitic clay foundation. The slip plane extended to 25 feet below the fill base. Sliding was occurring at the contact between the weak clay and underlying shale. The slide extended laterally several hundred feet downslope beyond the fill toe.

During the wet spring of 1983, slide movement accelerated to the point where the roadway was dropping at the rate of up to 1 foot per day. Continual maintenance was required that spring to maintain trafficability across the slide area. The accelerated slide movement necessitated development of a permanent correction scheme.

Preliminary design of a slide correction scheme called for subexcavation of the weak clay foundation below the fill toe to a depth below the slide failure plane and replacement with granular borrow to form a stabilizing shear key. The upper 15 feet of roadway fill, which was also comprised of clay, also was to be removed and

replaced with granular borrow. A deep drainage interceptor trench was proposed in the uphill ditch to intercept subsurface seepage. Estimated FOS = 1.5.

Because of previous successful sawdust fill slide stabilizations performed by the Washington State Department of Transportation, a sawdust fill alternative was considered. This alternative consisted only of removal and replacement with sawdust of the upper 15 feet of roadway clay fill and installation of the drainage interceptor trench (figure 6B.7). The shear key subexcavation was eliminated. Estimated FOS = 1.2.

The sawdust fill alternative was chosen and constructed, saving $185,000 compared with the shear key alternative.

The slide repair was completed in the fall of 1983 and has performed very successfully.

Figure 68.7.-Profile sketch for comparison of shear key to sawdust embankment.

Problem 3: Shifl in Horizontal Alignment

Case History: Vann SSI, USDA Forest Service, Wilfamette National Forest. Mitigation design by Mark A. Leverton, Engineering Geologist, August 1984.

Introduction

The Vann slope failure, on Road 1912 at milepost 0.68, is approximately 500 feet in length and extends from the road to the creek below with an estimated volume of 30,000 cubic yards. This failure has resulted in loss of road width and has been investigated four times (in 1980, 1983, 1985, and 1986) by the South Zone Geotechnical Group. This is a summary of the findings of those investigations and documents the work leading to the recommended design alternative made in the most recent investigation.

In December 1980, the geotechnical recommendation was to shift the road alignment 6 to 10 feet. The alignment shift was completed by approximately December 1981.

In February 1983, continued movement of the outside edge of the road and loss of width prompted another investigation. At this time, the site was recognized as a large area in movement. A subsurface investigation was begun. The recommended solution was to construct a retaining structure.

In March 1985, review of the previous analysis and projected lack of funding for a retaining structure led to exploration below the road to determine the feasibility of a rock buttress. The buttress was chosen as the new recommended alternative, and a design was produced.

In May 1986, the analysis was again reviewed. It was noted that there was insufficient confidence in the location of the failure surface near the road to proceed with the buttress design. Information gathered from inclinometer monitoring indicated the possibility of a deep failure. Stability analysis was performed using the deflection and water data gathered over the year. With this additional information, the decision was made to shift the road alignment into the hillside and avoid the slide area below the road.

Analvsis

The loss of road width is the result of a translational movement below the road extending to a slope break above the creek. The result of moment at the toe end of the slope is a response of the area immediately below the road. The continued loss of fill material has resulted in inadequate road width and horizontal alignment. Deep movement detected in the slope indicator casing installed within the slide mass produces major uncertainty in the foundation conditions for a retaining structure or buttress. Therefore, to meet the objective of acceptable horizontal alignment and road width, the alternative to shift the horizontal alignment was chosen (figure 68.8).

Figure 6B.8.-Plan view for the Vann SSI.

The height of cut involved in the new alignment required the combination of steep slopes and benching to produce an economical design. A bench cut was chosen to limit the free-fall height of material dislodged at the top of the cut. Slopes were chosen to parallel existing stable cut slopes, along with a bench width wide enough for a dozer, so that excavation could be accomplished efficiently. A critical section was analyzed using the program XSTABL (XSTABL v. 4.102 1992). The XSTABL input file is shown in figure 6B.9, and the XSTABL cross-section is shown in figure 6B.3.

Recommendation

The recommendation is to shift the horizontal alignment and construct a new cut slope using a combination of steep slopes and benching where necessary. Excess material should be removed from the outside road edge, and the slope below the road should be "rounded" from its near vertical slope to reduce the occurrence of sloughing if additional movement of the translational slide should occur.

PROFILE FILE: VANN3 4-07-93 1257 fl Sect R-R' Shift Alinnmcnt (bcnchcd)

Figure 6B.9.-XSTABL inputfile for sh@ed alignment.

Problem 4: Use of a Sag Vertical Curve

Case History: Bruler Slide. USDA Forest Service, Willamette National Forest. Original investigation by Cheryl L. Clark, Civil Engineer, August 7, 1984. Followup investigation by Rent! A. Renteria. Geotechnical Engineer, Febnrary 1991.

Introduction

Bmler Slide occurred at milepost 12 on the Quartzville Road in or around 1975. Each spring after the slide had moved during the winter, Forest Service maintenance crews "fi xed" the road by filling in the sag caused by the slide. This loaded the head of the slide where the road crosses it.

Field Investigation

The field investigation consisted of surface reconnaissance, surveying geologic cross- sections, subsurface drilling, and monitoring a slope indicator casing installed during the drilling operation. Seven holes were drilled at the site, and a record of the water level was kept for each hole.

Five material units (soil units A, B, C, D, and rock unit 12) were used for analysis. Their characteristics are summarized as:

Designation

I Plasticity I below the I BPL I non-plastic I above the plastic limit W L ) plastic limit I APL I

Description

Moisture

RU-12

I % Sand 1 90 1 55 1 4 0 1 80 1 60 1

SU-C SU-A

brown silty sand

wet

Consistency

Slide Mechanism

SU-B

silty sand with rock fragments

(RFs)

moist

loose

% Fines

Unified Classification (Field)

The topographies above and below Bmler Slide are indicative of a very large and very old slide consisting of a series of blocks. Two mechanisms were examined.

SU-D

Model 1. It appears that the "slide block" below Bmler Slide slipped out and was washed away by the stream. This removed the support for the next block (Bmler Slide). When combined with the added weight of road fill on the head of the slide, and a high water table, the loss of the abutment at its toe allowed Bmler Slide to move downhill. Slope indicator readings show a failure plane at approximately 25 feet below the road surface. However.

RFs with brown silty

sand

moist to wet

loose

10 % Coarse

10

SM

medium I soft I hard

light brown to yellow-brown silty sand with

RFs

wet

0

40

SM (MH fines)

remolds to silty sand with

RFs

moist

5 1 50+ 10

5

GP

10

SM (MH fines)

30

EEEE

tilting of the entire slope indicator casing suggests that a much deeper failure may exist.

Model 2. From the drill investigation, wet moisture conditions were noted only for SU-A, SU-C, and part of SU-D. This model assumes a perched water table on SU-B with minor surface saturation at the interface of SU-B. Water is allowed to rise into the fill (SU-D).

Based on monitoring for deflections in the slope indicator casing and observation pipes located below the road, model 2 was chosen for final analysis and design. Movement was detected under the road at a depth of 25 feet, but no movement was detected in the observation pipes downslope of the embankment. The deformation of the slope also suppotts the appearance of an embankment foundation failure. A plan view is shown in figure 6B.10.

Figure 66.10.-Plan view of Bruler slide.

Sloue Stabilitv Analvsis

A slope stability analysis performed by XSTABL (v. 4.10) indicates the stability of the slope to be most sensitive to the height of the phreatic surface or the height of

the embankment. The XSTABL input file is shown in figure 6B.11. The results for the analysis at failure are shown in figure 6B.12, the post-failure input file in figure 6B.13, and the post-failure results in figure 6B.14.

PROFILE Bruler Slide at Failun

17 13 132.0 3208.2 197.0 197.0 3220.8 245.0 245.0 3236.0 254.0 254.0 3248.4 292.0 292.0 3248.4 330.0 330.0 3257.8 344.0 344.0 3260.5 378.0 378.0 3283.0 399.0

WATER 1 62.40 9

132.0 3208.2 197.0 3220.8 245.0 3236.0 292.0 3247.0 344.0 3260.5 358.0 3267.0 382.0 3274.0 409.0 3275.6 420.0 3276.0

LIMITS 2 2

Figure 68.11.-XSTABL input file for Bruler Slide at failure.

Bruler Slide a t Failure 10 most critical surfacer, MINIMUM JANBU FOS = 9 9 6

3295

150 170 210 250 290 330 370 410 450 X-AXIS (feet)

Figure 68. I2.-XSTABL cross-section of Bruler Slide at failure.

Alternatives

Five alternatives were considered (costs are given in 1984 dollars).

- Trench drains (cost estimated at $10,000-$12,000). Trench drains would be a relatively simple way to lower the ground water level, particularly during the wet season when the slide moves. This should be an effective measure against large movements because the slide is most sensitive to the water level.

Lightweight fi l l with drain system (cost estimated at $l5,OOO-$l7,OOO). Although more expensive than the other alternatives, a lightweight fill combined with trench drains would not only "de-water" the slide, but also unload the head of it, thereby significantly reducing the driving forces. This alternative also produces the largest FOS against instability.

Road relocation (cost estimated at $9,000-$10,000). A road relocation would unload Bmler Slide, but the alignment of the road would be poor because it is advisable to avoid building the road through the marshy area above the slide.

Horizontal drains (cost estimate $1 1,000-$13.000). Drain holes would be drilled from below the road and pass through the slide plane. There is a risk of losing the drain system if the slide moves significantly.

Yearly maintenance. The existing culvert (damaged by slide movement) would be removed and replaced. Ditches to the culvert would be defined, and Forest Service maintenance crews would grade the road after movements but would not add additional material.

PROFILE FILE: BRULERI 4-07-93 7:43 ft Brulcr Slide Post-Failure

17 13 132.0 3208.2 197.0 197.0 3220.8 245.0 245.0 3236.0 254.0 254.0 3248.4 292.0 292.0 3248.4 330.0 330.0 3257.8 349.0 349.0 3270.0 382.0 382.0 3278.8 4W.O 404.0 3278.8 409.0 409.0 3275.6 419.0 419.0 3279.2 426.0 426.0 3279.6 445.0 445.0 3287.5 450.0 330.0 3257.8 409.0 292.0 3248.4 420.0 245.0 3236.0 420.0 197.0 3220.8 420.0

SOIL 5 IOOO 1100 .O 31.50 .WO 105.0 118.0 50.0 33.00 .WO 120.0 125.0 .O 34.00 ,000 105.0 115.0 .O 25.00 ,000 110.0 120.0 80.0 38.00 ,000

WATER 1 62.40 9

132.0 3208.2 197.0 3220.8 245.0 3236.0 292.0 3247.0 344.0 3260.5 358.0 3267.0 382.0 3274.0 409.0 3275.6 420.0 3276.0

LIMITS

Figure 6B. 13.-XSTABL input file for Bruler Slide postrfailure.

Recommended Alternative

Based on limited movement from 1985 to 1991, maintaining the road in the post- failure configuration of a sag vertical curve is recommended. New culverts should be placed at the low point in the ditch line and at a location to relieve ditch water before it enters the slide area. Water should be carried from the culvert outlets past the failed area limits with the use of cormgated plastic pipe. Without restoring the grade, a 27 percent increase is realized in the FOS without the additional costs of a lightweight fill or subsurface drainage.

Bruler Slide Post-Fai lure

3335 1 10 most critical surfaces. MINIMUM JAJVBU FOS = 1.271

3135

130 170 210 250 290 330 370 410 450

X-AXIS (feet)

Figure 6B.14.-XSTABL cross-section for Bruler Slide post-failure.

Problem 5: Use of a Crest Vertical Curve with a Buttress

Case history: Krassel Rock Slope Failure, South Fork Salmon River Road, USDA Forest Service, Payette National Forest. Original investigation by Michael Remboldt, Assistant Forest Engineer, September 9, 1992.

Problem

During road construction involving road widening excavation, the contractor undercut an existing rock slope by blasting. The failure was immediately adjacent to a major anadromous river which at the time had a spawning threatened and endangered (T&E) species. The failure also threatened the safety and stability of a major fire guard station and airstrip at the top of the slope.

Time was of the essence due to the T&E issue.

Investigation

Field-developed cross-sections were plotted to determine the size and extent of the failure and for design (see figure 6B.15). Road alignment at the failure was a long, straight tangent with minimal vertical grade.

The geology (Idaho Batholith) prevented stabilization alternatives, such as rock bolting, due to the random structure of the rock.

f CROSS SECTION STA. 0 1 5 0 - E- OF X,Vt-r-.

i- L CDLC

XSEC 0 + 3 5

CROSS SECTION STA. 0 + 3 5 (ROAD STA 1498+61) XSEC 0 + 5 0

(ROAD STA 1498+46)

Figure 68.15.-Field-developed cross-sections at Krassel rock slope failure. Slope angle approximately 40'. Rock type granite, Idaho Batholith.

There were ample sources of select fill. The ongoing road construction contract included a significant amount of H-pileltimber lagging retaining wall which had been recently installed, with some extra material left over due to contract modifications.

Solution

Several alternatives, including rock bolting and buttressing, were evaluated. An earth buttress was selected because of availability of materials, cost, road alignment, and time to construct.

(1) Failed mass end area (A) is approximately 200 square feet.

(2) Estimated rock density (y) is 155 pcf.

(3) Length through initial failure (0 is 30 feet.

(4) Estimated angle of internal friction (4) is 35".

(5) Backslope angle (8) is approximately 40".

(6) Calculated weight of failed mass is:

W = (200 sq. ft.)(155 pcf) = 31000 Iblft = 31 kipslft

(7) Calculated FOS against sliding of the failing mass is:

FOS = 1

W sin4O0

= 0.83 (<1.15) :. No good

(8) Let B be the mass of required earth buttress and calculate buttress size required for FOS = 1.15.

This type of problem is most readily solved by drawing a free body diagram showing the forces acting on the mass (see figure 6B.16). In equilibrium (FOS=I), the horizontal and vettical components of all forces acting on a body must sum to zero. Therefore a polygon comprised of all force vectors acting on the mass must close. (see figure 6B.17).

Figure 6B.16.-Free body diagram of sliding mass.

Figure 6B.17.-Force polygon,

In analyzing problems such as this, with rigid blocks acting against each other, the ordinary method of slices (OMS) can yield a FOS 8 to 10 percent greater than will a more rigorous force equilibrium analysis, resulting in a more costly, conservative solution. This is because of the simplifying assumptions of the OMS to satisfy only moment equilibrium.

From figure 6B.16, i t can be seen that movement will not occur without sliding at the block-to-block interface. At equilibrium (no sliding), there will be a force at the interface, the components of which are denoted Rx and Ry. It can also be seen that only horizontal movement can occur in BOTH blocks (assuming the buttress moves only on the underlying surface). For these reasons it is appropriate to use force equilibrium analysis rather than a moment equilibrium analysis.

The buttress mass solution is simplified by considering equilibrium in the horizontal (x) direction only. This is conservative, because the increase in the buttress normal force is greater than the corresponding decrease in normal force on the failing mass.

To solve for overall FOS, use the following equation for the forces acting in the horizontal direction:

FOS = 1.15 = C horizontal resisting forces C horizontal driving forces

Finally,

(9) Required minimum end area of buttress

I = riggo Iblft = 57.3 sq. ft. 120 pcf

One can check the effect of ignoring an interslice force by using a force equilibrium method such as Janbu's or Spencer's method. Output files from XSTABL (v. 4.1) using Janbu's method which illustrate the validity of the design assumptions are found in figure 6B.19.

The criteria for buttress geometry were that it: (1) must have end area greater than 57.3 sq. ft., (2) must support raised road subgrade (i.e., minimum width of 18 ft.), and (3) must be high enough to contain some large, loose rock fragments near the top of the existing cut slope.

A retaining wall system was constructed on the outside edge of the buttress because of geometry considerations (proximity to the river) and for protection of fisheries habitat (sediment delivery and disturbance). The wall system was constructed out of surplus H-piling and timber lagging and was anchored into the buttress itself.

Krassel Rock Failure with buttress Janbu Factor of Safety for Specified Surface = 1.163

Figure 68.18.-Krassel rock failure wirh buttress.

ISOTROPIC Soil P.rwf.r.

so11 u n i t 8.1- S-nr

Figure 68.19.-output file from XSTABL.

.................... ~~ilUnl...lght cohaslon W.t.r Unit Moist Sat. 1nt.rc.pt Mql. P a r u t e r constant 8urf.e.

NO. Ipcf) I ~ c t l 1p.f) 1d.q) Ru 1P.f) No.

Trial fallur. surfac. apc1fl.d by 4 swrdinat. point.

SLICE INTORIUIION ... cont1nu.d

Figure 68.19 (continued).-Output file from XSTABL.

I KRKSEL ROCK FAILURE 1 BUTTRESS DESIGN 1-1

Figure 6B.ZO.Selected stabilization alternative: modification of road geometry with earth buttress. Note: Retaining wall system for buttress face stabilization and protection of stream.

PROFILE - BUlTRESS/PILES KRASSEL ROCK FAILURE

---- - , . - m

Figure 6B.2I.-Selected stabilization alternative: plan view of vertical curve and buttress.

6C. Surface and Subsurface Drainage

Richard Van Dyke, Geotechnical Engineer, Siskiyou National Forest

6C.1 General There are two categories of water with which we will be concerned: surface water

hformation and ground water. Concentrations of surface water, seeps, springs, and vegetation changes indicate topographic changes that can provide critical clues about what may be happening with the ground water.

Water plays a very important role in the cause and mitigation of most landslides. It is important to learn as much as possible about surface water and ground water because changes in ground water levels and pore water pressures alter effective normal stress and, as a result, modify shear strength.

It is therefore critical that the source of ground water, changes in ground water levels, and the relationships among surface water, ground water, and the local geology be understood if landslide activity is to be managed.

6C.2 Surface The first step in understanding surface drainage is to obtain an accurate map that

Drainage depicts topography, geographic features, permanent and intermittent watercourses, seeps, and ponds (see sections 3 and 4 on field-developed cross-sections). Landslide magnitude and activity can sometimes be inferred by changes in drainage courses, blocked drainages, the deepness of incised runoff channels, and the location of seeps and springs. Thus, some type of record reporting the location of the hydrologic features and any changes should be started. Observations of surface runoff are best collected during periods of high rainfall.

Both natural and constructed changes to the affected areas must be taken into account. Construction and landslide activities may alter surface water courses, change ground water conditions, and activate otherwise stable areas. Evidence of previous locations of channels should be noted.

Maps of seeps and springs (caused by the intersection of an aquifer with the ground surface) and identification of the associated aquifers can be correlated to subsurface information and help establish the geologic map for the site. Changes in the flows of seeps or springs may be indicators that can be correlated to slide activity. Generally. if free flow is continued and is adequate to handle peak flow, the aquifer acts as a natural drain, and pore pressure cannot build up. However, if the volume of water exceeds the capacity of the aquifer, the pore pressure may increase to the point that a land mass becomes unstable.

A very important consideration in the design and construction of any repair work associated with a slope stability problem is that the surface water and any collected ground water should be carried away from the unstable or potentially unstable slopes.

It is almost always less costly to deal with the collection and control of surface water than of ground water. Some methods of dealing with surface water are:

Installing surface channels and trenches that redirect or intercept surface flows.

Sealing the exposed surface cracks and other locations where the surface water gains access to subsurface aquifers.

Reshaping the surface topography of a slide. This can eliminate ponding and seal unseen cracks. Care must be taken to ensure that the distribution of soil loads does not decrease any resisting forces or increase any driving forces. Reshaping an area may seal natural seepage, or it may redirect seepage paths and cause new problems.

Such radical methods as sealing the entire surface of a slide mass to prevent infiltration are generally only reasonable in extreme situations and have a very low success rate.

6C.3 It is generally agreed that ground water is the most important trigger for landslides.

Subsurface Thus, one of the main goals of a subsurface investigation should be to locate the

Drainage aquifers and to install equipment to monitor ground water activity (see section 3D). This equipment also can be used to evaluate the success of the repair. The drillers and geologist must be aware that locating the ground water is a main objective of the investigation. Accurate information must be available in order to perform any stability analysis and to design the control and possible changes to the ground water. Once the surface and subsurface water have been adequately mapped and the relationship between them determined, design alternatives can be developed and evaluated for cost and practicality.

There are inherent risks associated with subsurface drainage systems, due primarily to the difficulty in determining ground water flow patterns and locations and geology and soil types. The reader is referred to sections 3D and 6D.2 for investigation techniques that will aid in determining ground water location and soil parameters.

The success of a drainage system depends upon the decrease in pore pressure on the critical failure surface (see section 4E). Its rate of flow is not a reliable indicator of how successful the application is; some subsurface systems drain very little water but are very successful in stabilizing a failure. Flow rate is actually a function of soil type and whether or not the water is confined.

Two of the most common subsurface drain systems used to manage ground water are cutoff trenches and horizontal drains.

6C.3.1 Cutoff Cutoff trenches are ditches dug to intercept and drain ground water away from areas Trenches of instability. The trenches usually consist of perforated pipe with geotextile or drain

rock and divert the intercepted water to daylight at an acceptable location. Some typical cutoff trench systems are show in figure 6C.1.

Figure 6C. I.-Typical cutoff trench system.

If a critical aquifer is located within an excavatable distance of the ground surface, then a cutoff trench can be a very good solution. The technique and actual

construction are not easy and can be very expensive, especially if the trench is more than 10 feet deep; trenches greater than 10 to 15 feet deep become impractical to construct. If the water is perched on a surface of relatively low permeability and the trench can be constructed down to that level, then most of the ground water can be diverted from the area of instability. If the perched aquifer handling the ground water is very thick or deep, then a cutoff trench may divert only a portion of the aquifer water. However, if enough water is collected, then the slope still may be stabilized.

The cutoff trench design includes three elements: the separation layer, the trench geometry, and pipe size and slope.

Separation Layer. The function of the separation layer is to keep soil fines from clogging the drain system and reducing or eliminating the capacity to carry away intercepted water. At the same time, the separation layer must have adequate hydraulic capacity itself to allow the ground water to enter the drain freely.

The most economical and simple separation materials used today are geotextiles. The equivalent opening size (EOS) of a geotextile to provide adequate separation must be specified. The reader is referred to Christopher and Holtz (1984) for information about estimating the necessary EOS. For required flow characteristics, which should be at least as great as the soil surrounding the cutoff trench, the permeability of the geotextile must be specified. The designer also should consider the puncture strength characteristics of the geotextile in order to ensure minimal damage during installation.

Trench Geometry. The trench must extend to the minimum depth that will ensure desired ground water interception. In cases where the trench does not terminate in a low permeability layer, the response of the ground water on the downstream side of the trench must be estimated, and the trench depth must be adjusted as necessary. Trench length should allow for uncertainty in the geometry of the unstable or unsuitable geology.

Pipe Geometry. Most trenches will be drained using either a perforated pipe at the bottom of the trench or a composite vertical drain system. In either case, the hydraulic capacity of the pipe or composite drain must be at least as large as the calculated ground water recharge capacity for the area of interception. Section 6D.3 outlines how to determine ground water recharge capacity and discusses pipe hydraulic capacity. We recommend that the pipe be sized for a greater capacity than is estimated from ground water recharge calculations. The amount of additional design capacity depends on the probability and consequence of drain failure, such as separation layer failure (clogging the system) or ground movement (reducing cross-sectional area of the drain).

Sample Problem

The following sample problem is typical of stability problems that can be controlled with proper drainage methods. It is presented in a report format, including the background, survey data, investigation, testing, and exploration of the stability analysis. The stability analysis was performed using XSTABL (XSTABL v. 4.102 1992).

Descri~tion of Area

This shallow translational slide-12-18 feet deep, 5 W 0 0 feet long, and 150-200 feet wide-is located on a south-facing slope. Approximately 180-200 feet of roadway is affected by this failure. However, the failure itself extends 300-400 feet above and below the road (see Figure 6C.2). Movement of the area occurred yearly at varying rates depending upon the amount of precipitation. Many trees within the slide area had fallen over as a result of the slide. Several springs appeared above and below the road during the high precipitation season. In general, the ground water within the slide was very high during the winter and spring months. During years of extreme slide activity, road use was restricted due to vertical and horizontal displacements that occurred across the road prism.

Backeround

The first report concerning this area is dated 1980, when it was reported that the road had been closed due to displacements of as great as 6 feet. Several inconclusive investigations were conducted between 1980 and 1986. These consisted of seismic work, core drilling, and test pits. The road was realigned in 1983 approximately 15-20 feet uphill, but that winter the slide activity, which had been only below the road, progressed above the mad and took the present configuration shown in figure 6C.2.

Field Investieation

An additional subsurface investigation of the site was made in October 1986 using 8-inch hollow-flight augers. That investigation had several objectives: to characterize the subsurface geology; to install water monitoring pipes to allow observation of the fluctuations in ground water; and to monitor the movement of the slide and determine the depth of the slip surface. Three drill holes were located along the central axis of the slide (see figures 6C.2 and 6C.3). Evidence of sliding and movement was found just above the refusal layer (a layer of mudstone) in each hole; see the drill logs (figure 6C.4) and the test pit section (figure 625) .

Figure 6C.Z.Si te plan of the failure.

Figure 6C.3.-Cross-section of the failure with trenches installed.

/ Dr i l l Hole # I

1 Dri l l Hole #Z

Dri l l Hole #3

Laboratory Test Data

Figure 6C.4.-Drill informtion and laboratory data.

Ground Surfosc

FEET 0 5 10 15 20 25 30

Figure 6C.5.-Test pit section

Laboratow Testing

The initial testing on the samples obtained from the test pits in 1984-85 included classification, bulk densities, residual direct shear tests, Proctor tests, and shear tests on some remolded samples. Similar tests were conducted on some of the samples obtained from the augering.

Monitoring

Various types of monitoring were established to characterize ground water activity and slope movement, survey points, and stake arrays. Initially, ground water levels were read manually; eventually, electronic continuously reading ground water equipment was installed in several of the monitoring pipes. Figure 6C.6 shows the hydrograph for the monitoring of drill hole 3 during the period when ground water rose from the summer to winter extremes used in the stability analyses shown in figures 6C.7 and 6C.8.

Drill Hole 83 - Deep a n d Shallow October 19. 1988 t o Jonuory 5 . 1989

0 1

-20 1 1 1 1 1 1 / 1 / 1 1 1 1 / ~

I 9 25 1 5 10 I5 20 25 30 5 10 15 20 25 30 5 October or ember Dec.rnber

MONTHS -

Figure 6C.6,-Hydrograph of ground water monitoring data for drill hole 3.

BEAR CAMP BEFORE FAILURE/LOW WATER ~ m b u Foclor of Sofell for Specified Surfoc. = 1.207

1

15 75 135 195 255 315 375 435

X-AXIS (feet)

Figure 6C.7.-XSTABL srability analysis with low warer before failure (plor and input file).

BEAR CAMP BEFORE FAILURE/HIGH WATER Jonbu Foclor 01 S o f d y (or S p ~ i f i . d Surfas. = 0.992

0

15 75 135 195 255 315 375 X-AXIS ( feet )

Figure 6C.8.-XSTABL stabilify analysis with high water before failure (plot and input file). Slope is unstable.

Analvsis Procedures

Using field-generated cross-sections (figure 6C.3). several ground water conditions and various soil parameters were analyzed to determine the factors of safety for each situation. A parametric analysis was performed to estimate values of the shear strength and cohesion of the soil and to confirm the ground water conditions that caused failure. These soil parameter values were then used in analyzing conditions prior to failure (figures 6C.7 and 6C.8).

The goal was to stabilize the road. It was anticipated that by installing one trench drain above the road and a second at the inside edge of the road, most of the water that was causing the instability in this localized area could be diverted (see the XSTABL analysis in figure 6C.9). The construction of the trenches to intercept the mudstone also would allow the backfill material to act as a shear key (see section 6G). Note that at some distance below the downhill drainage trench the slope could still be unstable because the undiverted ground water could presumably rise to a critical level and cause the failure to become active once again.

BEAR CAMP H.w./DRAINS/SOIL COMPACTED Jonbu Foclor of 50faty f o r Specif ied Surfoc. = 1.267

0

I5 75 135 195 255 315 375 435 495

X-AXIS ( feet )

Figure 6C.9.-XSTABL analysis of site with a trench drain above the road and one at the inside edge of the road (plot and inputfile).

Four XSTABL analyses were run. Figure 6C.7 shows an analysis of the section with ground water in the summer condition (when no movement was occurring). Figure 6C.8 shows the slope with high ground water and an unstable slope, which is typical of a winter condition. Figure 6C.9 shows the same slope with the trenches installed and backfilled. This condition shows an increase in the FOS to 1.267 during winter ground water levels. The fourth analysis, of the effects of drainage alone, is shown in figure 6C.10, where the shear key effects have been eliminated from the analysis (that is, the trench material has the same strength properties as the in-place soils). For safety, the stability of the open-cut excavations should also be checked for conditions that might exist during construction.

BEAR CAMP H.W./DRAINS/NO SOIL CHANGE Jonbu F a d o r o f Saf.fy for Speci f ied Sur fac l = 1.123

1

0

15 7 5 135 195 255 315 3 7 5 4 3 5 4 9 5

X-AXIS ( feet )

Figure 6C.10.-XSTABL analysis for sire with drainage alone (plot and inpurjile)

6C.3.2 Horizontal Horizontal drains are also used extensively to drain ground water from unstable or Drains potentially unstable areas. On many landslides the ground water either does not

conform to an easily mapped or linear surface, or it is located at such a depth that draining it with surface trenches is impractical. The use of horizontal drains may solve such problems.

Horizontal drains consist of a system of slotted or perforated pipes installed horizontally, or nearly horizontally, into a slope (see figure 6C.l I) . They work by intercepting ground water and reducing pore pressures in zones of instability. Installation requires a slope face from which the drains can be installed and can daylight.

Contracts for the installation of horizontal drains usually provide that the exact location of the drains depends upon information gained during the placement of the drains. This may reduce some of the initial investigation costs. Costs for installing horizontal drains are generally less than for other subsurface drainage methods.

Section 6D provides a full discussion of horizontal drains.

Figure 6C. 11.-Horizontal drain

6D. Horizontal Drains

Michael T. Long, Engineering Geologist, Willamette National Forest

6D.1 General Section 6D discusses the advantages and disadvantages of using drilled horizontal

Information drains for landslide stabilization and offers some techniques for preconstmction, analysis, construction, and post-construction phases of a project. The term "horizontal" may be a misnomer in that the drilled-in drains are normally "inclined" to match the subsurface geologic conditions; however, because most practitioners are familiar with the term, it will be used throughout this section.

Many slope stabilization authors have said that the three most important factors in slope stability are "water, water, and water." The reader is strongly advised to refer to sections 4E, 3C, and 3D prior to committing resources to the methods in this section. Most of the following material is taken from Long (1986, 1991).

Historically, horizontal drain systems used for landslide correction have had varying degrees of success. This variability of success has also been apparent in pre- construction, exploration, engineering design, construction and post-construction monitoring. The USDA Forest Service Pacific Northwest Region has accomplished a number of successful horizontal drain projects. As a result, an approach and several low-cost alternative technology tools have been developed for completing the work. This section is intended to be a summary and guide to the project approach and a supplement to the current body of literature available on horizontal drain systems.

Horizontal drains have proven to be a cost-effective alternative to such major slope stabilization repairs as unloading and buttressing when subsurface water is involved in the mechanics of failure. In situations where interceptor (cutoff) trenches may have excessive depth or the material involved may need extensive construction shoring, horizontal drains present a more practical construction option. Horizontal drains have been installed into unstable slopes to lower the phreatic surface or relieve confined ground water pressures for some time. The California Department of Transportation pioneered drilled installations in 1939 (Royster, 1977). Since then, California has installed more than 1 million linear feet of horizontal drains, helping to develop state-of-the-art horizontal drilling and installation methods (Smith, 1980). The list of successful case histories throughout the United States is extensive. Material and site conditions in these cases have varied from discontinuities in overconsolidated clays to silty sands with rock fragments larger than 1 cubic yard (Vandre, 1975; Barrett, 1980; Trolinger, 1980; Royster, 1980; and Spitzer et al. 1986). Brawner et al. (1982) documented cases where an induced vacuum system improved performance of horizontal drains in both soil and rock slope applications.

Although the body of literature contains many horizontal drain case histories, there are few papers relating to pre-construction investigative techniques and quantitative design methods, especially in anisotropic heterogeneous material. A 1984 case

6D.2 Pre- Design Information- Investigation Techniques

6D.2.1 Area Reconnaissance

6D.2.2 Ground Control Suwey

history by Long discussed such methods on one project. The Pacific Northwest Region has accomplished a number of successful horizontal drain projects using contract as well as government force account drill crews. As a result of these successes, methods and procedures for accomplishing the investigation and design of this type of project have been refined. The project planning process now includes a variety of techniques, including geophysical, geochemical, and drilling exploration methods, and ground water and drain system modeling using Darcy's Law, Manning's Equation, and a method of drain end spacing developed by Prellwitz (1979). A number of inexpensive, low-technology, reliable exploration and installation techniques have also been developed (Neal and Williamson, 1980; Williamson, 1989; Prellwitz and Babbitt, 1984).

The success or failure of any geotechnical project depends on the degree of accuracy of the subsurface model (soil and rock characteristics and horizontal and vertical distributions) and the ground water regime (distribution, volume, and flow characteristics). The most sophisticated analysis and design efforts are useless, within reasonable economic limits (regarding over-conservative design parameters). unless there is an equal effort employed toward technical confidence in the subsurface investigation phase. The following techniques can be considered for any geotechnical exploration effort but should be considered more closely in a drainage design for slope stabilization. Depending on the project scope and budget, some methods may not be economically practical but are presented as part of a suggested list of available techniques.

This includes a complete literature search and aerial photo review as well as a general field reconnaissance to determine the history, process, and origin relating to the previous geologic and construction events that produced the present morphology. In particular, rock and soil units should be designated and classified and point sources for ground water infiltration should be identified.

This is best accomplished by a survey crew with an electronic distance measuring (EDM) device for precision and accuracy in control-point or hub-line monitoring of surface movements. The centerline and lateral cross-sections should be surveyed under the supervision of a qualified engineering geologist or geotechnical engineer, who can identify features important to the interpretation and stability analysis. These will be staked on the ground for future reference and included on the plan map as cross-section points. Aerial photography targets can be set and tied in at this time if the scope of the project warrants photogrammetric mapping. A topographic map and an adequate number of cross-sections for analysis and design should be generated from the survey. Consideration for periodic control-point and hub-line re-survey should be planned at this time.

Another survey technique that may be used is the field-developed cross-section method (appendix 3.5) in conjunction with traverses to tie together the sections, developed in the Pacific Northwest Region for internal geotechnical project support (see section 3C). This method uses a cloth tape, handheld compass, and clinometer and gives the investigator freedom of scale and the ability to operate independently of a survey crew. One limitation of this method is its tendency to compound errors in cross-sections over 200 feet long. This method is ideal for small organizations

without survey support, or with modest budgets, who need to begin or continue a project to completion without delay.

6D.2.3 Subsurface Prior to further exploration efforts, a first approximation of the subsurface material Interpretation distribution and slope failure geometry should be made on the centerline and lateral

cross-sections. This procedure compels the engineering geologist or geotechnical engineer to use. the scientific method approach of multiple working hypotheses leading to commitment to a working model. This, in turn, promotes confirmation or revision of the hypothesized model. As further exploration proceeds, this method provides a tool for preliminary slope stability analysis and helps to define further exploration efforts. This method has proven to be a valuable tool for learning interpretation and for providing project managers a process for serial review.

6D.2.4 Drive Another low-level-technology exploration device developed in the Pacific Northwest Probe Exploration Region is the portable drive probe assembly (Williamson, 1989). The probe

assembly is inexpensive, lightweight, and retrievable. It consists of Cfoot sections of 112-inch threaded galvanized pipe which are advanced below the surface by an 11-pound sliding hammer free-falling 41 inches. The lead section is closed by means of a pipe plug and perforated with 3116-inch drilled holes so that water levels may be checked as the pipe is advanced. Normally, blow counts are recorded at 6-inch intervals. Any change in the number of blow counts indicates a change in density and shear strength relative to the overlying or underlying soil profile. These data will assist in making an initial interpretation of the subsurface. Measuring the static water level in the hole through the pipe with a resistivity meter will also allow the investigator to determine exactly where ground water was encountered. When apparent refusal is reached, additional dynamic force (by means of physically adding acceleration to the hammer) may be applied to ensure that a rock fragment has not been encountered. The pipe may be left in place as an open stand pipe piezometer. This will provide an extension to the data obtained from more costly efforts of core drilling. This device has been used effectively to depths of 30 feet (see appendix 3.6).

6D.2.5 Electrical A portable electrical resistivity instrument with a simple Werner configuration can Resistivity help define areas of subsurface water concentrations. The profiles may also be used Profiling as an inexpensive method to plan drilling exploration to ensure optimum locations

for boreholes to intercept saturated zones. Profile lines with electrode spacings of 60 feet, recording drops in apparent resistivity 25 feet below the surface, have been used effectively.

6D.2.6 Drilling Exploration

Hollow-stem augers and continuous standard penetration test sampling should be used to obtain subsurface samples and soil strength estimates without introducing drilling fluids into the borehole. The hole should be advanced far enough beyond the interpreted failure plane, using a core barrel assembly if necessary, to properly seat and seal any borehole instrumentation (such as inclinometer casing or piezometers). At a minimum, open stand pipe observation wells should be installed. Observation wells must be installed within the failed mass at points near or adjacent to analysis cross-sections in order to determine the effectiveness of the horizontal drains and the final calculated FOS.

6D.2.7 Permeability Testing

To fully define the ground water model and obtain data to construct lateral cross- section end areas, we suggest additional borings and wells outside of the lateral failure limits. These additional borings and data will also help in horizontal flow net construction to plan for the most effective drain locations. As the borings and observation well installations are being completed, the design must be considered relative to in-situ permeability testing. Slug tests, or maintained-head tests, have proven to be efficient. This testing should he completed in order to obtain the hydraulic conductivity for hydraulic analysis. The upslope and downslope boreholes and measured static water levels should be plotted on the appropriate cross-sections in order to determine the hydraulic gradient for the model. All drilling exploration should be completed with a qualified on-site engineering geologist or geotechnical engineer as inspector.

In situ permeability tests should be performed in the material to be drained (U.S. Department of the Navy, 1982b). If, from the exploration borings, the material is determined to be hydraulically zoned containing perched water tables or "pockets" of isolated water, testing in an adjacent borehole is preferred in order to take advantage of the opportunity to test each zone separately. Testing can be accomplished in the observation wells if care is taken in placement and sealing of the wells within the zone to be tested, or if pneumatic borehole packers are used.

6D.2.8 Ground There are several methods for confirming a point source or sources of water Water Tracing infiltration into a system and for determining hydraulic conductivity between

observation wells. The tracer dyes rhodamine WT and fluorescein have been used successfully in ground water modeling efforts in the Pacific Northwest Region (Long, 1986; Stables, 1979; Turner Associates, 1971 and 1973). Both dyes can be used concurrently to determine separate point-sources at the surface or introduced into observation wells (which could discharge at the same location). Water samples can be collected directly or by placing a packet of activated charcoal at discharge points in the failed mass. Charcoal packets may also be connected to a line lowered into a borehole.

Sodium chloride has also been used to trace ground water and to determine hydraulic conductivity by measuring the relative electrical conductivity at the discharge point (Megahan and Clayton, 1983). This method is limited, however, to shorter travel distances due to the lower concentrations of sodium chloride achievable and detectable.

A weir should be const~cted at each spring and seep observed at the slide scarp, lateral margins, and toe. The weir can be constructed using natural material and fitting a 1-inch polyvinyl chloride (PVC) overflow pipe through the weir so that flow measurements can be taken of all the seeps and summed for water budget estimates. The weir discharge must be directed away from the failed mass.

6D.2.9 Water A general water surface contour map should be constructed from the static water Surface Contours level (SWL) readings, converted to elevations, from the observation wells. The water

surface contours then can be superimposed on the topographic contours and used to plan the drainage design. The water surface contours generally can be considered to be equipotential lines, and the general subsurface flow path or paths can be estimated

constructing flow lines perpendicular to the equipotential lines (figures 6D.1 and 6D.2).

WATER SURFACE

I INTERPRETED

I I

Figure 6D.I.--Geologic cross-section.

/

6D.2.10 Test Drain Test drains should be installed to confirm final drain locations. The test drains Installation should be located according to the water surface contour model in areas which

suggest "piezometric valleys," or concentrations or convergence of flow lines.

6D.3 Drainage System Design

Horizontal drains are installed to reduce excess pore-water pressure, thereby increasing stability. Excess pore-water pressure may be a result of confined ground water at or below the failure surface (piezometric) or unconfined ground water flow (phreatic). In the confined state, horizontal drains will relieve "pressure" on the failure surface by lowering the piezometric head. The initial discharge may be relatively small or large but then will dissipate quickly. In the unconfined state, horizontal drains may be used to intercept and divert ground water and lower the phreatic surface. The initial discharge will likely be continuous over time but will respond to seasonal rainfall.

In the unconfined state, in order to design a horizontal drainage system, the data obtained from the subsurface investigation and subsequent interpretation must be used to obtain the values necessary for Darcy's Law, Manning's Equation, and drain end-spacing equations. The following is a recommendation for the approach to the design (the number of drains needed, inclination, length, and effective end spacing).

6D.3.1 Ground To determine the volume of water entering the failure area, which ideally Water Recharge corresponds to the desired interception volume, Darcy's Law states: Capacity

Q = r n where:

Q = discharge in gallons per minute (gpm) K = hydraulic conductivity in feet per day i = AWAL hydraulic gradient (ftlft)

A = cross-sectional area in square feet.

The cross-sectional area (A) is determined from the interpreted cross- sections perpendicular to the long axis of the slide, taking into consideration the current (steady state flow) and potential (transient state flow) rainfall recharge areas.

The hydraulic gradient (i) can be obtained from the difference in head and horizontal distance between upslope and downslope observations wells or from the water surface contour map.

Hydraulic conductivity (K) is determined by borehole falling-head and maintained-head tests performed in the exploration phase.

This process should be used for each water zone, if possible, and summed for the total discharge.

Current cross-sectional area discharge calculations should then be compared with the sum of all seep discharge points to determine whether there is a reasonable correlation. If the correlation is not reasonable, engineering judgment and/or further investigation must be used to resolve the discrepancy (sum of all seeps do not approach the total discharge potential, or values assigned to the model parameters are high or low).

See also section 4E.1.4.

6D.3.2 Number of The number of drains necessary to accommodate the computed volume is based on Drains Needed Manning's Equation, which states:

where: V = discharge velocity in feet per second N = roughness coefficient = 0.009 S = percent slope: range 2 to 15 percent

R = hydraulic radius = area

wetted perimeter = 0.031 ft. (for 1-112 inch pipe flowing full).

The capacity of 1-112 inch inside diameter (ID) slotted PVC pipe flowing full in various slope gradient configurations can now be calculated by:

where:

Q = discharge rate V = discharge velocity (from Eq. 60.2) A = pipe end area (0.012 f t2 for 1-li2 inch pipe).

Drain grades and lengths are then determined from cross-section analysis to optimize the drain length in the saturated zones. An array of grades and lengths will probably be necessary to accommodate the geomehy of each site.

From experience, we suggest that the number of drains in the preliminary design be based on 25 percent flow capacity to compensate for poorly performing or nonfunctional drains in the system.

Consider incorporating interceptor drains above the failed mass, as well as relief drains within the failed mass.

6D.3.3 Slot Width The slot width of the drain pipe must be small enough to prevent piping of fines and Spacing through the opening, but large enough to prevent clogging. The U.S. Department of

Transportation recommended that the slot width be one-third of the D,, of the soil for slotted underdrain systems where D,, is defined as "the decimal number of the size of soil particles for which 85% of the soil is finer." (US. Department of Transportation, 1976).

Cedergren (1977) suggested

D= < 1.2. Slot Width

Driscoll (1986) recommended that the slot spacing match the porosity of the water- bearing layer. However, experience has shown that with Driscoll's spacing, if the

open area exceeds 20 percent, the strength of schedule 80 slotted PVC pipe may be reduced enough to break during installation.

60.3.4 Effective The ground water drawdown is determined by the distance between drains. In most End Spacing and instances, the drains are installed in a fan array to reduce drill setup time and Drawdown working area excavation. The maximum and minimum effective end spacing should

be calculated to ensure that the minimum required drawdown is being achieved within the failure mass. As spacing between the drains increases, drawdown decreases to zero. As spacing decreases, drawdown is affected by radial flow interference between drains.

The reader is referred to Prellwitz (1979) for a comprehensive method for determining end spacing and drawdown based on modifications of the Hooghoudt and Glover-Dumm equations for transient state flow and the site-specific slope and phreatic surface geometries. Figures 6D.3 through 6D.7 and table 6D.1 describe the mathematical methods involved, which are quite labor-intensive. Prellwitz (1990) has programmed all the algorithms for an HP41 programmable calculator, which greatly simplifies the process.

Notes for Hand Solution for End Spacing and Drawdown

1. Calculate X,, (figure 7 from Prellwitz, 1979) first to begin step 3 on analysis table.

2. For calculation 6 (equation 5). for negative values of Xu,

Y, = tan2Xt - 2YuJu + Y,,'

3. For calculation 10 (equation 9). for negative values of X,,

where:

Y , is the conversion to common axis (figure 6)

4. For calculation 1 l (drawdown), steps 25 and 31 must converge using a series of assumed values in step 12.

(Figure and equation numbers from Prellwitz, 1979.)

F o r MINIMUM E f f e c t i v e S p a c i n g

WHERE :

Figure 6D.3 .4eometry for horizontal drain and spacing for hand solution method.

M h x l n U n E f f e c t o r e E n d S p a c ~ n g

WHERE :

AYD - Y COORDINATE AT DRAIN /DRAINAGE BARRIER INTERCEPT

X COORDINATE AT DRAIN /DRAINAGE BARRIER INTERCEPT

X u o - X COORD l NATE AT ORA l N D l SCHARGE

X u - A X D - X u ,

2 2 Y u o = h S I n O C o s O I l + T a n O ) ( I * T a n 8 )

Yu 0 Xu, = -

T a n p

2 2 2 Yu s T a n 0 X u - 2 Y u o Xu + Y u o

THEN:

G 1 VEN L:, 1 0 1 - s ~ - l o ~ - l s ~ - 2 0 ~ A xu - C H M I N GIVEN INTERVAL

:ALCULATED h Y U = CWW3E I N CALCULATED INTERVAL

TUESE VALUES DEFINE THE CURVE OF PHREATIC SURFACE D

Figure 6D.4.-Maximum effective end spacing calculation.

8 Table 60.1.-"Table I" from Prellwitz (1979) for hand calcularion of horizontal drain end spacing and drawdown.

m Figure 6D.5.-"Figure 7"from Prellwirz (1979). u

Figure 6D.d-"Figure 2" from Prellwifz (1979). z

Figure 60.7.-"Figure 6"from Prellwitz (1979) 9

6D.3.5 Collector System

The collector system for a large array of drains may consist of several options; 12- or 8-inch cormgated pipe, anchored above ground with steel posts and 114-inch wire rope, with Zinch feeder hoses clamped to the drain ends, has been used successfully (see figure 6D.8). A buried pipe may be considered if further movement is not expected in which locating a break would be difficult. Cormgated polyethylene pipe may be used in a subsurface installation. Surface installations require material strength adequate to withstand animal traffic and vandalism. If freezing conditions are likely at the site, a covered manifold system with a concrete drop collection box at the drain discharge points may be designed. It is important, however, to consider access to each individual drain in order to monitor post-construction discharge and to facilitate cleanout. A sudden decrease or increase in flow or a change in water color may indicate further movement.

'. CTCR WIT ASSEMBLY

0.D. WCWRILOHTALORUN PIPE

Figure 6D.8.-Typical collecror sysrem details.

6D.4 Construction Considerations

6D.4.1 Suggested Construction Practices

Horizontal drains are commonly constructed using rotary drilling methods. Drains are installed by advancing a Cinch drill casing to the desired length with a knock off tri-cone roller bit. At the end point, the casing is rotated in the reverse direction and the slotted PVC pipe is inserted through the casing, thereby knocking off the bit. As the casing is removed, the PVC pipe and roller bit remain in the drill bole.

Construct the collector system prior to drilling to accommodate anticipated drain flow.

If excavation is necessary for constructing drilling pads, ensure that the drill pad cut slope will continue to be stable under long-term leaky drainage conditions. If necessary, design and construct a rock buttress for local stability

prior to drilling. Cut slope failure after construction can shear the drains and cause system failure.

If possible, drilling should take place during wet weather in order to make field judgments for modifications on drain locations, concentrations, direction, and inclination based on observed discharge from completed drains.

Determine drain end-elevation using a manometer (a hose connected to the end of a flowing drain and elevated to equilibrium) or a pressure meter to check for vertical casing drift. Correct the next hole as necessary.

Absorbent wipes, stream booms, silt fences, and straw bales are effective in controlling sediment from drill cuttings and machine fluid leaks. If the site is adjacent to an environmentally sensitive area, a spill plan may be necessary.

Slots have been installed successfully in both the up and down positions; however, consideration must be given to segments of blank pipe in the drain. Blanks should be installed in any segment that is not penetrating the water- bearing zone to prevent migration of ground water into otherwise dry areas. Blanks should be installed in at least the last 20 feet of drain if toe or cut slope stability is a concern. Drains should not penetrate further than 15 feet beyond the failure surface or the drainage barrier.

For discharge end protection, and to prevent root growth in the drain, a 3-inch galvanized metal pipe should be installed over the discharge end into the drill hole to a minimum of 5 feet, then grouted in place.

6D.4.2 Inspector A qualified engineering geologist or geotechnical engineer should direct the drilling Duties installation to ensure that design criteria (angle, elevation, location) are met, and to

make any field modifications. The inspector should be responsible for the following duties:

Setting fore and aft site stakes for hole alignment prior to arrival of large metal objects that will affect compass bearings.

Measuring drill casing slope as the hole is "collared in."

Recording advance rate, water return, and water color. Monitoring path of drill casing for surface indications of drilling fluids in any adjacent or upslope tension cracks.

Noting material changes with casing advancement.

Sampling drill cuttings in order to be aware of when the failure surface or the soil/rock interface has been reached.

Having predictive tools (interpreted cross-sections and drill logs) at hand to assist with estimating failure plane and material boundaries.

Recording final length, slope, end elevation, and discharge rate from each completed drain.

6D.4.3 Alternative Construction Method

6D.5 Post- Construction Monitoring

Ensuring that the drain number is marked with a metal stamp on the galvanized sleeve for future reference.

The Oregon State Department of Forestry has experimented with hand-driven horizontal well points as an alternative low-technology method of installing horizontal drains into slopes. The technique uses sections of 112-inch steel pipe inserted into a perforated 1-inch PVC pilot sleeve with a hard plastic well point. The well point is then advanced into the slope by means of a slide hammer acting against the ID-inch steel pipe in a manner similar to the drive probe method. Sections of PVC and drive pipe are added as the drain is advanced. The drive pipe is then rotated and removed from the PVC upon completion. This system has been used effectively with horizontal advancement up to 40 feet. With some modification, this system could be used with a head-frame assembly and power cathead to drive steel well points and galvanized pipe to greater depths with faster advancement rates.

Post-construction monitoring of a drainage system should be calculated into the project budget and schedule. Monitoring should be done weekly for the first month after project completion and monthly thereafter until there is a high degree of confidence in long-term stability. Monitoring frequency may be increased during periods of excessive rain, rain-on-snow events, or spring runoff. Recommended monitoring activities include:

Measuring static water levels in all observation wells.

Obtaining inclinometer readings.

Re-surveying hub lines and control points.

Measuring discharge from each individual drain.

Measuring total system discharge.

Installing rain gauge on-site and recording precipitation.

All water levels, total system discharge, and rainfall can be recorded automatically using pressure transducers calibrated to the head of water in a casing, flume, or rain gauge and connected to a battery powered automated data logger. Prellwitz and Babbitt (1984) gave detailed instructions on construction of these devices.

A low-technology method for recording the highest water level reading in an observation well is to place finely ground cork in a length (equal to the depth of the borehole) of 3/8-inch clear flexible plastic tubing with a piece of sponge to close the bottom end. The riser tube is then lowered to the bottom of the observation well and fastened in place with tape. As the water rises in the observation well, the cork in the plastic tubing rises. As the water level decreases, the cork adheres to the sides. The tubing can then be removed from the observation well at any time, with the highest level of cork representing the highest level of water that has been in the well. The cork can then be flushed back to the bottom of the riser tube, and the tubing re- inserted in the well.

6D.6 Case History Summaries

Drain effectiveness is often reduced within 5 to 10 years due to root growth, piping of fines, and bacteria. The California Department of Transportation recommends that an ongoing inspection program be initiated and a cleaning schedule established if reduced discharge volume is noted (Smith, 1980). Cleaning is accomplished using a high pressure water pump with a self-propelling jet nozzle attached to a length of hose inserted the full length of each drain.

The five case histories summarized below represent between $300,000 and $400,000 in cost savings over the next lowest-cost stabilization alternatives considered.

Camp Five Slide

Location: Failure Mass: Install Dates: Linear Feet: No. Drains: No. Locations: Drain Length: Drain Slope: Slot Size: Soil Type: SWL Drop: Total Discharge: Install Cost: Final FOS: Investigation:

Drill Holes: Drive Probe: Other:

Willamette National Forest, Oakridge, OR 250,000 yd3 12/83 to 1/84 7,800 feet 52 7 65 to 240 feet 2% to 15% 0.050 inch Silty Sand (SM) 14 feet High of 576 gpm $150,000 with 2,000 yd3 buttress 1.20

2 1 0 Resistivity Profiling, Dye Tracing, Permeability Testing, EDM Survey, Aerial Photogrammetry

The slide increased in size from 30,000 cubic yards to 250,000 cubic yards in the final failure prior to drain installation.

Fairview Sanitary Landfill

Location: Failure Mass: Install Dates: Linear Feet: No. Drains: No. Locations: Drain Length: Drain Slope: Slot Size: Soil Type: SWL Drop: Total Discharge: Install Cost: Final FOS:

Bureau of Land Management, Coquille, OR 80,000 yd3 4/87 to 5/87 3,337 feet 19 3 175 feet 3 degrees 0.051 inch Sandy Silt (ML) 5 feet 10 gpm $45.000 1.25

Investigation: Drill Holes: 9 Drive Probe: 0 Other: Survey

Drains were placed to intercept subsurface water in siltstone and into landfill "pits" to drain infiltrated water. The majority of flow was from intercepted subsurface water. Immediate maintenance concern arose with buildup of iron bacteria in the drains.

Powder Creek Slide

Location: Failure Mass: Install Dates: Linear Feet: No. Drains: No. Locations: Drain Length: Drain Slope: Slot Size: Soil Type: SWL Drop: Total Discharge: Install Cost: Final FOS.: Investigation:

Drill Holes: Drive Probe: Other:

Willamette National Forest, Oakridge, OR 55,500 yd3 8/88 to 10188 2,754 feet 20 1 110 to 225 feet 3% to 10% 0.090 inch Silty Sand (SM) 10 feet 4 to I6 gpm $30,000 1.35

L

Survey

Drill pad construction at the toe of the slide was difficult due to saturated conditions. The drains have effectively stabilized the road prism, which had been moving for 20 years despite previous attempts at stabilization (piles, relocation, syphon wells). The drill pad backslope was not buttressed, and drain slots were installed the entire length. As a result, the pad backslope failed and sheared all drains 15 feet behind the discharge point. The drains are still effective; however, the toe will have to be restabilized.

Quentin Slide

Location: Failure Mass: Install Dates: Linear Feet: No. Drains: No. Locations: Drain Length: Drain Slope: Slot Size: Soil Type: SWL Drop:

Willamette National Forest, Blue River, OR 66,600 yd' 2/87 to 3/87 4,087 feet 19 2 215 to 296 feet 2% to 14% 0.050 inch Silty Sand (SM) 10 feet

Total Discharge: High of 40 gpm Install Cost: $60,000 Final FOS: 1 .05 Investigation:

Drill Holes: 13 Drive Probe: 10 Other: In situ permeability testing

The system has not been monitored since installation. To achieve original road alignment, further stabilization methods must be considered.

Highlands Interchange Slide

Location: Failure Mass: Install Dates: Linear Feet: No. Drains: No. Locations: Drain Length: Drain Slope: Slot Size: Soil Type: SWL Drop: Total Discharge: Install Cost: Final FOS: Investigation:

Drill Holes: Drive Probe: Other:

Oregon DOT, Sunset Hwy., Portland, OR 300,000 yd3 10158 to 11/58 5,900 feet 22 7 80 to 450 feet 3% to 5% l/2-inch drilled holes Sandy Silt (MH) 30 feet 100 gpm Unknown Unknown

17 zero Unknown

This system was installed in 1958 using Zinch iron pipe with drilled holes to stabilize a slope failure in a residential neighborhood as a result of highway widening. The drains are still operational today after 31 years of service. Maximum discharge from one drain is still 15 gpm.

6D.7 Design Appendix 6.1 presents a design example for a major slope failure which was

Example-- stabilized in December 1983 on the Willamette National Forest (in the central portion

Camp Five of the Cascade Range in western Oregon) with a network of horizontal drains that

Slide continually intercept and divert a localized aquifer.

6E. Buttresses

Richard Van Dyke, Geotechnical Engineer, Siskiyau National Forest

6E.1 General One of the more common and effective methods of stabilizing an unstable or

hformation potentially unstable slope. is to increase the resisting forces by installing a buttress at the toe of the failure. This method works particularly well on small- to medium-size slides, the most common size along Forest Service roads.

The resistive-force buttress placed at the toe of the failure can take various forms, such as riprap rock buttresses, gabion baskets, gravity retaining walls, reinforced soil walls, counterweight fills, pile systems retaining walls, and anchored wall systems. Even though the concept is rather elementary, the design, stability analysis, and construction considerations should be approached with the same consideration afforded more complex engineering problems.

The engineer should address the following concerns when designing a buttress:

Foundation bearing capacity

External stability

Internal stability (if the buttress is a mechanically stabilized embankment)

Surface and subsurface drainage

Construction

6E.2 Rock Buttresses

6E.2.1 General One of the most common and cost-effective buttresses used in the Forest Service is one constructed out of rock. It is generally used where additional road width is required or a failure has occurred above the road and the existing material will not stand at the desired cut slope angle. In mountainous areas on road cuts where the natural sideslope. may be 1-1/4:1 or I:], a buttress must be ~0I IS t~c ted at 3/4:1 or steeper to attain minimum road width.

The buttresses typically are used where the road cut intersects soil pockets that occur in a more stable formation; thus, the areas that need supporting are relatively short and do not involve long stretches of the road. Rock buttresses are commonly analyzed as gravity retaining walls. A typical rock buttress is shown in figure 6E.1.

Figure 6E.l.-A typical rock buttress.

Maximum face slope angle of the rock buttress is a function of rock shape (angular rock interlocks better than does rounded rock), rock size, and the skill of the equipment operator. Face slopes of rock buttresses can be constructed as steep as 114H:l V for large, angular rock. Density of rock buttresses is a function of the rock specific gravity (G,) and void ratio. Typical rock buttress densities are 120 to 130 pcf for G, of 2.65 and void ratios of 0.2 to 0.3. Rock buttresses work because of the high internal angle of friction within the buttress and their drainage characteristics.

6E.2.2 Foundation As noted above, the foundation must be adequate to support the weight of the Bearing Capacity buttress. The foundation should be analyzed using the general bearing capacity

equation (US. Department of the Navy, 1982a):

B q.,, = CNc + y D N , + Y N , (6E. 1)

2 where:

q,, = ultimate bearing C = cohesive strength of the soil y = unit weight of the soil D = depth below ground surface B = width of the buttress and No N, and N , are dimensionless bearing-

capacity factors that depend on the soil friction angle and the shape of the assumed failure zone.

6E.2.3 External The combination of the rock buttress and the slope should be analyzed for overall Stability external slope stability. Particular attention should be given to the changes in ground

water behind the buttress. The buttress also should be checked for stability against overturning and sliding along its base.

6E.2.4 Drainage Because water plays such a major role in the stability of slopes, drainage concerns must always be addressed. A separation layer of geotextile should be placed at the back and sides of the rock buttress. Geotextile selection criteria include soil gradation, ground water recharge rate, and construction.

With a buttress of rock with a geotextile appropriate for the soil conditions, drainage at the toe should not be a concern as long as the water is directed away from the base of the buttress.

6E.2.5 Construction

It is very important that the stability of the slope be considered during all phases of construction. The construction may have to be timed to correspond to a period when the water level and other parameters are at the least critical point. The buttress may have to be designed to allow construction in stages so that the entire toe of the slide is not removed at one time. Each of the various stages of construction should be subjected to separate stability analyses.

Buttress construction considerations include stability of temporary excavation and support of the geotextile separation barrier. Temporary backslope excavation stability must be assessed. Large excavations for the placement of the buttress in poor, wet soils are hazardous. This limits the maximum length of excavation at any given time. The geotextile must be supported against the temporary backslope so that it is not tom or punctured and seams are not pulled apart during placement of the rock.

6E.3 Earth Buttress

6E.3.1 General An earth buttress placed at the toe of the slide mass can also be effective in providing resistant force. This type of buttress can be used in the same situations as rock buttresses: existing cut slopes, new cut slopes, existing slide stabilization, and constructing an over-steepened fill (see section 6F). The earth buttress is analyzed as a gravity retaining structure and shares most of the design considerations of a rock buttress. The engineer should evaluate local foundation bearing capacity and external stability in the same manner as for rock buttresses. Figure 6E.2 shows an earth buttress.

6E.3.2 Internal The earth buttress can be either a layer-placed earth embankment or a mechanically Stability stabilized embankment (MSE). In each case, the internal stability of the buttress

must be assessed. The stability of the face against slope failure should be checked for simple embankments, whereas the stability of the internal reinforcement against pullout, shear, and corrosion must be evaluated for MSE's.

Figure 6E.2.-Earth buffress.

6E.3.3 Drainage The main difference between earth buttresses and rock buttresses is that a positive drainage method generally has to be incorporated into the design of the earth buttress to provide internal drainage and stability, allowing the use of the total weight rather than the submerged unit weight in the analysis (see section 4B). These drainage methods usually involve installing a backslope drain along the contact of the natural material and the compacted fill (see section 6C).

There are two common types of backslope drain-a prefabricated drain, consisting of drainage fabric and an impermeable three-dimensional hard plastic core sheeting, and a traditional rock drain with fabric. It is more economical and practical in most Forest Service repa i rc in which the repair and slide are relatively small (100 to 200 feet across)-to use the prefabricated drain. This is due in part to the fact that most of these failures are distant from adequate commercial gravel sources (resulting in high haul costs) and to the difficulty in constructing a good rock drain on slopes.

An advantage of the prefabricated drains is that they provide a positive impermeable surface if they are installed correctly. Care must be taken in the selection of the drainage core material and the correct drainage fabric.

6E.3.4 Concerns regarding backslope stability during excavation are essentially the same as Construction those outlined for rock buttresses. However, even more caution must be exercised in

placement of the drainage system. Damage to the drain may have severe consequences because saturation of the earth buttress may cause internal failure to the buttress and subsequent mass failure of the slope.

Both the rock and earth buttresses can be constructed with common earth- and rock- moving equipment and, in many instances, using material that is close at hand, thus reducing the costs of construction in remote areas.

6E.4 Retaining Walls

6E.4.1 General In cases with limited access and right-of-way, rock and earth buttresses may not be feasible. Various types of retaining wall systems can be used in these situations to provide the resisting force.

Stabilizing a slide mass with a retaining wall requires consideration of backslope geometry and surcharge loading that might not be considered in an ordinary retaining wall design.

Two common types of non-gravity walls used for buttressing are H-pilehimber lagging and tied-back walls. H-piles can be driven or placed and grouted in pre- drilled holes. Tied-back walls can be reinforced concrete, timber lagging, or other facing materials tied into the backslope using drilled and grouted steel ties to provide what is essentially a massive gravity structure.

Criteria for selecting the type of retaining wall include required temporary backslope excavation (both lateral and height requirements), drainage characteristics, availability of materials, cost, and ease of construction. The reader is referred to the Forest Service Retaining Wall Design Guide (1979) for specific information regarding retaining wall selection and design (a revised edition is in preparation).

6E.4.2 Foundation The foundation soils beneath the retaining wall footing should be evaluated for local Bearing Capacity bearing capacity.

6E.4.3 External Overall stability of the wall and surrounding slopes should be evaluated. The Stability designer should check for adequate factors of safety against sliding of the wall

against the base and against overturning around the lower outside comer of the wall. In the case of pile walls, the piles must develop adequate passive resistance (below the ground surface at the wall face) to resist overturning and sliding.

6E.4.4 Internal Internal stability for mechanically stabilized walls has been discussed in section 6E.3. Stability Tie-back walls need to be evaluated for tie-back resistance to pullout.

6E.4.5 Drainage Drainage must be provided if there is any chance that the retained backslope will be saturated. Drainage can be provided by geo-composite drain systems, rock drains, or openings in the wall face. Drain design criteria include soil gradation and ground water recharge capacity. Constructing drainage appurtenances behind certain wall structures may be difficult and in some cases impossible. The reader is referred to section 6D for information on drain design.

6E.4.6 Construction

Construction considerations are the same as for rock buttresses (section 6E.2).

6E.5 Buttress Sample Problem

6E.5.1 Description of Area

6E.5.2 Background

This sample problem is an example of road-related stability problems that can be stabilized by means of a buttress. The example is presented in a format similar to a report, including the background, survey data, investigation, testing, and explanation of the stability analysis. The stability analysis was performed using XSTABL (XSTABL Ver. 4.102 1992).

The failure site is located on a N35E-facing slope. The road is a paved two-lane major access road to the forest. The failure is in a through-fill section that was originally constructed with a 1-1/4:1 slope that extended to the edge of Elk River, approximately 40 feet below the road. The area uphill from the road is on a 35-percent to 40-percent slope and shows evidence of past failures in the form of hummocky and broken ground. Considerable surface water and shallow subsurface flows not confined to drainage channels are evident above the road. Figure 6E.3 shows a plan view of the site.

L E G E N D

ULLb SCARPS 0 10 20 30

SITE PLAN

Figure 6E.3.7.Site plan of the Elk River failure.

The site initially was investigated after a 4-inch vertical displacement of the road surface disrupted traffic. That investigation consisted of creating several shallow auger holes and doing a seismic survey. The resulting recommendation was to install a 6- to 8-foot-deep cutoff trench along the inside edge of the road to control the ground water that was assumed to have caused the failure. Such a drain was installed in August 1987.

6E.5.3 Field suwey

Another failure, involving 55 linear feet of the road and extending 17 feet into the traveled roadway, occurred in December 1987 following 2 days of approximately 12 inches of rain. Subsequent investigation revealed that the end of the drainage pipe for the trench drain and the drain rock for the trench were left exposed in the ditch line. Neither of these practices is standard procedure for cutoff trench installation, and they may very well have caused the injection of surface water into the trench, resulting in the second failure. Very little horizontal displacement was evident at the top of the failureindicating a relatively steep headscarp and a deep failure surface. The total vertical displacement at the scarp was 6 feet. The failure widened downslope, and near the river the affected area was 90 to 100 feet wide. Evidence of the location of the toe of the failure and where it had been displaced was visible near the river. Several smaller surface failures had occurred within the larger failed area.

Shortly after the second failure, a site survey was conducted to help plan the drilling investigation and to gather information to be used in the stability analysis. A site map (figure 6E.3) used to design the drilling plan was constructed, a baseline was established, critical surface features were noted, and field-developed cross-sections before and after failure (figures 6E.4 and 6E.5) were developed.

- - - - ASWYID IO~L umt - arrro ow o w

U O w..YI , ~ I I I U 1 , 0 * .

ASSUMED SLOPE BEFORE FAILURE

A

/ I ; I I I I I I I / ; I I I I 0 0 10 2 0 30 4 0 50 60 70 80 90 100 110 120 130 140 150 160

FEET

Figure 6E.4.-Cross-section A-A before failure.

6E.5.4 Field Investigation

6E.5.5 Laboratory Testing

6E.5.6 Analysis Procedures

AFTER FAILURE

EXISTING GROUND

,,-BCOROC' SC"8ST

BEDROCK SCHlSl

1 ASSUMED FAILURE ~ r w n mw* SURFACE

- -

FEET

Figure 6E.5.-Cross-section A-A after failure.

Several soil samples were collected from the site to be used in laboratory testing to determine the parameters to be used in the stability analysis. A sample of what was believed to be slip surface material was collected from below the slide area. Other material to be used in determining unit weights for the soils was also collected.

The subsurface investigation was started in January 1988 and consisted of augering and sampling five drill holes at the site. The drill hole locations are indicated on the site map (figure 6E.3). A summary of the information from the two pertinent holes is given in table 6E.1 and is shown on the cross-sections (figure 6E.5).

Testing of the materials collected during the surface and subsurface investigation was completed in January 1988. Residual direct shear testing was done on the material thought to have come from the slip surface. The test results indicated that the soil had a friction angle of 25 degrees and a cohesion of 70 pounds per square foot.

The first step of the analysis is to verify field and laboratory data and to establish the soil parameter values to be used in the design. This was done by making assumptions of the variation of the FOS of the slope as a function of seasonal changes in ground water. Starting with the initial field and laboratory values for the soil parameters, slope stability analysis was performed using the seasonal ground water assumptions. Repeated analysis was done using variations of the soil parameter values until the assumed seasonal variation in the FOS was satisfied.

6E.5.7 Construction

The assumptions for FOS were that during summer ground water levels, the FOS is greater than 1 .O, and for winter ground water conditions FOS is less than or equal to 1.0. Assumptions were also made regarding the variation of the ground water profile and FOS afer failure occurred.

The soil conditions that represented the problem were established:

@ cs Y Unit - (decrees) (osn Soil 1 25 0 125 Soil 2 25 0 115 Soil 3 30 0 128 Soil 4 30 50 124 Rock Buttress 45 0 145

Experience as to what values and conditions are realistic is invaluable because they vary from one geographic location to another. Once a realistic combination of values for the soil parameters and existing conditions was established, various possible solutions were explored to determine how best to stabilize the failed slope. Figure 6E.6 shows a cross-section with a buttress and road fill in place. Figure 6E.7 shows the XSTABL plot of the slope before failure, figure 6E.8 shows the slope after failure when it has reached a temporarily stable condition, and figure 6E.9 shows the analysis of the slope stabilization using a rock buttress.

The initial FOS during winter conditions was increased from less than 1 to nearly 1.13 with the installation of the rock buttress. The buttress also protects the slope from erosion.

Although analysis showed that the buttress would have stabilized the slope (FOS = 1.125) without additional drainage, a prefabricated backslope drainage blanket was installed. Because the road material needed to be recompacted anyway, the drainage system was justified and increased the FOS. The road was removed to a depth of approximately 12 feet, and the drainage system was constructed using a geo- composite drain.

Several intermediate stability runs not illustrated here were made to verify the stability of the slope during construction. It was assumed that construction would take place when the ground water was at its lowest level and that some of the material at the toe would have to be removed to place the rock buttress. If the stability of the slope could not have been maintained during the construction phase, then either alternative methods of construction (such as stage construction, where only portions of the slope are worked on at a time) or an alternative stabilization method would have been required.

Table 6E. I.-Logs for Drill Holes 1 and 5,

Drill Hole No. 1:

DEPTH (FT) DESCRIPTION BLOWS16 IN. MOISTURE

&6 Road filVrocky 4-6 Damp

6-12 Highly disturbed 4-8 schist/old fill material

Damp

12-16 Top soillroots, etc. 5-8 Damp

16-17 Impermeable black1 15-17 grey siltylclay very plastic

Damp

17-18 Saturated gravel 9-10 Wet

18-23 Decomposed schist 18-22 Damp rock in-place

2 3 4 2 Hard rock schist no STPlaugered

Vote the saturated gravels directly below the impermeable clay layer at 17 feet,

4 water monitoring tube was installed in the drill hole and the water level itabilized at 16 feet. The water rose 4 feet to 11.8 feet as a result of 9-112 inches )f rainfall on January 11. Ground water had fallen to a depth of 15.8 feet when it vas checked on January 27.

Drill Hole No. 5:

DEPTH (FT) DESCRIPTION BLOWS16 IN. MOISTURE

&5 Sandyhopsoil 3 4 Wet

5-5.5 Gravelly -- Wet

5.5-1 1 Decomposed schist 13-20 rock in-place

Damp

1 1 4 8 Hard rock schist no STP/augered

BUTTRESS AND ROAD FILL IN PLACE C

0

0 10 20 30 40 50 60 70 80 90 100 110 120 130 140 150 160

FEET

Figure 6E.6.-Cross-section A-A with buttress and road f i N in place

ELK RIVER BEFORE FAILURE Jonbu Foclor o l Safely lor Sprcilied Surlacs = 0.991

loo 1

P R O ~ I L ELK RIVER BEFORE TAILML 11 I 11.60 13.10 n.50 17.SO 17.00 21.00 11.00 18.00 78.00 78.00 5 1 . 4 0 98.00 $ * . D O 51.50 107.50 107.50 51.90 116.00 ll6.00 51.90 111.00 71.20 19.60 6I.60 6,. ' 0 21.00 1SJ.00 26.50 ll.50 68.IO 61.80 11.10 111.00 24.10 17.10 69.10 69.10 11.40 111.00

S O I L

60 80 100 120 I 4 0 1 GO

X - A X I S ( feet )

Figure 6E.7.-XSTABL analysis of the Elk River site before failure.

ELK RIVER AFTER FAILURE 100 , 3 mart cr i l icol surlnces. MlNlKUM JANBU FOS 1 1.055

0

0 20 LO 60 80 I00 120 140 160 X - A X I S ( f ee l )

FlOIIL CLX 1 I V U A R U FIILVIIZ 15 15

13.60 15.20 1s.10 20.60 17.10 12.50 37.00 27.40 41.00 11.60 51.<0 15.10 60.20 11.90 73.00 41.80 78.70 0.70 Sl.<O 50.10 94.SO 50.10 94.40 50.40 s*.*o l5.SO

107.10 56,lO 115.CO 15.60

SOIL

Figure 6E.8.-XSTABL analysis of the Elk River site afler failure.

ELK RIVER AFTER FAILURE W/ BUTTRESS 3 most crilicol surfoc... MINIMUM JANBU f O S = 1 . 1 2 5

t 0 20 A0 60 8 0 1 0 0 1 2 0 1 4 0 160

X-.AXIS ( feet )

Figure 6E.9.-XSTABL analysis of the Elk River site after failure with buttress.

6F. Soil Slope Stabilization-Reinforced Fills

Gordon Keller, Geotechnical Engineer, Plumas National Forest

6F.l General Information

6F.l.l Introduction

Reinforced embankments (reinforced fills or mechanically stabilized earth slopes) consist of an embankment fill built up in compacted lifts, with layers of a reinforcing material (geogrid, welded wire, or geotextile) placed throughout the embankment. The reinforcing material adds tensile resistance to local (face) or deep-seated shear failure in the embankment. Reinforced fills placed with a 1H:lV or steeper face slope can offer an economical alternative to retaining structures for those sites where the ground is too steep to catch a conventional 1-112:l fill slope, yet is flat enough to catch an oversteep reinforced fill. A slope. range from 67 to over 150 percent can be achieved, depending on the reinforcement and facing measures used. Reinforced fill heights have ranged from 15 to 50 feet (5 to 15 m) on forest projects, and the highest fill built anywhere to date is a 115-foot-high (38-m-high) 1:l fill in California.

The spacing of the primary reinforcement is chosen to add the tensile strength needed to support the oversteepened fill slope and to prevent a deep-seated slope failure. Spacing typically varies between 2 feet (0.6 m) and 5 feet (1.5 m) and depends on soil parameters, height of the fill, and strength of the reinforcing material. Intermediate reinforcement, placed between layers of the primary reinforcement, typically consists of narrow, e.g., 5-foot-wide (1.5-m-wide), strips of low-strength geogrid placed along the fill face on a 1-foot (0.3-m) vertical spacing. They prevent local failure on the oversteep face between the primary reinforcement layers and prevent failures due to construction equipment loading. Figure 6F.1 shows typical reinforced embankment configurations.

6F.1.2 The primary advantage of a reinforced fill is cost savings through the ability to fit a Advantages and fill slope on a tight (steep) location. A stable reinforced embankment can be Disadvantages constructed in a location where a conventional 1-1/21 fill will not catch, yet a

retaining wall and its associated facing costs can be avoided. The cost for reinforced fills has been between that of a conventional fill and a retaining structure needed for that site. Reinforcing materials needed are roughly similar in quantity to those used in a wall, so the principal cost savings come from avoiding use of forms or installation of facing members, both of which are typical in wall construction. Installation of the geogrid is quite easy, but commonly geogrid placement requires hand labor. It is important to note that the geogrid must be correctly oriented.

Incorrect orientation may lead to a material strength lower than that required by design and can waste money. Some geogrids are biaxial, with the same or similar design tensile strengths in either direction. Others are uniaxial, where one direction is substantially stronger than the other, so correct orientation of this material is critical. All materials should be installed in accordance with the manufacturer's recommendations.

-

Potential Failure

Surface .. . . - - _ _ _ - - -

Conventional Unreinforced Slope

. Reinforced Slope Slide Stabilization

Figure 6F. /.-Typical reinforced slope embankment configurations.

One disadvantage of geotextiles and geogrids is the unknown long-term behavior of the material. Different manufacturers have different performance histories for their products, but neither the long-term creep characteristics nor the long-term environmental degradation of the synthetics is well known. This issue is a subject of considerable current research and is currently addressed by the practice of applying various environmental and construction factors of safety (FOS's).

An advantage of reinforced fills is their ability to increase the stability of any slope, particularly after a failure, to improve on the construction behavior of poor-quality soils, such as silts and clays, and to make slopes "fit" when constrained by space. Improved compaction at the edge of the slope by equipment operating on the secondary reinforcement decreases the tendency for surface sloughing and face erosion.

Finally, use of reinforced fills can save on materials if fill material and polymer reinforcement materials can be used in corrosive soils or harsh acidic, saline, or alkaline environments because of their general resistance to chemical degradation.

6F.1.3 The primary application of reinforced embankments is to construct an embankment Applications with an oversteepened face slope to fit a particular site and avoid the use of a

retaining wall. This is primarily a cost consideration. Other applications include sites where the area of disturbance must be minimized, a slope toe is constrained to a particular elevation or limit, or additional fill bench width is needed (with a constrained toe location). Reinforced embankments built into a roadcut have also been used as a drained "buttress" in a limited space application.

Reinforcement is also used in the base of embankments as a way to improve the overall embankment foundation stability by creating a "wide footing" over a soft subgrade or poor foundation soils. This application is discussed in such publications as Christopher and Holtz (1984) and is not commonly considered a "slope stabilization" technique, so it will not be further discussed here.

Although reinforced embankments are typically constructed with select frictional backfill, successful projects have also been constructed using marginal, silty backfill materials as well as clay-rich plastic materials. The successful use of poor backfill material requires special design and construction considerations discussed later.

The reinforcement concept can also be applied in biotechnical slope stabilization where layers of brush or vegetation, such as willows, are built into the fill. This concept is commonly known as brush layering. Vegetation adds slope reinforcement from the woody debris, breaks up the slope for erosion control purposes, and offers long-term root strength for slope stability. Brush layering should be regarded as a superficial slope stabilization measure, rather than a "designed" deep-seated slope reinforcement. However, the concept of primary reinforcement with material such as geogrid plus the use of vegetation in the outer several feet of the lift or as secondary reinforcement is likely ideal.

6F.2 Necessary Predesign Information

Information needed for predesign and design of reinforced fills includes site and materials information similar to that needed for most other slope stabilization alternatives. This includes knowledge of the local site geology, ground water conditions, and mechanism of failure, and having an accurate site cross-section, as outlined in section 3. Information on backfill materials' strength and characteristics-including frictional and cohesive strength parameters, gradation, plasticity, chemical characteristics, and other properties-is typically necessary. Methods of obtaining this information are discussed in section 4 and such other references as Forest Service Retaining Wall Design Guide (1979).

Information on materials properties such as design long-term strength, creep characteristics, stress-strain relationships, durability, corrosion resistance, and resistance to installation damage is needed on the reinforcing material used in the slope. Much of this information is available from manufacturers' literature, and the topic is the subject of ongoing research.

Pullout resistance properties of the reinforcement material in contact with the soil are particularly important for the function of the structure. Pullout resistance is mobilized through a combination of both interface friction and passive soil resistance against transverse elements of a reinforcing grid. Geogrids with a high percentage of open area and thickness to the grid offer the highest passive resistance, while a geotextile primarily mobilizes interface frictional resistance. Pullout resistance design

values or factors are typically obtained from pullout tests performed by manufacturers on a range of soil types or on backfill to be used on a large project. Standardized pullout test methods are currently being developed.

6F.3 The Stability of reinforced slopes and fills is typically analyzed using versions of

Stability Design conventional limit equilibrium analysis, modified to account for the added tensile strength of the reinforcing material. A circular or wedge-shaped potential failure surface is assumed, and the relationship between resisting and driving forces or moments determines the FOS. Detailed design procedures for reinforced fills have been well summarized by Christopher et al. (1990) and outlined in text by Koerner (1990). Tall reinforced structures--over roughly 20 feet high--and those in critical areas should be designed using comprehensive slope stability analysis. Computer programs that handle reinforced slopes include PCSTABL6 (1980). TENSLO (TENSLOI ver. 2 1991). STABGM (STABGM ver. 9.85 1985). UTEXAS2 (UTEXAS2 ver. 1.21 1 1987). and a comprehensive design program currently being developed by the Federal Highway Administration (FHWA). XSTABL (XSTABL ver 4.102 1992) can be used for reinforced fill analysis, as shown in the design examples.

Simplified hand solutions have been developed and published by the FHWA, and step-by-step design charts are available from various manufacturers, such as Tensar Corporation. Design charts are commonly used without detailed stability analysis to design small projects. This approach appears appropriate for rural, non-critical applications on relatively small fills (under 20 feet high) in good soils. Also, the cost of reinforcing materials is a relatively small percentage of the total repair cost of a site, so optimization of a design is not critical and may not be cost effective.

Conventional design of a reinforced fill involves determining:

Desired final slope geometry and loading conditions;

Selection of fill;

Determination of fill and soil properties;

Forces acting on the soil structure (which must be resisted with reinforcement for internal stability);

Required number and type of reinforcement layers;

Vertical spacing and embedment lengths of the reinforcement layers; and

Overall external stability of the reinforced mass against sliding, deep-seated failure, and settlement.

Appropriate minimum factors of safety for the various aspects of internal and external design are available in the FHWA design procedures. Recommended seismic design procedures use a pseudo-static analysis outlined by the FHWA. Specific design procedures and examples are also available in the US. Forest Service Retaining Wall Design Guide (1979). A new edition of this guide is in process.

Design procedures developed to date are typically suited to sites which have deep-seated rotational failure surfaces or to new fill embankments where long reinforcing members can be used. However, the most common type of slope failure in mountainous terrain is a shallow debris slide, so the design procedure must be adapted to repair shallow fill failures over bedrock. Basically, the reinforcement material and layer spacing is used as required from published design procedures for the fill height needed, but the reinforcement length is shortened from the recommended value to fit the site geometry. Reinforcement length is determined to prevent sliding or overturning of the reinforced mass. Because reinforcement spacing is developed based on the most critical, typically deep-seated, failure surface, and because only a shallow failure is likely to occur on a debris slide, the design spacing is conservative. Internal stability of the reinforced mass is thus conservative.

External stability, however, may be marginal. The critical mode of failure for this type of design has been sliding along the fill-native material or fill-bedrock interface. Limit equilibrium analysis of the fill as a sliding block has shown an FOS with respect to sliding of less than one unless the interface is stepped, forcing a failure through more competent in-place material or through the reinforcing members. If a geocomposite drain is used under the fill and placed along the fill-native material interface as is commonly done, sliding failure is even more likely because the soil-to-geosynthetic friction angle is typically assumed to be only 60 to 90 percent of the friction angle of the soil itself (Eigenbrod and Locker, 1987). A graded aggregate drain can minimize this problem.

As determined on reinforced fills designed to date with a benched soil-fill interface, adequate terracing of the interface can achieve a desired FOS against sliding of 1.5. Terracing, however, produces additional construction difficulties, particularly for installation of a drain, which may increase the total project cost.

6F.4 Unique Construction and Preconstruction Considerations

6F.4.1 Internal Internal reinforcement material most commonly used in reinforced fills is a uniaxial Reinforcement or biaxial geogrid material made of polymers, such as polyester, polypropylene, or Materials polyethylene. Each has its own creep, pullout, strength, and elongation

characteristics that should be considered in design, Ideally, stress-strain characteristics of polymer geogrids or geotextiles should be as similar to the soil as practical. Materials characteristics are commonly available from the manufacturer, but test methods should follow those developed as industry standards, such as Task Force 27 (AASHTO) and Geosynthetic Research Institute (GRI-Drexel University) standards of practice and test methods.

Allowable design tensile strength of reinforcing materials, particularly geogrids, is determined from an ultimate strength laboratory value modified by several FOS, including factors for creep deformation, installation damage, chemical and environmental degradation, and weakness at seams and joints. Creep tests should be conducted for a minimum duration of 10,000 hours and extrapolated to a minimum 75-year design life.

Pullout resistance, which reflects the soil-reinforcement interaction, is a significant design consideration. An advantage of geogrids or wire mesh is the high percentage of soil-to-soil contact through the grid. Long-term pullout performance is a function of pullout capacity (preferably determined from pullout tests), allowable displace- ment, and long-term displacement (a function of creep characteristics).

Welded wire mesh or other metals, such as chain link fencing, that are considered inextensible will have less creep and long-term displacement, different load-deformation characteristics and will be more susceptible to corrosion than are extensible polymers (geogrids). Galvanized coating or additional sacrificial metal are used to resist corrosion. Corrosion issues are addressed in Elias (1990) and other publications.

In reinforced embankments constructed with marginal or clay-rich embankment, material nonwoven geotextiles are desirable due to their "wicking" or transmissibility characteristics. The theory is that the nonwoven geotextile will provide both reinforcement and drainage in the embankment. Drainage is desirable to dissipate pore pressure during constmction and to provide for a long-term drained condition, as well as to reorient seepage forces to have a net stabilizing influence on the slope. In cases where nonwoven fabrics lack needed tensile strength, composite woven and nonwoven geotextiles have been used.

Expanded fiberglass strands mixed pneumatically with backfill soil have been used both as primary internal reinforcement for the entire reinforced fill and for reinforcement of the zone at the fill face. This method, originally developed in Europe under the name "Texsol," has been used by the Forest Service in field fill face applications, and laboratory testing of its soil-interaction properties has been conducted by the California Department of Transportation Materials Laboratory. To date the design methodology is evolving, but the concept appears promising.

6F.4.2 Standard Recommended backfill gradation requirements for reinforced slopes are: and Marginal Backfill Material Sieve Size Percent Passing

4-inch 75-100 No. 4 20-100 No. 40 0-60 No. 200 &50

These gradation requirements are broad and can include excellent to marginal quality material. "Marginal" soils are defined as fine grained, low plasticity materials which may be difficult to compact, have poor drainage, or have strength parameters sensitive to density. Plasticity index should not exceed 20. Soil pH should be between 3 and 9 to avoid excessive corrosion.

On-site or local materials, often of marginal quality, are consistently used by the Forest Service for backfill in retaining walls and fills. Ideal coarse granular free- draining backfill material often is not available at construction sites or is very expensive to import. The coarse rockfill material, which is occasionally available, often has enough oversize material to make layer placement difficult, and rocks can damage the reinforcement material.

Local materials used have varied from silty sands to silts and clays (SM, SC, ML, and CL Unified Soil Classifications) with over 50 percent fines (passing the No. 200 sieve). Marginal materials should be tested to determine their strength properties and strength-density relationships. Many soils found in a montane environment have good frictional properties, usually exceeding a 30° effective friction angle (0) at the specified compaction density, and so are desirable backfill materials. Note that peak shear strength parameters should be used for analysis. Strength parameters should be determined using direct shear or consolidated-drained triaxial tests.

Use of marginal backfill has been acceptable, but its use can present problems in construction and long-term performance. Compaction of fine-grained soils is sensitive to moisture content, so close construction control is needed to ensure that specified densities are achieved. Typically, a specified density equal to 95 percent of the AASHTO T-99 maximum density has produced satisfactory results. Unless settlement is severe it normally does not present a structural problem for a reinforced fill. Cohesive soils have not yet been used in Forest Service applications. Creep characteristics would have to be carefully evaluated prior to use of a very cohesive backfill material.

6F.4.3 Fill Face With reinforcement, the final fill face most commonly designed and achieved is a Slope and Facing 1H:lV slope. However, this slope can vary between 1-114H:lV and 112H:lV. Needs depending on the soil type used and extra measures taken. A I H: I V slope is about

the steepest slope face achieved using a dominantly granular, low plasticity backfill material typical of mountainous terrain. Because the outer edge of the fill face is unsupported, good compaction in this area is very difficult to achieve. Without adequate density, soils placed on a steeper slope will typically not hold, and local fill face instability will occur. Attempts to construct an unsupported 112H:lV fill face have failed, even using a slightly clayey soil, and a 1H:I V slope was the result.

Use of a material with significant clay content may allow a somewhat steeper slope to be constructed, but performance probably is still controlled by construction limitations at the fill face. In very granular, non-plastic material, such as decomposed granitic soil, experience has shown that a 1-114H:lV face is the most appropriate stable slope. A steeper slope either will ravel or will need some additional type of support, such as wrapping the reinforcement material around the face. Wrapping material has included either the geogrid being used for reinforcement or a heavy erosion control matting or netting. Tensar Corporation has constructed 314H: 1 V slopes using the primary reinforcement grid wrapped around the face in up to 3-foot lifts. However, if much wrapping is needed, a retaining structure may become more economical. Eight-inch (20 cm) quarry rock has been used on fill faces to achieve a stable 1H:I V slope. Also, a variety of prefabricated concrete blocks is available for facing material.

Some promise exists for achieving a 1/2H:1 V slope face on low fills with the use of biotechnical measures such as vegetal stabilization. Experience with fills up to 11 feet (3.3 m) high has shown that a 112H:lV slope can be constructed, without forms, by placing a mixture of straw, clay-rich soil, manure, and seed in 1-foot (0.3-m) lifts along the outer couple feet of the fill face between geogrid reinforcing layers. The straw provides both tensile strength to support the steep slope and erosion protection until the seed germinates (which further adds root support). These measures can become labor intensive, and cost currently has been $9.50 per square foot. An aesthetic, well-stabilized face has been produced. Also, using layers of brush as secondary reinforcement on any slope face will likely improve the surface stabilization and provide erosion control for the face.

A 314H: 1 V slope face has been achieved by pneumatically shooting expanded fiberglass strands into the outer couple feet of the fill face as fill material was being dumped with a backhoe. This method appears promising and inexpensive. Only about 0.1 percent by weight of fiberglass reinforcement was added to the soil, and cost was about $3.50 per square foot, including reinforcing geogrid within the fill.

On any newly constructed fill face, particularly when it is oversteepened, both surface water control and erosion protection are needed. Surface water should be collected above the slope and channeled around it. At a minimum, some type of erosion control blanket or matting should be placed on the slope. Erosion control and revegetation measures to protect the fill face should be an integral part of the reinforced slope design, and should not be left to the discretion of the contractor.

6F.4.4 Drainage Most reinforced fills have had drainage provisions added, either to remove local Requirements ground water or to ensure that the backfill remains in a drained condition (as

assumed for the design). Typically, chimney drains or geocomposite drains are installed behind the fill. Alternatively, a layer of drainage material could be incorporated into the fill.

Geocomposite drains have been used extensively by the Forest Service. They are particularly applicable behind fills and retaining structures where the excavation backslope is steep or nearly vertical, making conventional gravel drains very difficult to construct. Numerous manufacturers produce geocomposite drains, and a variety of models has been used. A good study of their long-term performance has not been done, but experience to date has been satisfactory. Cost of the drains themselves is around $2 to $4 per square foot, installed.

Where considerable ground water flow is encountered and gradients are low, such as in the flatter toe-area of a slide, the capacity of some commonly available drains may be exceeded. Conventional gravel drains enveloped in geotextile, or a double layer of geocomposite drain, may be best suited for this area. A combination of gravel drains and geocomposite drains has also been used on one slide repair project to achieve adequate drainage in a cost-effective manner.

6F.4.5 Typlcal Construction Sequence

In specifying geocomposite drains for wall or fill applications, the items that appear most critical are crushing strength, flow capacity, and use of a geotextile which will satisfy needed filter criteria. Testing by the Forest Service and others shows that many geocomposite drains available today have both good crushing strength propexties and high flow rates. However, some do not, and products vary considerably. Frequently the Forest Service specifies a drain that can maintain a flow of at least 1 gallon per minute per linear foot of drain under a hydraulic gradient of 1. Crushing strength must exceed 1.5 tons per square foot. Either woven or nonwoven geotextiles are specified, depending on local soils. The geotextile should be tight or glued to the fin material for best performance.

Construction of reinforced fills is relatively simple and rapid once the design is well laid out. A typical construction sequence is:

First, the area is excavated to grade, and the subgrade is smoothed and compacted. To ensure that a suitable foundation exists, the excavation should be inspected by a geologist or geotechnical engineer.

Typically a horizontal layer of the reinforcing material is placed on the subgrade to the dimensions and orientation shown on the drawings or as directed by the engineer. Correct orientation should be confirmed by the manufacturer to ensure that the direction of maximum tensile strength is toward the embankment face. The material may be secured with staples, pins, stakes. or backfill material.

Drainage, typically a chimney drain using gravel or geocomposite material, is placed against the back of the excavation and brought up as necessary to stay above the lifts of backfill material. Perforated collection pipes and nonperforated drain pipes are installed as required on the drawings.

Layers of backfill material are placed and compacted similarly to normal earthwork lift operations. Backfill should be spread and compacted in such a manner that the reinforcing fabric or geogrid does not move and equipment does not operate directly on the material. Compaction on clay-rich soils should be in 6- to 8-inch lifts; on granular soils, 9- to 12-inch lifts. Lift thickness may be governed by reinforcement spacing. Soil is normally compacted to 95 percent of the AASHTO T-99 maximum density. Large, smooth drum vibratory or rubber tire rollers should be used. Sheepsfoot rollers may damage the reinforcement.

Facing elements or erosion control measures, or both, are either brought up or wrapped around as the lift sequence progresses, or facing is added upon completion of the lifts in the structure.

The sequence, spacing, and length of reinforcement layers and fill material lifts is continued as shown on the drawings until the structure is completed. Secondary reinforcement strips are commonly placed along the edge of the fill face as each lift or two are placed.

Correct orientation of the reinforcement is critical to the design of the structure, so it should be double-checked and confirmed. Also, because it is almost impossible to work with a single reinforcement roll width, either overlapping or splicing must be considered. Splices should be avoided along the reinforcement in the direction of the principal stresses (towards the face). The type of splice or overlap distance will depend upon the type of reinforcing material and direction of the splice. They should be installed in accordance with the manufacturer's directions.

6F.5 Relative Because many contractors are not familiar with this design concept, construction

Costs costs have not been as low as expected. Bid prices in 1987-92 have averaged $1&$15 per cubic yard for controlled compaction of the material, plus $4-$8 per square yard for the reinforcing geogrid, installed, including both the primary and intermediate reinforcement members. These same costs have ranged between $3.50 and $12 per square foot of vertical fill face. The lower cost can be realized on wide sites where equipment can operate efficiently, while small sites, common to many Forest Service applications, will likely continue to see relatively high unit costs.

6F.6 Sample Two design examples are examined. The Willow Slide on the Plumas National

Problems Forest in California is a storm damage site where a reinforced fill was designed and constructed. The external stability of the design and the gross internal stability have been examined using XSTABL.

The second design example, also analyzed using XSTABL, is the buttress sample problem of section 6E. The reinforced fill can be used as an alternative design solution to the buttress, but in this case it probably would not be as cost effective for the Elk River site as for the steeper Willow Slide site.

6F.6.1 Willow Willow Slide is a reinforced fill site that originally failed during the major storm Slide Reinforced event of February 1986. It is located in moderately steep terrain on Road 22N85Y in Fill the Plumas National Forest on the La Porte Ranger District. The area that failed was

the outer 15 feet of roadway and shoulder material for a distance of 75 feet along the road. Maximum depth of the debris slide, on a near-1:1 natural and fill slope, was approximately 8 feet, with a length downslope of over 80 feet (the plan view is shown in figure 6F.2). Over 1,000 cubic yards of slide material moved downslope in a narrow drainage chute for several hundred feet and entered Willow Creek.

CONSTRUCTION DETAIL

Figure 6F.2.-Plan view of the Willow Slide site. m N 4

Because of the relatively small size of the site and because hundreds of storm damage sites had to be repaired, no subsurface investigation or detailed soils testing was done at this site. Field mapping and examination of the site was done, as was a field classification of the material that failed. The failure occurred in the mantle of weathered soil and fill over an irregular weathered bedrock surface.

A reinforced fill design was chosen for the repair of this site because additional road width was needed, and a 50-foot-high 1:l fill could be rebuilt on the slope and could catch on a bench area near the toe of the slide feature. The initial reinforced fill design was done using design charts and the procedure given by Tensar Corporation (1986bsee figures 6F.3 and 6F.4a through 6F.4~-and was supplemented by the method outlined by Christopher (see figure 6F.5). Because backfill material for the repairs was to come from two local roadcut areas with fine to coarse nonplastic soils with some rock, design strength parameters were conservatively chosen with $I = 30'. C = 0, and moist density = 120 pcf.

Chart Procedure

Limiting Assumptions:

Extensible reinforcement. Slopes constructed with uniform, cohesionless soil (c-0). No pore pressures within the slope. Competent, level foundation soils. No seismic forces. Uniform surcharge no greater than 0.2m. Relatively high soil/reinforcement interface friction angle + - 0 . + (may not be appropriate for some geotextiles).

Determine force coefficient R from Fig. A above where +', - tan- (tan +r/FS, ) .

Determine Tm. 2 - 0.5Kvr H'

where ti' - H + q/yr q is a uniform surcharge.

Determine length of reinforcement L, and L, required from chart 8.

Figure 6F.3.-Chart procedure for confirming reinforced slope design (reprinted with permission of Tensar Corporation from Christopher et al., 1990).

30 40 50 60 70 BC a) Slope angle B

!O 4 0 50 65 7C 5 0

b) slope angle B

Figure 6F.4a.- Simplified design chart (reprinted from Mitchell and Villet, 1987). (a) with pore pressure coefficient of 0. (b) with pore pressure coejj7cient of 0.25.

The ciample ir takcn from the technical guidchncr for lhc dcrign and conrfivclm of rlrcp rctnlorccd crnbnnlmcntr o w rrablc foundations irsucd by Srrlon Lmircd j1981bl

T h e problem a to dclcrminr n rumble rcinfmcrment ls).oul to probide q ~ i l i b n u m in a M-8 high rmbanlmmt a l r h a dope angle 70 dcg built from compacted granular soil.

I The cmbanlmenr dimcnrionr are rhoun in F w r c 8-86 and thcrc is no surcharge loading.

1 Thr large w a r n uduc of shear rtrcngth i3 takcn to be c' = 0, m. = 19: and the msximum d m s w y = I21 pcf. The rlopc ir fully drrtned and r. = 0.

3 T h e c m h p insure c a f k i c n t from Ftgvrc B-83 i> K = 025. Thc rctnforccmenl length to rmbanlvncnt height r a m L / H = 0 6 9 from Figurc 8-83, giving a rcinforccmcn! length L = 11 5 fl.

4 The in-service rharsclctirlic rtrmgzh ruggcrtrd by the manufacturer of Tcnrar SR2 m grmular roils is the i d 4 lab- orator) value (1.W lb/fI) dirlded by a p s n d f a c t o r 1.1 lo 1.4 d c ~ n d e n t on rod rypc (re* Sw. 44 .3) to account lor parrlble conslrucl~on damage. creep. and lang.trm lor5 of rtrcnglh. ln rhm c e falung a panial factor 1.3 gira an in-rervicc d c r i p

2.m strength. - = I.JlO lb/fr.

1.3

..,, 6. An assumed minimum spacing for the r c i d o m m ~ n f S...,.

p - = 9 in g i v e a \ d u e for thc rparmg conrflnl Q = -- - KyS..,.

50 6 ft. 1. The dcprhr and tbiclneir of thr zonn of equal reinforce-

mrnt ~ p l c i n g m y bc ralculalrd (see Fig 8.82) and thnr nrr s h o r n in Figurc 8-87,

8. The tom1 hotimntnl form tcquircd lo provide equilibrium is T = !4 KrH = 5,860 1b/fl.

9 , The cslculnted number of rcinfoilrcrmcnl grids is 8 (Table 5 860 Ib/*

8.7. Fig. 8-88), giving thr chcck = 731 I b / h < 8

1.130 Ib/ft, which is sufiicirnt.

Figure 6F.4b .S impl i j i ed design chan with pore pressure coeflcient of 0.50, and a design chart example lfrom Mitchell and Villet, 1987).

Spailng of Grids Depth t o B o t t o m Thickness o f i n Zone i f 1 1 of Zone (111 Zone i f t l

P r a c t i c a l spacings Depths of s p a c i n g zones of grids i f t ) i n i f t ) i n m u l t i p l e s of p r n u l t i ~ l e r o f Smin a

Figure 6F.4c.Simplified design chart (reprinted from Mitchell and Villet, 1987).

The design resulted in Tensar SR-2 geogrid for the primary reinforcement, with vertical spacings of 3 or 5 feet. SS-I biaxial geogrid was used for secondary reinforcement at I-foot spacing along the fill face. Concern for block sliding of the reinforced mass (and subsequent block stability analysis) resulted in the addition of other benches along the fill-bedrock contact. Figures 6F.6 and 6F.7 show the final reinforced fill cross-section constructed with a multiple-benched interface and drainage with geocomposite drains. Note that a soil-geocomposite friction angle of 2/3 $I was used, resulting in an FOS of 1.2 with the benches.

CENTER OF ROTATION

I - Le EMBEDMENT LENGTH

L

Factor of safety of unreinforced slope:

L, m Resisting Moment (M,) - ', T~ . R dL F.S.,, - (5'3) Driving Moment (M,) (Wx + Aq - d)

where: w - weight of sliding earth mass L,, - length of slip plane d q - surcharge r, - shear strength of soil

Factor of safety of reinforced slope:

T, - D F.S. - F.S." + - (59)

", where: T, - sum of available tensile force per width

of reinforcement for all reinforcement layers

D - moment arm of T, about the center of rotation - R for extensible reinforcement - Y for inextensible reinforcement

Rotational shear approach to required strength of reinforcements.

Figure 6F.5.-Rotational shear approach to required strength of reinforcements (from Christopher er al., 1990).

Figure 6F.6.-Reinforcedfill cross-section with drainage and a benchedfill-native soil interface to prevent sliding failure.

Stability analysis using XSTABL showed an unreinforced 1:l slope FOS of 0.58 (figure 6F.8). Application of reinforcement in the 1 : 1 slope, with the benched interface as constructed, produced an FOS of 1.30 (figure 6F.9).

SECTION 6-8

SIATIW 11.20

Figure 6F.7.-Construction detail for the Willow Slide site. 0

%

willow s l ide , 1 : 1 unre in forced 3 most critical surfaces, MINIMUM BISHOP FOS = .581

Figure 6F.R-willow Slide, I : / unreinforced; three nwsr criricul surjhces, minimum BlSffOP FOS = 0 . ~ 8 1

willow slide, 1: 1 reinforced force Maximum Rernr. Force (BISHOP, FOS = 1.30) - 11794. Ib.

Figure 6F.9.-Willow Creek XSTABL run for reinforced slope with the benched interface as constructed.

6F.6.2 Elk River Refer to the Elk River buttress example in section 6E.5 and figures 6F. 10 and 6F.I I . Buttress vs. Reinforced Fill Figure 6F.10 shows an XSTABL run for unreinforced fill, with a Bishop analysis

reporting a minimum factor of safety of 0.502.

Figure 6F.11 shows an XSTABL run for the same location with a reinforced fill with a Bishop analysis reporting an FOS of 1.185.

elk river, reinforced fill, unreinf 3 most critical surfaces. YINIYUM BISHOP FOS = 502

5

0 20 40 60 80 100 120 140 1 0 X-AXIS (feet)

Figure 6F.10.-Elk River, unreinforcedfill; fhree most critical surfaces, minimum BISHOP FOS = 0.502

elk river, reinforced fill, x=30 f t 3 most critical surfaces, YlNlYUY BISHOP FOS = 1.185

, 20 4U 0 80 IW 120 140 1 0

X-AXIS (feet)

Figure 6F.11.-Elk River, reinforcedfill, x=30ff.; three most critical surfaces, minimum BISHOP FOS = 1.185.

6G. Shear Trenches

Cliff Denning, Geotechnical Engineer, Mt. Hood National Forest

6G.1 General Shear keys are commonly used with retaining walls to resist sliding failure. A shear trench, as described in this section, is similar to a shear key and consists of a trench excavated through a potential or existing slope failure surface. Material that has failed or may fail is then replaced with stronger material. Shear trenches can be used alone or as a contributing portion of a combination of actions, such as with a counterbalance weight (buttress).

The purpose of using a shear trench is to increase the resisting force along a failure surface. A shear trench will accomplish this in two ways. The first is by replacing the material comprising the failure surface with stronger material, usually compacted granular material possessing significantly higher shear strength (angle of internal friction). The depth and width of the trench are determined by stability analysis. The second is by acting as a ground water intercept trench, lowering perched ground water, thereby reducing pore pressures acting on the failure surface. Methods of determining ground water drawdown are discussed further in section 4E.2.

6G.2 Pre- The shear trench is used for shallow, planar-type failures which would be analyzed

Con~truction using a sliding block or a noncircular procedure as discussed in section 5. As with any slope stability analysis, developing an accurate model to analyze is important. The need for an adequate site investigation to determine such things as the location of soil and rock units, soil and rock strength parameters, ground water location, and the location of potential or existing failure surfaces cannot be overemphasized. The reader is referred to section 3 for guidance in evaluating site conditions and developing a model.

Items to consider when thinking about using a shear trench for slide stabilization include:

* The location of a flatter portion of the slide where construction equipment may gain access to the rest of the slide.

The existence of a thin portion of the slide where excavation can be minimized. Benching the excavation could allow the depth to be increased. The Oregon Department of Transportation (ODOT) has found shear trench excavations approaching 40 feet in depth become very massive and difficult to construct (ODOT, 1991); therefore, they limit shear trench depth to 40 feet as a rule of thumb.

Filling or partially filling in the excavation to satisfy stage construction requirements.

Often, shear strength along the failure surface and pore pressures acting on the failure surface are the variables most difficult to define. Movement often lowers shear strength to residual values. Repeated (reversed) direct shear tests can be used to estimate residual friction angles. Charts relating simple index tests, such as plasticity index (PI), to residual friction angle are available to estimate residual shear strength. Additional discussion of residual shear strength, tests, and charts can be found in section 4C.3.4. Estimated ground water level can be used with the measured residual shear strength to calculate a factor of safety (FOS) along the failure surface. The ground water table can be raised or lowered until the FOS is approximately 1. If the investigation has provided some confidence on the ground water conditions at the site, the residual friction angle can be back-calculated by solving for the residual friction angle for an FOS of 0.95 to 1.00. In this back-analysis, the portion of the residual shear strength contributed by cohesion is assumed to be very small or zero (section 4C.3.4). Figure 6G.1 shows a slope with the failure surface location inferred by a site investigation. Figure 6G.2 shows the FOS from several XSTABL runs with friction angle plotted against FOS. An estimate of the residual friction angle can be made by setting the FOS to 1 and solving for the friction angle as shown in figure 6G.2.

Figure 6G.I.-A slope with the failure surface location inferred by a site investigation (after ODOT, 1985)

Figure 6G.2.-The FOS from several XSTABL runs with friction angle plotted against FOS. An estimate of the residual friction angle can be made by setting the FOS to one and solving for the friction angle.

In practice, many or all of these methods are used and the results are compared with each other. The designerlinvestigator uses experience with similar conditions or sites and familiarity with the materials, along with results from the analysis, to select a model for remedial analysis.

There are several stability analyses for shear trenches presented in this section. It is not the intent to develop a "cookbook" to cover all situations, but rather to present a general guide for the reader's consideration.

66.3 Shear Figure 6G.3 shows the slope presented in figure 6G.1 with a shear trench located

Trench Width near the toe of the failure surface. Note that the ground water intercept curve has been plotted on the figure. Several stability analyses are run by varying the trench width and solving for the FOS with respect to sliding through the shear key. The shear trench width is adjusted until the FOS calculated equals the FOS required. Figure 6G.4 shows the results graphically.

Figure 6G.3.-The slope presented in figure 6G.1 with a shear trench located near the toe of the failure surface. The ground water intercept curve has been plotted on thefigure (after ODOT, 1985).

Figure 6G.4.-Results of several stability analyses run by varying the trench width and solving for the FOS with respect to sliding through the shear key. The shear trench width is adjusted until the calculated FOS equals the FOS required.

G.4 Shear Figure 6G.5 shows the slope with the failure surface passing below the shear trench.

~~~~~h ~ ~ ~ t h Figure 6G.6 is a closer view of the shear trench area. The shear trench depth is adjusted until the FOS calculated equals the FOS required. In many instances, the trench will penetrate into a lower, stronger unit (such as a rock unit) requiring only a minimum depth below the failure surface. ODOT (1991) recommends at least a 5-foot depth below the failure surface.

Figure 6G.5.-The slope with the failure surface passing below the shear trench (after ODOT, 1985).

Figure 6G.6.-A closer view of the shear trench area.

66.5 Analysis Figure 6G.7 shows a closer view of the shear trench area with the passive wedge of with Passive the block failure surface exiting up through the shear trench. The location of the Wedge Exiting passive block within the trench can be varied as shown in this figure.

Through Shear Trench

-,~- ..-- L 1- - ,+o <m , & ,.. l ( o - ~ ~ r- -~*-

a r r y r r w + E CFIITI

?nsbl#~. ' ~ E O C E ELIWC TWP.YW SWIQ ~ ~ E * I C H

Figure 6G.7.-A closer view of the shear trench area with the passive wedge of the block failure su$ace exiting up through the shear trench. The location of the passive block within the trench can be varied (afer ODOT, 1985).

6G.6 Analysis A change in strength along the Failure plane may initiate a Failure behind the shear

with Passive trench. Figure 6G.8 shows a passive wedge toeing out behind the shear trench.

Wedge Exiting Experience has shown that failures of this type are possible, especially when the depth to failure behind the trench is 15 feet or less. The location of the passive

Behind Shear block can be varied as shown in this figure. Trench

Figure 6G.8.-A passive wedge toeing out behind the shear trench. Experience has shown that failures of this type are possible, especially when the depth to failure behind the trench is 15 feet or less. The location of the passive block can be varied (after ODOT, 1985).

66.7 Analysis Figure 6G.9 shows an active wedge of a block failure surface located within the with Active shear trench. The active wedge can be varied as shown to check for toe kickout Wedge Exiting below the shear trench.

Shear Trench

Figure 6G.9.-An active wedge of a block failure surface located within the shear trench. The active wedge can be varied to check for toe kickout below the shear trench (ajier ODOT, 1985).

6G.8 Figure 6G.10 shows the slope with a counterbalance fill. In this case the fill Counterbalance increased the FOS of the slope and provided a waste area for material from the shear Fill trench excavation, eliminating haul costs. Analyses similar to those discussed above

must be run on the counterbalance till and shear trench combination.

I SWRR T~EIICU CI~TH COWTSKBQLWEE FIL

Figure 6G. I0.-The slope with a counterbalance fill (ajier ODOT, 1985).

6G.9 An analysis of an open shear trench would show a very low FOS. For this reason,

Con~truction shear trench construction should be done only during the dry summer months. Excavation done in other months, when the ground water levels are higher, would have a greater risk of initiating slide movements. In addition, the shear trench should be constructed in stages. ODOT (1991) generally recommends that the length of an unfilled shear trench section not exceed 20 feet along the foundation at any given time (this recommendation has varied depending on the geotechnical engineer's judgment based on open excavation movement observations during construction). A recommended sequence for the shear trench excavation and backfill is shown in figure 6G.11. The excavation should be backfilled with shear trench material by the end of each working day.

STEP 2 - lnltml Backlill and Dram

STEP 3 -Excavate Second Smge

STEP 4 - S1.p. 2 0.eLllll and Drain Inslallalion

Figure 6G.11.-A recommended sequence for the shear trench (reprinted from ODT, 1985).

Excavation for the shear trench should be observed by a geologist or geotechnical engineer to confirm that the base of the shear trench is the required depth below the actual failure surface and into competent material based upon the analysis.

In general, shear trench material should consist of clean, hard, durable, free-draining, angular gravel or rock, reasonably well-graded from coarse to fine (ODOT, 1988). The shear trench rock should be placed and manipulated to provide a dense and well-filled mass. Geotextile may be required along the shear trench if the trench is also functioning as a ground water cutoff trench. The reader is directed to section 6F and Christopher and Holtz (1984) for further discussion on geotextile design. Collected ground water within the shear trench should be directed away from the slide area. This often can be done by placing a french drain or a perforated pipe along the toe of the shear trench cut slope and carrying the collected ground water off-site in a nonperforated pipe.

6G.10 Post- Signs of movement above the shear trench (such as surface cracking) should be

C ~ n ~ t r ~ ~ t i ~ n expected for 2 to 4 years after construction. Minor movement is required to develop shear strength within the shear trench. During slide movement prior to shear trench construction, voids often form in the active wedge. Rearrangement of soil is to be expected, because these voids may surface after construction. Post-construction maintenance should fill in any cracks to reduce voids and water infiltration.

6G.11 Sample A sample problem using shear trenches is presented in appendix 6.2.

Problem

6H. Rock Slope Stabilization

Michael T. Long, Engineering Geologist, Willamette National Forest

6H.1 General This section discusses alternatives, advantages, and disadvantages for stabilizing mass

Introduction wasting processes in natural and constructed rock slope environments. Stabilization measures will depend heavily on the information gathered from the procedures in sections 3C.2, 4D, 4E, SH, and 6D.

6H.2 In a general sense, the advantages of stabilizing a failing rock slope or one which has

Advantages and a high potential for failure may be obvious: it reduces risk of injury to the public,

Disadvantages reduces maintenance costs, and reduces the possibility of road closure. In more specific cases, stabilizing or rehabilitating aging rock slopes may enhance fisheries' environments by reducing sediment yield to streams (shotcrete applications and scaling), may reduce the agency's liability in the event of a tort claim from a rockfall injury, and may prevent severe environmental damage by preventing large-scale structural discontinuities from daylighting in the slope due to small-scale failures.

There are some disadvantages that must be considered in choosing a design alternative. Rock bolt installation often involves a drill rig suspended from a large crane, which may block traffic. Scaling operations involve mad closure as well due to loose rock falling on the roadway. Artificial support and containment systems, such as screening, bolting, and shotcrete applications, although well engineered, are obviously not a natural component of the slope and can be visually disadvantageous.

6H.3 Necessary The most crucial information for stabilization design is also that needed in the

Design stability analysis. Sizes, lengths, angles, and loads on bolts; square feet of retaining

Information screen; cubic yards of shotcrete; and crew scaling hours all depend on an accurate representation of the slope geometry, defined by design segments (see figure 6H.1). These design segments are defined during the mapping phase of the project by horizontal and vertical determination of the mass structural geology characteristics, local discontinuities, and individual structural blocks. Outcrop mapping upslope, and discontinuity mapping by horizontaVvertical line and cell methods, must be used in order to define the geometry.

Canyon Creek Bluf Fs Construction Zones / \

AFFGUXIMATE Y N E

Ez a 1 a . m

CCNSTRKTlCN Z C N 3 Scalc sod chute bladmp ad mchsnhcml scol rq na,l bc used a. -04 & L k F r e d Ssvrcc Const r~c t slope reinfarcrn~ mat ss shown on Shcct 15 hsta l l so11 -hers ar shorn on Y c c t 16

CaVSTRLTTICN Z C N 4 Scalc. ~ns to l l 35 TF I rock bolts 15 F t

CWTRUCTICN Z M 6 Scdc mstoll approxt-tcly 4 T p I rack bolts I2 ft Imp as d rcc t cd & the -lncr Plocc free hap1-p slop= contat-t mat os shorn m Skcts 11 B 14

cmmnm ZM 8 Scdc mstall 22 T p 2 rockbolts 23 ft 1- cn 10 l vc r t l bg B l h r m l pat tern as sho.n a, Ynt 7 Plocs Frcc howlno slope contot-nt mat aa she- on S k e t s 11 8 14

Figure 6H.I.-An example of construction zones and design segments defined during mapping.

6H.4 Stabilization Components and Systems

6H.4.1 Reduce Driving Forces

6H.4.2 Increase Resisting Forces

Parametric analysis of the strength characteristics of the discontinuities is equally important, and these strength characteristics are usually determined during the mapping phase (see section 4D for typical values). Construction zones are developed from the design segments based on similar activities. A rock bolt design segment may appear in several locations on the slope but may only be zoned a unique construction segment number for project scheduling and payment (see figure 6H.1). Project economics, weighed against risk, will also be a factor in choosing stability systems and combinations of systems.

Rock slope stabilization systems and components fall into three categories: those which reduce driving forces; those which increase resisting forces; and those that redirect, contain, or dissipate rockfall. Often, a combination of systems is used on a single project. The following is a discussion of each category. Sample specifications are found in appendix 6.4.

Drainage

Hydrostatic forces created by water-filled tension cracks can cause rock slope failure. Slopes should be evaluated for the presence of tension cracks and for water sources upslope. Simple upslope interceptor and diversion trenches can often improve slope drainage. Care must be taken to consider long-term maintenance of open trenches. Cut-and-cover designs using clean open-graded fabric-wrapped gravel will reduce long-term maintenance problems.

Geometry

By changing the geometry (steepness and alignment) of the construction slope, the resulting intersections of planar, wedge, and toppling discontinuities may be designed to avoid critical dip angles. Markland's test for kinematic possibility of failure is useful for determining critical slope angles and directions but does not account for apparent cohesion or pore pressures (see section 3C.2 for a structural analysis discussion).

Changing geometry may require blasting followed by mechanical scaling of the slope to remove loose rock fragments. Controlled blasting methods using pre-split designs should be used to prevent over-break of the final dope face. Hand scaling of the slope should be completed after blasting in order to remove all of the loose rock and vegetation and prevent root-wedging. Any trees growing within 10 to 15 feet of the slope crest should be removed to prevent wind-levering of rocks at the slope crest.

Drainage

Excess pore-water pressures are often the cause of rock slope failure. Reducing the excess pore pressure results in an increase in normal force and stability. Slopes should be evaluated for the presence of seepage on the face through the discontinuities. Horizontal drains are also effective in reducing pore pressures and improving overall drainage. If horizontal drains are anticipated with the installation of rock bolts or dowels, it is advantageous to locate specific sites during the drilling phase of rock bolt installation (see section 6D for drainage discussion).

Tensioned Rock Bolts

As discussed in section 5H, tensioned rock bolts improve stability by applying an additional normal and upslope force component on the failure plane by transfer of load through tension on the bolt and compression on the face bearing plate analogous to a pair of pliers pinching a deck of playing cards. Two popular types of tensioned bolts are:

Those with mechanical expansion shell anchors located at the base of the bolt which anchor it in place at the bottom of the drill hole and allow for tensioning. Corrosion protection is accomplished by pumping cement grout through a small-diameter center hole in hollow-core bolts (see figure 6H.2).

' (, EXPANslON SHELL

GROUT RE TURN TUBE

GROUT HOLE ----

/'

WASHER ANO NLlT -' /

6 5 3 / 8 STEEL REARING PLATE

Figure 6H.2.-A tensioned rock bolt with a mechanical expansion shell anchor.

Those which use a resin cartridge system with quick- and slow-setting resins. The quick-setting resin cartridges are inserted in the hole first, followed by the slow-setting resin cartridges. As the bolt is spun into the hole, the cartridges break and encapsulate the bolt. The quick-setting resin hardens in less than 3 minutes, allowing the bolt to be tensioned and locked into place as the slow- setting resin cures.

Load testing of the first three installations and at least 10 percent of the remaining bolts is suggested in order to maintain quality control and jack calibration. Bolts should extend 4 to 5 feet beyond the anticipated failure surface.

Untensioned Rock Bolts

Untensioned or post-tensioned bolts may be installed like tensioned bolts, but without a load being applied to the bolt. This type of installation is more for containment purposes than for reinforcement (see figure 6H.3). Due to the lack of tensioning and equipment setup, unit cost is lower.

EXISTING FACE

v 4 THREAD LENGTH =

WASHER AND NUT

HOLE DIMENSIONS 4 5 SPEClFlED 8, GROUT MANUFACTURER A N n APPRnVFn BY FOREST SERVICE

POLYESTER RE5114 EP3XY OR OTHER PRE APPROVE6 GROUT

8- X 8' X 1/4 1 n4 REBAR HOT DIP GALVANIZEO GALVANIZEL 5TEEL1 )\ k WITH THREADEC ENE PLATE

Figure 6H.3.-An untensioned rock bolt, used primarily for containment.

Shear Dowels

Untensioned rock bolts may be installed perpendicular to a potential failure plane in front of individual blocks with sliding potential (see figure 6H.4). Rules of thumb for 1-112- to 2-inch-diameter dowels are: 12-inch stickup; each dowel will hold 40 to 60 tons; a crew can install 100 per day; and do not use on slopes greater than 60".

bdlvldwl dowels and cuncrete ~ o p s

Note: Concrete shall esxtend a mlnlmum o f 12' laterally on each slde o f the dowel.

Potmt/al failure plane

P l m dowel as clam to rwk fm as passlbIe

Sectlon 602 concreh fill vvdds

Figure 6H.4.-lnstallarion derails for shear dowels.

Buttress

In some situations where removal of failed material could trigger further upslope banslation of failure, a toe buttress should be considered. If it is in a road template, it may be designed as a vertical curve through the section to be stabilized (see section 6E).

6H.4.3 Rockfall Shotcrete Mitigation

Portland cement concrete applied to the slope by pneumatically blowing it through a nozzle is termed shotcrete, or gunite, depending on whether the concrete is premixed or mixed with water as it is applied to the slope. The process requires operator experience and skill in order to prevent rebound (mix that bounces off of the slope due to either excessive pressure or the nozzle being too close to the slope). Quality control must be maintained by gauging thickness and sampling for unconfined compressive strength.

Shotcrete is applied to slopes with weaker interbedded material to prevent spalling and eventual development of overhangs and to slopes with a large number of open discontinuities that produce a "diced" effect that spalls easily. Steel, polypropylene, or fiberglass reinforcing fibers are mixed with the concrete to increase strength. Short sections of rebar are installed in the slope prior to application to act as hangars. Weep holes should be provided to prevent hydrostatic forces from building behind the shotcrete. Rules of thumb are: a maximum 12 to 15 percent rebound is allowed; the nozzle should be approximately 5 feet from the slope and remain horizontal; and the final thickness should be greater than 3-112 inches to prevent spalling of the concrete (see figure 6H.5).

Figure 6H.5.-Placement of shotcrete.

Wire Mesh Screening

Free-hanging or reinforced wire screening can be installed on rock slopes which have a potential for continual rockfall and ravel. Use of "chain link" type of fabric is not recommended due to the potential of complete vertical disassembly in the event of a tear. Double- or triple-twisted gabion wire is now widely used due to its flexibility, ease of handling, and durability. When a tear or puncture occurs, it does not expand further than the diameter of the hole (see figure 6H.6).

TYPE I ROCK DOWEL INSTALLATION

em- mV(Cm PI* ---

<. .m.-w M I W D t l r n -7- -

DETAIL OF TENSION WIRE AN0 MESH INSTALLATICN WIRE MESH C O M C T l M \ I DETAIL

Figure 6H.6.-Details of wire mesh screening.

Energy-Absorbing Barriers

Energy-absorbing barriers at mid-slope or slope toes may be installed in areas of critical rock fall with long upslope runout areas. CRSP (Colorado Rockfall Simulation Program ver. 4.0 1993) is useful in evaluating these types of slopes and in designing barriers. Some barriers consist of heavy wire rope netting, such as that used by defense agencies as submarine nets, supported by anchored steel supports. Others consist of simple free-standing bin-type walls filled with earth.

Fallout Areas

Fallout zones at the toe of the slope can be designed so that the depth to width ratio will be adequate to prevent a falling, rolling, bouncing rock from going beyond the catchment area. The nomograph shown in figure 6H.7 is derived from the criteria developed by Ritchie (1963). Depth may be increased by excavation or by installing gabion baskets or Jersey-type concrete barriers along the outside of the fallout area.

6H.5 Example Example problems are provided in section 5H and include geometry modification,

Problems drainage, and artificial support.

Figure 6H.7.-Fallout areas: (a) ditch design chart; (b) rockfalls on various slopes (from Colder Associates, 1989).

61. Comparison of Alternatives and Decision Analysis

RenL Renteria, Geotechnical Engineer, Intermountain Regional Ofice

61.1 Introduction

61.2 Methods for Equivalent Comparison of Alternatives

61.2.1 Defining Mutually Exclusive Alternatives

Section 61 discusses the process whereby the geotechnical specialist communicates information on mitigation alternatives, associated costs, risks of failure, and potential benefits to the decision maker. It is also within the scope of this analysis to present a recommended alternative and the basis for that recommendation.

A set of alternatives must be evaluated and compared before a preferred alternative can be recommended. These alternatives should be mutually exclusive, such that only one alternative will be chosen. In order for the comparison to be fair, the unit of comparison (typically monetary value) must have an equivalent basis. This is usually accomplished under the principle of discounted cash flows of the time value of money.

A convenient way to document, analyze, compare, and select a preferred alternative is to use a decision analysis. This can be as simple as eliminating non-feasible alternatives or as complex as solving a multiple-branched decision tree. If possible, probabilistic techniques should be used to better define the risks associated with an alternative. It is the task of the geotechnical specialist to provide the decision maker a clear picture of the possible outcomes for each alternative and which alternative provides the best combination of minimized risk and maximum benefit.

A standard must be chosen for an equitable comparison before a decision can be made to recommend an alternative. In engineering economics, alternatives are compared based on their dollar value. This dollar value is "discounted" over the life of the project to account for the time value of money (a dollar today is not necessarily worth a dollar a year from now). Most resource agencies must consider additional "values" for comparison, such as resource, social, and agency impacts. This section will provide a discussion on quantifiable values-mainly providing a least-cost, low-impact alternative.

One of the duties of a geotechnical specialist is to recommend a stabilization plan to the land manager (decision maker). The first step in creating alternatives is to develop proposals. The proposal stage is much like "brainstorming." These proposals provide the basis for an alternative, such as drainage, increase in resisting strength, or decrease in driving force. According to White et al. (1977),

The difference between a proposal and an alternative is that an alternative requires a decision. A mutually exclusive alternative requires that no more than one alternative can be chosen.

Problem 1: An analysis of a cut slope failure has generated the following feasible proposals:

No action (or minimum action-maintenance)

Horizontal drains

Rock buttress

From the proposals, the following alternatives are developed. The no-action alternative is listed because it is always an option.

(1) No action

(2) Horizontal drains

(3) Rock buttress

(4) Horizontal drains and a rock buttress

As can be seen, some proposals are mutually exclusive, such as the no-action proposal with any other proposal. More than one proposal (horizontal drains and a rock buttress) can be chosen to form an alternative. However, because alternatives provide the basis for a specific decision for the slope mitigation, the choice must be to do nothing (no action), use horizontal drains, install a rock buttress, or use both horizontal drains and a rock buttress. The alternatives are mutually exclusive because only one alternative will be chosen for the final stabilization measure.

61.2.2 Setting To make an objective comparison of mitigation alternatives, the comparison should Project Life for use a common period of time. This common period of time is called the planning Comparison horizon. Some common methods of defining the planning horizon are (White et al.,

1977):

Least common multiple of lives for the alternatives

Shortest design life among the alternatives

Longest design life among the alternatives

A standard design life

When alternatives of unequal life are compared, an estimate must be made for either the remaining value of the alternative (salvage value) or the cost to continue providing the same service.

61.2.3 Determining Equivalence Using the Time Value of Money

Problem 2: Define the planning horizon for the following two alternatives:

(A) Reinforced embankment, design life of 20 years. (B) Welded-wire retaining wall, design life of 50 years.

I shortest design life 1 20 yr 1 I life 1 1 life + SV 1

Method

least common multiple

The shortest design life method will require an estimate of salvage value (SV) from year 20 to year 50 for alternative B. The longest design life method requires an estimate of the cost to provide the same level of service from alternative A from year 40 (2 design lives) to year 50.

Cost A

5 lives

Planning Horizon

100 vr

longest design life I 50 yr

The initial investment and any monetary benefits or maintenance cash flows required over the planning horizon must be considered to make an economic comparison of alternatives. These cash flows and their timing will be quite different for each alternative; therefore, a method is needed to convert the cash flows to a common measure. This requires a consideration of the time value of money.

Cost B

2 lives

2 lives+lO yr 1 life

The time value of money provides a distinction between what a dollar is worth today versus what it will be worth 1 year from now. The process uses compounded interest calculations to convert future cash flows into present dollars or equivalent annual dollars for a given time period.

The three methods for time value of money presented here are present worth, annual worth, and future worth. All three of these. methods are equivalent for comparison and will produce the same recommendation in a cost-comparison decision analysis. The choice of which method to use should be based on how the decision maker prefers the economic value to be represented.

standard life of 20 yr 20 yr

The interest rate to use for comparison is typically referred to as the cost of capital, or the minimum rate of return. Questions have been raised as to the validity of Government projects using an interest rate. Standard practice for Forest Service projects has been to use a typical value of 4 percent.

Table 61.1 presents a summary of the discrete compounding formulas. The chart solutions for the formulas at a 4 percent interest rate are presented in table 61.2. Symbols are traditionally used to represent the various cash flows over an analysis period. The "years" of analysis is referred to as n. The interest rate for analysis is referred to as i. A cash flow at the beginning of an analysis is represented by P, for present value. Cash flows occurring at a future time are represented by F. A series of "annual" cash flows of the same value is represented by A. For a more extensive

1 life 1 life + SV

discussion of the time value of money, the reader is referred to textbooks on engineering economic analysis, such as that of White et al. (1977).

Table 61.1.-Discount factors for discrete compounding.

Present Worth (PW)

Annual Wonh (AW)

Future Wonh (FW)

Table 61.2.-Compound interest table for 4 percent interest rate.

1 Discrete Corn~oundine at i=4% Interest Rate "

n

1

2

3

(PlF) 0.9615

0.9246

0.8890

(AIF) 1.0000

0.4920

0.3203

(PIN 0.9615

1.8861

2.7751

(Alp) 1.0400

0.5302

0.3603

n

1

2

3

(FIP) 1.0400

1.0816

1.1249

VIA) 1.0000

2.0400

3.1216

Table 61.2.-Compound interest table for 4 percent interest rate (cont'd).

61.2.3.1. Present Worth

The present worth method converts all cash flows to a single value at a referenced beginning (value at year 0).

61.2.3.2. Annual Worth

The annual worth method converts all cash flows to an equivalent uniform annual series of cash flows for a common period of time (the planning horizon).

61.2.3.3. Future Worth

The future worth method converts all cash flows to a single value at a referenced future time (value at end of planning horizon).

Problem 3:

Two alternatives are being considered for an unstable embankment along a Forest Service single-lane asphalt concrete paved road with turnouts. A popular campground is located at the end of the road beyond the embankment. Each year the road grade drops through the unstable area, requiring $1,200 of maintenance patching to transition over the cracks. This "minimum action" alternative also requires an overlay to be placed every 5 years to bridge the sag created by the drop in grade to allow clearance for passenger cars and trailers (no maintenance patching is done when an overlay is constructed). The estimated cost of the overlay is $3,500.

Continued instability is expected unless the organic materials at the embankment/ground contact are removed. The cost of the removal of organics and the rebuilding of a single lane 1:l embankment is estimated to be $12,500.

The district requires dust-free vehicle passage until after Labor Day. Long-range plans are to widen the road to two lanes in 10 years.

Required:

Compare the two alternatives using present worth, annual worth, and future worth (round to the nearest whole dollar).

Solution:

Choose a 10-year project life based on the end of a single-lane road standard, which will require modification of the road template. No overlay will be placed in year 10, but early season maintenance will be required for vehicle passage before reconstruction will begin after the summer visitor season. Use the Government interest rate of 4 percent.

Alternative A. Minimum action of regular yearly maintenance through year 10 and an additional expense in year 5 for an overlay instead of patching.

FW = $1,200*(FIA,4%,10)+$F*(P/F,4%,5)*(NP,4%,10) = $17,206 also

Alternative B. Excavate and rebuild steepened embankment.

Alternative

1 I

On the basis of cost alone, alternative A is indicated.

A. Minimum Action

B. ExcavatelRebuild

Total Cost (9

i=O%

$14,300

$12,500

i=4%

F W

$17,206

$18,503

PW

$1 1,623

$12,500

A W

$1,433

$1,541

61.3 Comparison Under Conditions of Uncertainty and Risk

61.3.1 Break-Even Analysis

Usually in slope stability analysis, few of the analysis parameters are known with certainty. Some uncertainty often is associated with material properties, ground water conditions, major storm events, project life expectancy, project costs, and failure mechanism. This section will examine some methods to observe the effects an analysis parameter can have on a solution over the range of uncertainty.

Break-even and sensitivity analyses can be considered to be two-dimensional analyses because only one analysis parameter is allowed to vary at a time. Probabilistic risk analysis can be considered to be multi-dimensional because several analysis parameters can be varied simultaneously, as in using the process of probabilistic model simulation. The selection of an alternative under uncertainty and considering risk will be discussed in section 61.4.

Break-even analysis is used when an accurate estimate of an analysis parameter cannot be made, but a reasonable estimate can be made as to whether the value is above or below a break-even point. A definition is given by White et al. (1977):

When we are completely unceaain of the possible values a parameter can take on, we will be interested in determining the set of values for which an investment alternative is justified economically and the set of values for which an alternative is not justified; this process is called break-even analysis.

Problem 4:

An unstable embankment on a Forest Service road requires yearly maintenance of $1,000. The embankment has been identified for major reconstruction along with improvements to the road standard. The total cost of the reconstruction will be $9,000. At this time, full funding is not available. Subsurface water has been confirmed within the embankment. Horizontal drains are considered the most feasible temporary alternative for this site before major reconstruction occurs. Installation cost of the drains is estimated at $3,800.

Required:

Using the time value of money (neglect inflation), recommend a temporary repair if major reconstruction is expected to occur in (1) 3 years; (2) 4 years; (3) 5 years.

Solution:

Use the present worth method at a 4 percent interest rate. Assume no salvage value when reconstruction occurs (temporary repair destroyed as part of reconstruction).

Maintenance:

PW = $1,000 * (P/A,4%,n) (formula from table 61.1)

Horizontal Drains:

Find the "break-even" point by mathematically equating the two formulas and solving for n, or use the preferred method of graphical representation shown in figure 61.1.

Figure 61.1.-Break-even analysis of maintenance and horizontal drains

61.3.2 Sensitivity Sensitivity analysis is used to examine the effects of varying an analysis parameter. Analysis A definition is given by White et al. (1977):

When we are reasonably sure of the possible values a parameter can take on, but uncertain of their chances of occurrence, we will be interested in the sensitivity of the measure of merit to various parameter values; this process is referred to as sensitivity analysis.

Although it is similar to break-even analysis, sensitivity analysis is quite useful for examining the effects of errors in estimating the value of a given analysis parameter. It is often used for non-economic analysis.

Problem 5:

A mining company has submitted a proposal for a waste rock dump (embankment). The rock dump will be constructed using end-dump placement. The dimensions for the proposed dump are shown in figure 61.2. The stability analysis was conducted using Janbu's simplified method. The friction angle was assumed to be the 35" angle-of-repose for the rock material. The seismic coefficient was assumed to be 0.30 using Army Corps of Engineers zonation maps.

Find:

What is the effect on the calculated factor of safety (FOS) if a 10 percent error was made in estimating the value of either input parameter?

Solution:

To examine the effects of an estimation error, a sensitivity analysis is performed. The program XSTABL (XSTABL ver 4.102 1992) was used to compute a range of values for the sensitivity analysis. One input parameter of interest was varied while all other input parameters were held constant. Figure 61.3 presents a graph of the effect each input parameter has on the resulting FOS.

From the graph it can be seen that a 10 percent reduction in the friction angle will result in a 12 percent reduction in the FOS. However, a 10 percent increase in the seismic coefficient will result in only a 5 percent reduction in the FOS. Therefore, the input parameter of most concern for this particular review is the friction angle.

PROFIL FILE LM3E3P35 2-1 l 93 10:31 H Lucky Moon Waste 3:l SC=0.30 phi=35

8 5 150.0 7318.0 175.0 7320.0 2 175.0 7320.0 930.0 7575.0 1 930.0 7575.0 1010.0 7572.0 1

1010.0 7572.0 1190.0 7518.0 1 1190.0 7518.0 1200.0 7520.0 2 175.0 7320.0 550.0 7350.0 2 550.0 7350.0 1035.0 7505.0 2 1035.0 7505.0 1190.0 7518.0 2

SOIL 2 120.0 120.0 .0 35.00 .WO .0 0 120.0 I200 .0 35.W .OW .o 0

EQVAKE ,300 .WO

SURFAC

Lucky Moon Waste 3: l SC=0.30 phi=35 I m b u factor of Safety for Specified Surface - 1 l G Z

Figure 61.2.-XSTABL stability analysis: ( a ) input file and (b) analysis profile.

- - x l m l c coefficient - - frlctlon angle

Figure 61.3.Sensitivily analysis of seismic coefficient and friction angle showing the effect each has on FOS.

61.3.3 Hazard Hazard assessment is a process used to quantify the uncertainty of input parameters Assessment and for a model. Risk analysis considers the consequences of undesirable outcomes by Risk Analysis comparing hazards and the associated outcomes in a decision analysis process. The

differences between a hazard assessment and a risk analysis is given by Miller (1988) as:

. . . hazard assessment differs from risk analysis. Two areas may have similar estimates of predicted hazard but entirely different risks. For example, the slopes in two different land-type areas may have similar probability of failure values, but one of the areas may have a much greater associated risk due to a greater potential for losing an important transportation route, recreation benefits, or natural resources such as timber.

For limit-equilibrium slope stability analyses that compare resisting forces to driving forces, the uncertainty to quantify is typically associated with material properties, hydrostatic conditions, and location of the failure surface. Some input parameters may be known with some certainty and can therefore be considered deterministic without introducing much error. A sensitivity analysis can help to identify those input parameters most influential on the stability analysis results. The probability of failure most often is calculated by using a stochastic simulation method.

The benefits of risk analysis over deterministic analysis are given by Klausner as cited by White et al. (1977):

1. Uncertainty Made Explicit. The uncertainty which an estimator feels about his estimate of an element value is brought out into

the open and incorporated into the ... analysis. The analysis technique permits maximum information utilization by providing a vehicle for the inclusion of "less likely" estimates in the analysis.

2. More Comprehensive Analysis. This technique permits a determination of the effect of simultaneous variation of all the element values on the outcome of an [alternative]. This approximates the "real world" conditions under which an actual [alternative's] outcome will be determined ...

3. Variability of Outcome Measured. One of the most significant advantages of this analysis technique is that it gives a measure of the dispersion around the [alternative] outcome based on the expected [parameters] ...

4. Promotes More Reasoned Estimating Procedures. By requiring that element values be given as probability distributions rather than as single values, more reasoned consideration is given to the estimating procedure. Judgment is applied to the individual element values rather than to the [alternative's] outcome which is jointly determined by all of the elements. Thinking through the uncertainties in a project and recognizing what is known and unknown will go far toward ensuring the best [alternative] decision. Understanding and dealing effectively with uncertainty and risk is the key to rational decision making.

61.3.3.1. Types of Uncertainty

Uncertainty (hazard or risk) can be described as objective or subjective (Palisade Corp., 1992). For there to be risk, there must also be a chance for loss. The geotechnical specialist must decide what type of hazard and risk is being analyzed and the resulting consequences of that risk. This process may involve defining many different outcomes for a given event.

An objective uncertainty or hazard is one that can be described by a repeatable process. Although the outcome is uncertain, the probability of an outcome is known (Palisade Corp., 1992). An example of an objective hazard is the roll of a die. Although it is not known with certainty that a roll of a die will result in the number 5, it is known that there is a 1 in 6 chance to roll the number 5.

A subjective uncerrainty or hazard may use models and analysis, but personal judgment and interpretation are used to define the uncertainty (Palisade Corp., 1992). For a subjective uncertainty, two people may assign completely different probabilities of occurrence and both be correct. It is also acceptable to "update" the uncertainty of a subjective hazard upon analysis. This may seem like manipulating the data to get the desired result, but this can be said about any process requiring subjective data. The idea is to select a value of uncertainty that truly reflects the views of the assignor given the available information. An example of a subjective hazard is defining the uncertainty of buying a used car. There is a chance the car could stop working the day after purchase or run as well as a new car. It is up to the buyer to assess the vehicle and subjectively judge these chances given personal experience, third-party research, and the proverbial "best guess." The risk to the

buyer is purchasing a vehicle that does not function properly; additional expense may be required or the vehicle may have to be re-sold at a sizable loss.

61.3.3.2. Quantifying Uncertainty

How does one quantify uncertainty? Uncertainty can be described using a probability distribution. The tools for describing a distribution are statistics and probability theory, with reliance on:

Proper model selection

Use of relevant sample data

Confirmation through field observations

Subjective estimates

Professional experience and judgment.

Data are required for statistical inferences (which is objective) but not entirely necessary for probability theory (which is subjective).

Statistical analysis is performed using representative samples from one or more populations. A sample of at least 30 values from a population is often considered appropriate to make reasonable estimates of the population's parameters. The ability to collect sample data is influenced by physical or legal access and by the availability of resources, such as funds, personnel, and time. A sensitivity analysis can be used to assist in the selection of input parameters to investigate.

When parameter estimates (e.g., mean, standard deviation, histogram) of a site- specific population are unavailable, general data published in the literature can be supplemented by professional judgment. The most common method is to initially quantify soil or rock properties based on some general classification method, such as the Unified Soil Classification System. These estimates can be updated as more site- specific data are collected, and computations such as back-analysis can provide additional insight on characterizing uncertainties on input variables (parameters).

A probability theory review is presented in chapter 2 of the LISA program documentation (Hammond et al., 1992), including several probability distribution types. Some helpful hints on selecting slope stability input distributions are presented in chapter 5 of the LISA manual.

61.3.3.3. Simulation Techniques

Simulation techniques allow for consideration of variability and uncertainty of model parameters. The input parameters are represented by random variables, which can be modeled by probability distributions as discussed in the previous section. These input distributions are repeatedly sampled using one of two common sampling methods, Monte Carlo and Latin Hypercube sampling, to generate repeated outputs of the model being simulated. Some advantages and disadvantages of simulation are given by White et al. (1977):

Advantages:

1. Analytical solutions can be impossible to obtain without great difficulty.

2. Simulation is useful in selling a system modification to management.

3. Simulation can be used as a verification of analytical solutions.

4. Less background in mathematical analysis and probability theory is generally required.

Disadvantages:

1. Simulation can be quite expensive.

2. Simulations introduce a source of randomness not present in analytic solutions.

3. Simulations do not reproduce the input distributions exactly (especially the tails of distributions).

4. Validation is easily overlooked in using simulation.

5. Simulation is so easily applied that it is often used when analytic solutions can be easily obtained at considerably less cost.

FOS is commonly used as the measure of slope stability. According to Hammond et al. (1992):

If we want to predict a possible value of the factor of safety, we take a possible value for each input variable and use the appropriate performance function (a stability equation) to calculate the corresponding value of the factor of safety. This is known as one Monte Carlo pass or iteration. In Monte Carlo simulation we generate a large number of factor of safety values (say 1,000) by repeated, random, independent samplings of a set of possible input values, and calculate a corresponding factor of safety for each pass. The set of possible input values for each input parameter is described by a probability distribution.

The @RISK program (@RISK ver 1.1 1992) is a spreadsheet add-in for risk analysis that uses either Monte Carlo or Latin Hypercube simulation. The LISA program documentation makes reference to the need for 1,000 iterations for Monte Carlo sampling stability (Hammond et al., 1992). The Latin Hypercube method makes use of sampling technology that provides stable results in fewer iterations. The difference in sampling techniques is that the Latin Hypercube method uses stratified sampling, rather than random sampling, to obtain a more consistent distribution of sampled values for fewer iterations. The use of Latin Hypercube sampling can reduce the desired number of iterations from 1,000 to 3W500 (Palisade Cop. , 1992).

Problem 6: This example problem will use the information from the Powder Creek SSI demonstration problem presented in appendix 6.4. A simulation will be performed to compare how the uncertainties of the model will influence the FOS for the horizontal drain alternative.

The probabilities of the input parameters will be selected objectively using deterministic information (the mean value) as a basis and applying the LaPlace principle of equal likelihood (see section 61.4.4.1). The wedge method of analysis (Duncan and Buchignani, 1975) was chosen for ease in creating a spreadsheet for use with the @RISK vet. 1.1 program add-in to Microsoft Excel. Three discrete water profiles were selected to represent the most likely locations of the water pressure profile, which was assumed to be phreatic (see figure 61.4).

3315

1340 1385 1430 1475 1520 1565 1610 1655 1700 X-AXIS (feet)

Figure 61.4.-Comparison of possible water profiles where W l =failure condition, W2 = partial drawdown, W3 =full horizontal drain drawdown.

Using the assumed soil strength input parameter distributions along with the assumed water profile at failure, the FOS was simulated. Once a water profile was selected for analysis, it was allowed to vary due to the uncertainty of the exact location. Figure 6L5 shows the resulting histogram of FOS values. It can be seen that the probability of the FOS being less than 1.0 is approximately 50 percent (expected value).

Figure 61.5.-Simulation for FOS at failure

To simulate the effects of drainage due to the horizontal drains, three possible water profiles are chosen. Water surface 1 assumes that the drains are ineffective and there is no drawdown; water surface 2 assumes partial drawdown; and water surface 3 assumes draw-down as designed. The distributions for the parameters are shown in figure 61.6. A sample simulation worksheet is shown in figure 61.7. Figure 61.8a shows the resulting histogram for FOS after all the input parameters are included to simulate the use of horizontal drains. By using the program's ability to solve for a target value, we find that the probability of failure (FOS less than 1 .O) is 18.66 percent, as shown in figure 61.8b.

Figure 61.6.-Input parameter distributions used for simulation: (a) soil units I and 2; ( b ) soil unit 3; (c) water profile; (d ) water profile changes.

I[ Ordlnary and Wedge Method of Analys~s (2-6-93 RAR)

I : ORD. F.S.. 1 2 1 l I I I ! :WEDGE F.s.. i r . 2 0 ~ , $ I

I 1 I

Figure 61.7.4ample simulation worksheet.

Figure 61.8-Horizontal drains alternative FOS simulation results: (a) histogram and (b) simulation statistics

61.4 Decision Analysis- How to Select the Preferred Alternative

61.4.1 Why DO a A decision analysis can be a useful tool in analyzing all aspects of a project, from Decision choosing a subsurface investigation method to selecting a mitigation alternative. Analysis? According to Miller et al. (1987):

Decision analysis is an extension of risk analysis wherein the evaluated risks of various decision alternatives are compared to provide input for decision making. A decision analysis requires that values (monetary or otherwise) be placed on consequences of the decision alternatives; thus, it is almost always project specific.

Reasons to perform a decision analysis include (Prellwitz and Hammond, 1989):

1. Risk and uncertainty can be systematically considered in a decision analysis.

2. Any decision problem can be analyzed.

3. All possible decision alternatives and contingencies can be documented and analyzed in a rational manner prior to taking any action.

4. A decision tree is a "progress map" to follow the sequence of events in a project and to re-evaluate remaining alternatives as new information is obtained.

5. A decision analysis provides a mechanism for consistent, rational action in achieving a goal over a series of decisions.

If nothing else, a decision analysis provides a means for documenting the assumptions and rationale used in making decisions.

The steps to performing a decision analysis include (Prellwitz and Hammond, 1989):

1. Define mutually exclusive alternatives and a method for equivalent comparison.

2. Define mutually exclusive and exhaustive outcomes resulting from each alternative.

3. Estimate the probability of each outcome and conditional benefits andlor costs of each outcome.

4. Compare the alternatives using an appropriate method that considers the uncertainty and risk assigned to the possible outcomes.

61.4.2 Decisions When performing an analysis of an alternative, a geotechnical specialist selects a Uslng Assumed model, method of analysis, various analysis parameters, and construction cost values. Certainty This is typically done by selecting single values for each of the required inputs

necessaq to perform the analysis. When a comparison is made between alternatives, the decision is made under the assumed certainty of the selected input values. We generally do not know exactly what all the input values are, but for relative comparison purposes we assume with certainty that the input values are correct and that the resulting model used for analysis will occur. This is commonly referred to as a deterministic analysis.

61.4.3 Decisions Most stability problems examined by the geotechnical specialist can have more than Consldering Risk one outcome. This can be the result of uncertainty in the model or its input

parameters, which often is due to lack of knowledge or to spatial and temporal natural variations in the parameters. The geotechnical specialist must decide which model outcomes and input variabilities are significant enough to examine pmbabilistically, as opposed to assuming certainty, when considering the consequences of an event.

61.4.3.1. Dominance

The dominance principle can be used to eliminate certain alternatives. If an alternative exists such that for any possible outcome it will always be selected over a second alternative, then that second alternative may be eliminated. This method is commonly applied during an analysis but is typically not documented as such. Cost is the most common comparison factor used when applying the dominance principle.

61.4.3.2. Expected Value

If more than one outcome ( j ) is to be examined, alternatives ( i ) can be compared based on their expected value (EV). The expected value of an event (A,) for all possible values of outcomes (VU) is given by the equation

The term p(j) is defined as the probability of outcome j occurring. The expected value can be thought of as the long-term resulting mean value for repeated sampling events or occurrences. This definition also provides a means for the geotechnical specialist to define the probability of an outcome by examining long-term records and using the frequency of occurrence. The most common criterion in selecting an alternative is to maximize expected gain or minimize expected costs.

61.4.3.3. Most Probable Future Principle

If an outcome has a considerably greater probability of occumng compared to any other outcome, the most probablefuture principle (MPFP) considers that outcome as certain and all other outcomes as having zero probability of occurrence. The result is a decision under assumed certainty. This is a common method of making daily decisions on a personal as well as professional level. The idea is to weigh the consequences of "ignoring" other outcomes to simplify the decision process. The

geotechnical specialist must be careful not to use this principle to eliminate outcomes with associated risks that may seem unimportant from a geotechnical viewpoint but may be considered extremely important to the resource decision maker.

Problem 7: Using the information from the Powder Creek SSI demonstration problem presented in appendix 6.4, create a summary table of alternatives.

I Design Life 25 Year Outcome Alternative Possible Outcomes Probability

1 I I Non-maint. failure 1 5096 1 85,000

I I I I

B - Shear Trench

C - Riprap Buttress

I success I I 45mo

54,700 50% A - Maintenance

I failure I 10% 1 249,700

1.25

< 1.00 "status quo"

failure

success

D - Horizontal Drains

I I ( failure 1 5% 1 345,000

I I I I

Use the methods of section 61.4.3 to select a preferred alternative considering risk.

10%

90%

1.19

82,200

249,700

300,000

IA - Maintenance ] A before C & E 1 69,850 1 use EV 1

success

failure

95% E - Drilled Piles

I B - Shear Trench I B before C & E 1 48,720 1 45.000 1

Alternative

80%

20%

1 S O

Expected Value ($) Dominance

I Prefemd Alternative by stated method I D I D I

37,200

91,900

success

MPFP ($)

-

C - Riprap Buttress

D - Horizontal Drains

E - Drilled Piles

61.4.4 Decisions 61.4.4.1. The LaPlace Principle Using Uncertain Outcomes The LaPlace Principle is a good beginning point for any probabilistic analysis.

Simply stated, if one cannot determine the likelihood of any outcome, then the

C before E

D before C & E

eliminate

249,700

48,140

302,250

249,700

37,200

300,000

outcomes should be considered equally probable. Once a probability has been assigned, the analysis becomes a decision considering risk.

61.4.4.2. The Optimist Principle

The optimist principle chooses the "best" outcome for each event, then chooses the "best" event. This method tends to negate the consequences of certain outcomes and is thus not recommended for geotechnical analysis.

61.4.4.3. The Pessimist Principle

The pessimist principle chooses the "worst" outcome for each event, then chooses the "best of the worst." This principle attempts to limit your losses. This method may be appropriate when the geotechnical specialist wants to emphasize the worst consequences of the alternatives and recommend "the lesser of two evils."

- -

Problem 8:

Use the information from the summary of alternatives from problem 7, except consider the outcome probabilities as uncertain. Show the preferred alternative using LaPlace, optimist, and pessimist principles.

Required:

Use the methods of section 61.4.4 to select the preferred alternative using uncertain outcomes.

Solution:

Create a table comparing the alternative methods and specifying the preferred alternative.

I Alternative I =lace EV I optin& Cost I Pessimist Cost I

I A - Maintenance 1 69,850 1 54.700 1 85,000 1

I D - Horizontal Drains 1 64,550 1 37,200 1 91,900 1

B - Shear Trench

C - Riprap Buttress

63,300

249,700

E - Drilled Piles

Preferred alternative by stated method

45,000

249,700

322,500

B

82,200

249,700

300,000

D

345,000

B

61.4.5 The Use of 61.4.5.1. What Is EMV? Decision Trees for EMV Analysis EMV, or expected monetary value, is the product of the probability that an outcome

occurs and the conditional value of that outcome. The conditional value is the value that is realized only if a particular outcome occurs (Prellwitz and Hammond, 1989). A value can be a benefit or a cost. The idea is to maximize the expected monetary gain. If costs are presented as positive values, then the idea is to minimize the monetary cost.

EMV is interpreted as the average monetary value per decision that is expected to occur if that same alternative is selected over a series of repeated trials. In real- world problems, a decision is generally made only once, and every decision alternative has different possible outcomes and probabilities. The "repeated trials" aspect is satisfied when the decision maker consistently selects the alternative which optimizes the EMV (highest benefits or lowest costs). The total benefits for all decisions over the long-term will exceed the losses using any other strategy for making a decision under uncertainty. Selecting the alternative with the optimum EMV is a strategy or philosophy for decisionmaking rather than an absolute measure of profitability or cost (Prellwitz and Hammond, 1989).

Problem 9:

Examine a data base of 200 similarly interpreted cut slope failures. One hundred failures were repaired using a similar buttress design, while the remaining one hundred were treated as needed (maintenance). For the buttress alternative, the average cost per site for 60 of the 100 was $20,000; 25 sites needed additional treatment at an average total cost of $25,000 per site; the remaining 15 of 100 sites needed treatment in addition to the buttress, for a total average cost of $30,000 per site. For the maintenance repairs, the average cost per site for 60 of the 100 was $15,000; 25 sites needed additional treatment at an average total cost of $30,000 per site; the remaining 15 sites needed additional treatment at an average total cost of $60,000 per site (major reconstruction after failure). Upon first comparison, the cost of maintenance appears less than constructing a buttress. Now consider the total and average long-term costs as follows:

all buttresses = 60 x $20,000 + 25 x $25,000 + 15 x $30,000 = $2,275,000

average buttress = $2,275,0001100 = $22,750

all maintenance = 60 x $15,000 + 25 x $30,000 + 15 x $60,000 = $2,550,000

average maintenance = $2,550,0001100 = $25,500

total cost savings of using a buttress = $2,550,000 - $2,275,000 = $275,000

average savings using a buttress = $275,0001100 = $2,750

The EMV equation for a decision alternative is defined as the sum of the probability of an outcome i(p(i)) multiplied by the value (Vi) of the outcome for all mutually exclusive outcomes (i=l to n). The fraction of cut slope failures in each cost category can be taken as the probability of occurrence, resulting in the following EMV's:

expected cost per buttress = 0.60 x $20,000 + 0.25 x $25,000 + 0.15 x $30,000 = $22,750

expected cost of maintaining a cut slope failure site = 0.60 x $15,000 + 0.25 x $30,000 + 0.15 x $60,000 = $25,500

In conclusion, over the long term the buttress repair is the preferred alternative to optimize cost savings. It is also shown that the average long-term costs are equivalent to the EMV costs.

61.4.5.2. How to Use a Decision Tree

A decision tree is a graphical representation of sequential decisions. Symbols are used to distinguish the significance of a branch. Branching occurs from left to right, with probability of occurrence assigned to each outcome branch. Values are placed on a branch as appropriate. Expected values are used to solve the sequential decisions. To solve a decision tree, work proceeds along each branch from right to left. Some common symbols used are (White et al., 1977; Prellwitz and Hammond, 1989):

A = a decision point

A-- = a branch from a decision point representing an alternative that can be chosen at the decision point

0 = a node in the tree where chance outcomes exist

0--= a branch representing a probabilistic outcome for a given event

V = a known value associated with a particular branch

EV = expected value at a node of the outcomes

EMV = expected monetary value of a decision alternative

Some rules for decision tree analysis are given by Prellwitz and Hammond (1989):

1. Decision alternatives are mutually exclusive, which means that accepting one alternative precludes accepting any other at a given decision node.

2. Outcomes are mutually exclusive and exhaustive. This means that two outcomes cannot occur simultaneously and that all possible

outcomes must be considered. Obviously every possible outcome cannot be considered, but the full range of outcomes should be considered in at least a discretized form. An example may be that an aggregate source might yield all, some, or none of the rock required. In reality, the rock source could yield any amount between all or none. The exhaustive requirement is satisfied mathematically when the probabilities of the outcomes sum to 1.

3. The decision alternative which optimizes the EMV should be selected consistently to maximize gain over the long term. As appropriate, the highest EMV should be chosen when the analysis is in terms of positive benefits or the least EMV should be chosen when the analysis is in terms of positive costs.

4. A decision tree is commonly used to display the decision analysis Draw a decision tree and solve from right to left. Use different symbols for decision and outcome nodes.

Problem 10:

The simplified decision tree shown below illustrates the procedure of creating the tree and calculating EMV's. The steps are:

(1) The two mutually exclusive alternatives for the site are: (a) use a 3/4:1 cut slope or (b) use a I:1 cut slope.

(2) The two possible stability outcomes are: (a) routine maintenance (minor slough material) or (b) failure (remove large slide material).

(3) The probabilities of each outcome are estimated from past experience based on similar site geology and terrain. The probabilities may also be estimated using Monte Carlo simulation (see section 61.3.3.3). The initial cost of each alternative is for excavation per station. The cost of each outcome varies by the additional cost of maintenance.

(4) The decision tree (dollar amounts are positive costs in thousands):

murine ($20) $710

~ 0 . 9 5

murine ($20) $590

314: 1 cutslope ($570)

pO.85

failure ($290) PO. 15 $860

(a) EMV of 3/4:1 = 0.85*$590 + 0.15*$860 = $630.50

(b) EMV of 1:l = 0.95*$710 + 0.05*$790 = $714.00

Conclusion: This decision analysis indicates that using a 3/4:1 cut slope is the preferred alternative.

61.4.5.3. EMV Analysis Using Imperfect Information

The first step in any subsurface investigation problem is to formulate a subsurface model based on past experience in a given geologic environment and on a surface investigation (such as a field-generated cross-section). We are using the decision analysis to answer the question, "How much will drilling reduce the expected costs for a project?" Drilling or any subsurface investigation technique, such as drive probes or geophysics, should help refine the conceptual subsurface model, increasing our confidence that it is close to the true state of nature and thus ultimately enabling a "correct" design decision to be made. However, subsurface investigation techniques, unskilled drilling inspection or logging of the drill hole, or incorrect interpretation of the significant subsurface features can prevent the subsurface investigation from revealing the "true" state of nature.

The probability that the investigation method supports your conceptual model-given that your conceptual model adequately describes the true state of nature-is termed the reliabiliry or confidence of the investigation method. This reliability is a conditional probability. The reliability of drilling and the initial probabilities about the true state of nature are used to calculate the probability that drilling will indicate a particular state of nature using the Total Probability Theorem. In other words, there are two ways that drilling might indicate a particular state of nature: (1) drilling might support a conceptual model that adequately describes the true state of nature, or (2) drilling might support a conceptual model when the true state of nature is different. The Conditional Probability Theorem can be used to calculate the probability that your model is the true state of nature given that drilling supports your model as the true state of nature. The use of conditional probability is further demonstrated in appendix 6.5.

Problem 11:

Consider the data presented in problem 9. Upon further review by a geotechnical specialist, three outcomes (models) can be identified. Model 1, a low subsurface water condition, has a 60 percent probability of occurrence. Model 2, a high subsurface water condition, has a 25 percent probability of occurrence. Model 3, a failure surface progressing below the roadbed, has a 15 percent probability of occurrence. These models and probabilities are based on past experience in similar geologic terrain. If we had "perfect" knowledge using 100 percent reliable investigation techniques, we could visit a new site and know which model represented the failure. Then, knowing the model from which to design, we would choose the least costly alternative. Under the repeated trials concept, the result would be an EMV of $19,750 as shown in the decision tree for perfect knowledge (figure 61.9). (Note that the monetary values in the figure are in thousands of dollars.

Figure 61.9.-Decision tree using "perject" knowledge for Problem 11

However, our field investigation techniques are not 100 percent reliable. Because we do not possess perfect knowledge of the subsurface conditions, we use conditional probability to predict our probability of a given outcome. Our confidence in a non- drilling estimate depends on experience and proper use of such tools as the field developed cross-section, drive probes, and geophysics. A possible range of confidence might be from 50-70 percent using these methods. Using a value of 60 percent, the resulting EMV is $24,150 (figure 61.10).

IHVLSTk.AmN INDICATES SELECT

Figure 61.10.-Decision tree using 60 percent confidence for Problem 11

Now we want to evaluate the cost of obtaining more reliable information, such as by drilling. Again, confidence will depend on the knowledge and skill level of the drill inspector during planning, drilling, and interpretation of the data obtained. Confidence levels may range from 7C-95 percent. The differential cost of drilling must be added to all of the possible outcomes. Using a reliability of 90 percent results in an EMV of $23,850 (figure 61.1 1).

PXOk%mLITI~ MODEL I.&

CONFIDENCE M@NII= . . . . . . . .

LTF.RI-*T,,%S S W . 1.00

Figure 61.1 1.-Decision tree using 90 percent confidence for Problem 1 1

The results of this analysis indicate that increasing the confidence from 60 percent to 90 percent is worth the $3,000 cost of drilling. The result is to decrease the EMV cost from $24,150 to $23,850.

61.4.6 Decisions Prellwitz and Hammond (1989) stated that: Using Preference Theory It should be noted that it is perfectly legitimate for there to be other over-

riding concerns which cause the decision maker to reject the alternative having the optimum economic benefit. For example, the district ranger might wish to minimize excessive environmental impacts and therefore may elect selective cutting just because the probability of having excessive environmental impacts is less, and the costs estimated for those impacts are less than for clearcutting.

Preference theory is used to represent quantitatively decision parameters upon which such judgments are based. That is, for a particular district ranger the amount of environmental impact (in terms of dollars, cubic yards of sediment, or some measurable quantity) that cannot be tolerated is quantified and used as a decision criterion.

Miller and others (1987) noted that:

The concept of expected preference value (EPV) is similar to that of EMV, except that values are expressed in terms of subjective preferences rather than in terms of money.

EPV is a relative value that reflects a decision maker's attitudes about gains and losses. These attitudes are affected by things such as personal risk preferences, amounts of money involved, current financial status, and any current or long-term objectives.

Values of the EPV for a given decision analysis are expressed in terms of a preference function (see figure 61.12). This function can be developed (derived) in several ways, such as using historical behavior of the decision-maker (i.e., which alternatives were selected under what conditions in the past) and using an interview procedure whereby the decision-maker is presented with various hypothetical situations and requested to make choices.

situation for choice (i.e., monetary, resource)

Figure 61.12.-Example preference function. - - - p p p p p p -

EPV is preferable to EMV in circumstances where:

EMV is not relevant or cannot be calculated for the attribute(s) of interest.

The overall EMV for a decision alternative is positive, but one or more of the potential losses is greater than the decision maker can afford.

Current and forecasted socioeconomic conditions may make certain decision alternatives quite favorable, even though they have relatively low EMV's.

The goal is to evaluate the outcomes of a series of decision alternatives by comparing a conservative attitude against a gambling attitude.

6J. Construction Control

Ed Rose, Georechnical Engineer, Klainath National Forest

6J.1 The material in this section has been liberally borrowed from publications of the

Acknowledg- United States Department of Agriculture. United States Department of Transportation,

ment the United States Department of the Interior, the United States Department of Labor, and the Transportation Research Board. The specific publications are listed in the reference section, and, if more detail is needed regarding material in this section, reading these excellent publications is highly recommended.

6J.2 General The proper design of a slope stabilization structure requires that the materials used

hiformation (whether earth, manufactured, or both) have adequate strength to withstand the forces to which they may be subjected during its lifetime. These materials should be used such that the cost of the structure will be reasonable for the conditions present at the site.

In slope stabilization construction, it is common practice to use material that is available on-site or nearby rather than specifying one that may have the desired properties but is located a considerable distance from the site. A number of different procedures can be used to enable earth materials to be used satisfactorily in slope stabilization. It is important when using earth as material in slope stabilization construction that the personnel in charge of construction control be familiar with the design requirements and that they ensure that the structures meet these requirements (US. Dept. of the Interior, 1974). The geotechnical engineer needs to make sure that the official in charge of inspection and construction control (this may well be the geotechnical engineer) understands any special geotechnical design or construction requirements before construction proceeds. This information should also be discussed at pre-bid and pre-work meetings so that the (prospective) contractor understands the project's special functional requirements.

Slope stabilization projects require a close relationship between design and construction. During construction, the construction engineer and project designer need to compare the design assumptions with the actual conditions observed in the field during construction. Differences between assumed and actual conditions and associated construction problems need to be resolved. This may require additional geotechnical analysis (using the actual parameters discovered during construction) and contract changes.

In summary, a high degree of cooperation and coordination is required of the contractor, construction engineer, design engineer, and geotechnical engineer. Field inspection during construction is vital to the success of the stabilization project.

65.3 Process Slope stabilization structures designed by Forest Service geotechnical engineers are. usually built under contract. The foundation of the contract is a set of plans and specifications containing a schedule of work items. Based on the information provided in the plans and specifications, the contractor submits a set of prices for accomplishing the items of work which, if accepted, becomes a part of the agreement between the Forest Service and the contractor. The primary purpose of the construction control organization is to ensure that the structure is built according to the plans and specifications; this includes tracking the items of work required and the payment due the contractor. Sometimes conditions in the field differ from those assumed to exist during preparation of the plans and specifications (see section 6J.2). If a changed condition does exist and adjustments to the plans and specifications are required, a change order is agreed upon by the contractor and contracting officer. It is imperative that the geotechnical engineer keep abreast of the project, either by being on the site during critical phases of the project or by being kept informed by the construction engineer. If a changed condition exists, then he or she can be involved in the technical aspects of the changes to the plans and specifications.

6J.4 Organization

6J.4.1 Forest Construction quality assurance is performed by three primary functional positions: Service the contracting officer (CO), the contracting officer's representative (C0R)--usually Construction Staff the construction engineer-and the inspector or inspectors. The inspector reports to

the COR, who represents the CO in the field and has limited authority with respect to contractual changes. All major changes must by authorized by the CO.

The Forest Service construction staff administers the contract (submits payment vouchers, keeps daily diaries, records quantities, etc.), inspects the work to ensure contract compliance, and provides contract quality assurance.

65.4.2 During construction, the geotechnical specialist usually functions as technical advisor Geotechnlcal to the COR or the CO. In some cases, the geotechnical specialist is designated as an Specialist inspector; in special cases, if he or she is certified as such, a COR.

65.5 plans and The natural variability of soil, and the large number of methods used to treat the soil,

Specifications make it impractical to write specifications to cover every condition that might exist for a slope stabilization project. Also, the experiences of the contractor and Forest Service construction staff are probably not the same; therefore, they will interpret the requirements of the work at hand in the context of their respective backgrounds. It is not unusual to find differences of opinion between the contractor and Forest Service construction staff concerning the intent and requirements of the plans and specifications, especially at the beginning (US. Dept, of the Interior, 1974). If this occurs, the geotechnical engineer may have to explain the requirements for the work in a way that is acceptable to both the contractor and contracting officer. This problem often can be avoided by good pre-work communication.

Specification requirements for earthwork construction are of two types: those based on performance (also known as end-result specifications) and those based on

6J.6 Inspection

procedures ( U S Dept. of the Interior, 1974). It is important to know the distinction between the two. Usually, when performance is the requirement, it is not correct for the project engineer to demand that a particular construction procedure or equipment be used to produce a specific result. When procedure is the basic requirement, the project engineer may not demand that a specific performance (or end result) be attained.

Specification requirements are also of two other kinds: explicit requirements and those characterized by such phrases as "as directed by the contracting officer or project engineer." The second type of requirement is usually avoided wherever possible. There are three cases in which the non-explicit type of requirement is used: (1) producing minor dimensions in areas where investigations to determine such dimensions are not warranted; (2) areas where any of several possible methods would be satisfactory; and (3) new conditions for which requirements have not been established (US. Dept. of the Interior, 1974).

Contract compliance is determined by visual examination, measurement, and testing (U.S. Dept. of the Interior, 1974). The extent to which each of these methods is used will depend on local conditions, the value and importance of the work being inspected, and the skill and experience of the inspector. The balance of these methods also will vary as the project progresses. During the beginning of the project, intuitive (experiential) and visual judgment should be verified at frequent intervals by tests and measurements. In some cases, the amount of measuring and testing can be reduced as the work progresses, but it should never be entirely eliminated (U.S. Dept. of the Interior, 1974).

The inspector's job is to check for contractual compliance, not to create contractual requirements (specifications). Questions regarding compliance should be referred to the COR, who, in turn, may request interpretation or other technical advice from the geotechnical specialist.

Contract specifications outline dimensional, quality, and product requirements. Some specifications have target values for the dimensions or quality with specified tolerances on one or both sides of the target value. Thus, the inspector is also responsible for documenting the extent of contractual compliance (the amount of work falling within the acceptable range of compliance).

The inspector should see that safety requirements are met and that no unlawful practices are being followed (US. Dept. of the Interior, 1974). He or she should be aware of the progress of the work and the future schedule of work and should prevent any part of the work from being omitted.

65.7 Contract Elements

6J.7.1 Erosion Control

Before construction is started, the engineer and the contractor should review the topography and the erosion control plan for the project. As the representative of the Forest Service, it is the engineer's responsibility to protect the property and facilities

adjacent to the project from environmental pollution originating from the project (Transportation Research Board, 1990).

It is in the contractor's best interest to control surface erosion so that damage to completed work is minimized and present and future work are not jeopardized. The benefits of erosion control are large. From the Forest Service point of view, the adjacent lands and streams need to be protected from construction pollutants; from the contractor's point of view, surface water is prevented from saturating the foundation and embankment area and facilitates such operations as trench excavation and subsurface drain installation.

65.7.2 Earthwork Satisfactory earthwork is most easily accomplished by careful monitoring of contract compliance early in the contract (Transportation Research Board, 1990). Spending a significant amount of time closely inspecting the work at this stage will be time well spent because, once the contractor has established a satisfactory earthwork operation, the project generally runs smoothly. Input from the geotechnical engineer can be highly beneficial at this stage of construction. For example, he or she may alert the inspector to the fact that the material being compacted is a clayey silt which is highly moisture-sensitive, Knowing this, the inspector would carefully monitor the process or end result to ensure that excess water is not added to the material. It is important for the inspector to be familiar with the contractor's proposed earthwork construction plan because this will help in scheduling the inspector's duties. This in turn results in more efficient inspection requiring less field testing.

65.7.3 Clearing Clearing is the removal of all trees, brush, etc., within the work area. Grubbing is and Grubbing the removal of stumps and roots (Transportation Research Board, 1990).

Specifications should be read carefully to determine clearing and grubbing requirements.

Limits of clearing and grubbing are generally noted on the plans or in the standard specifications handbook. Areas outside the work limits are usually left in their natural condition unless otherwise designated in the plans. The geotechnical engineer usually does not get involved in the clearing and gmbbing portion of the contract.

6J.7.4 Slope Ditches

The specifications may require that top- or mid-slope ditches be completed before materials are removed from the cut. Some contractors like to ignore this requirement because it is not a pay item (as are excavation and fill embankment), assuming that they can complete this work in their spare time (Transportation Research Board, 1990). If the ditch work is not done in the beginning, problems that are harmful to the work may occur and expensive corrective procedures may be required.

Two problems may occur if ditches are not completed during the correct phase of construction. First, the ditches are designed to intercept surface runoff that would otherwise flow down the slope and into the excavation. This water could cause serious erosion and problems with cut slope stability. Second, if the water is allowed to flow into the excavation, the soil in the excavation and cutbank may become saturated, which will make it difficult to achieve specified compaction (Transportation Research Board, 1990).

To prevent such problems, the contractor should be required to submit a work plan describing progression of work. This work plan should be discussed during the pre- work meeting between the contractor and the Forest Service.

6J.7.5 Excavation The topsoil is usually unsuitable for use in compacted earth fills; therefore, it is normally stockpiled if it will be needed for use in landscaping of the project. The limits and depths of topsoil removal, where required, are usually included in the plans.

Because the properties of naturally occurring geologic deposits are highly variable, specifications for roadway excavation often consider all soil and rock as unclassified excavation. This avoids such problems as which bid price should apply to which material and whether a deposit can be ripped or must be blasted (Transportation Research Board, 1990).

6J.7.6 Rock Excavation

6J.7.7 Embankment Foundations

6J.7.8 Compaction

If the material is too competent to be ripped, then before drilling and blasting operations are begun, the COR, the contractor, and the owner's or agency's geotechnical specialist experienced in blasting should hold a meeting at the project site. All parts of the blasting operation should be discussed, and the specifications and rules for the contractor's operation and the basis for inspection of the work should be reviewed. The blasting specialist can determine what work needs the most inspection, what to look for, and how it should look (Transportation Research Board, 1990).

The embankment foundation provides the base upon which the fill is constructed. After clearing and grubbing, the contractor should establish the foundation according to the plans, specifications, standards, and any special provisions. Preparing the foundation may require removing the topsoil and organic deposits, adding material to establish a stable foundation, and compacting the ground surface. A geotechnical engineer should inspect the prepared foundation before the fill is placed on it. For problems encountered during the design, exact treatment will be shown on the plans or described in the contract documents. If foundation conditions differ from what was anticipated, the inspector should consult the design engineer and the geotechnical engineer. At this point, some revision to the design and construction plans may be required (Transportation Research Board, 1990).

Soil compaction is one of the most important parts of the earthwork portion of a slope stabilization project. An evenly and densely compacted embankment will provide a reasonable platform upon which to place running courses (in the case of road construction). If the embankment is uniformly compacted during construction, then, assuming the foundation is stable, the running course will behave uniformly. Compaction increases bearing capacity, slope stability, and resistance to frost action and decreases settlement and permeability. Poor compaction may result in general and differential settlement, which cause depressions, premature failure of the running course, and possibly failure of the embankment itself.

The contract should specify a maximum lift thickness. The specified thickness depends on the soils in the project area.

The contractor should be persuaded to drive hauling equipment as evenly as possible over the entire surface of the embankment during fill placement. This helps reduce the total amount of compaction required and minimizes rutting and damage that might be caused by heavy repetitive concentrated tracking by equipment (Transportation Research Board, 1990). Large loaded scrapers may overload even a densely compacted embankment. The performance of the embankment under the tires of heavy equipment will give a good indication of the uniformity and quality of compaction.

65.7.8.1 Compaction-Moisture Control

All projects that have the AASHTO compaction specifications should receive from the laboratory a set of laboratoly compaction control curves developed from standard (AASHTO T-99) or modified (AASHTO T-180) Proctor values representing the soil being used in the embankment. These curves are plots of density (dry unit weight) versus moisture (water content) and provide a standard of acceptability for the field tests (Transportation Research Board. 1990). When field compaction control tests are performed, the results are compared with the laboratory compaction control values to determine the applicable percent of maximum density, called the relative density or the percent compaction. The specifications will give the percent compaction required (i.e., 95 percent of T-99). If this value is not obtained, more compactive effort or a change in the water content will be required to reach the minimum specified value (Transportation Research Board, 1990).

Moisture control becomes more important as the particle size of the material being compacted decreases. Clays are more sensitive to changes in moisture content than are sands. Small amounts of fines in granular materials will have a marked effect on the moisture requirements. Well-graded materials will usually have steep, sharp compaction control curves showing a well-defined optimum moisture content (OMC). Uniform materials, especially sands, will show flat curves without a well-defined OMC. Uniform materials may be compacted properly at a large range of moisture contents (Transportation Research Board, 1990). For further discussion on moisture control, see section 4B.5.

65.7.8.2. Compaction-Weaving and Pumping

Weaving-and-pumping is an elastic-type deformation of soil. When loaded, the material deforms, and when the load is removed, the material springs back to its original position. The construction equipment looks as if it were riding on a wave as it travels over the fill. The soil will deflect and a wave will be created ahead of the wheel, but once the equipment moves on, the area looks unchanged (although there may be some cracking of the surface).

Weaving occurs when the soil is too moist and does not have time to drain as the load is applied, thus creating excessive pore water pressure. The load is shared by the soil structure and the pore water pressure. This gives a temporary elasticity to the soil and creates the weaving or pumping effect. In this condition the strength of the soil is dramatically lowered. The best way to solve the problem is to stay off the affected area and allow the excess pore water pressures to dissipate naturally (Transportation Research Board, 1990). The soil will then regain its strength. If the fill is weaving as a result of the compaction equipment, a smaller compactor will result in lower pore water pressures and help reduce weaving and pumping.

Initially the weaving does not damage the embankment, but if the excess loading is repeated, pore pressures will continue to build and may eventually result in shear failure or rutting. If a weaving condition exists, a geotechnical engineer should be consulted. Weaving conditions can be verified by testing for moisture-density specifications. Weaving could be caused by saturation of embankment material by ground water flow or surface drainage.

65.7.8.3. Rutting

Rutting is a surface shear or bearing failure caused by equipment moving across the embankment. When the loading exceeds the shear strength of the soil, the soil is displaced, the wheel sinks, and deep ruts occur. Rutting destroys uniform compaction and makes it difficult to place the next lift to uniform thickness. The quality of the work suffers. It is the contractor's responsibility to take corrective measures, such as changing the method of operation, materials, or loading (Transportation Research Board, 1990).

65.7.8.4. Compaction Equipment

The compaction equipment the contractor uses must follow the specification requirements and be approved by the COR. The type of equipment should be determined by the type of soils encountered; however, most of the time, the contractor will use whatever equipment he already owns or has leased (Transportation Research Board, 1990).

There are three general types of compactors: pneumatic-tired static, pneumatic-tired vibratory steel drum, and sheepsfoot roller. For additional information concerning compactors, see section 4B.5.

65.7.8.5. Compaction in Confined Areas

Compaction projects may be in large areas that are accessible to normal-sized compactors or in confined areas accessible only to smaller, highly maneuverable, or hand-operated mechanical compactors (Transportation Research Board, 1990). There is no precise definition of confined areas; therefore, each case should be considered on its own. Backfill behind retaining walls and minor structures are common examples of projects that require confined-area compaction.

Compaction equipment typically used in small, confined areas includes compactors with a tamping foot that is actuated by a small two-cycle engine, static vibratory plate compactors (that must be manually pushed or pulled), and small self-propelled static or vibratory rollers controlled by an operator walking behind.

Equipment that is not acceptable in some situations may be acceptable in confined areas. The basic question for a confined-area compactor is: "Will it do the job?" (Transportation Research Board, 1990). Material compacted in confined areas should be limited to lifts of not more than 6 to 8 inches before compaction. Some equipment may require thinner lifts. The material should be compacted with a sufficient number of passes to meet the specification density requirements (Transportation Research Board, 1990). The inspector should consult with the geotechnical engineer if he or she has any questions or problems with compaction in confined areas.

65.7.8.6. Compaction Inspection

65.8 Construction Control- Mechanically Stabilized Embankments (MSE's) and Walls

The field inspector has a number of things to check during compaction operations. There may be different density requirements for different parts of the stabilization project, and certain types of compaction equipment may work better with one soil type than another. Diligent inspection can ensure that the work is done according to the specifications.

Various methods are commonly used to specify compaction control. Performance of moisture-density measurements may be required at randomly selected sampling sites. If the project is large, the embankment to be compacted may be divided into lots and the acceptance of each lot determined by the statistical outcome of random sampling within it.

Embankment compaction may also be controlled by the use of test strips. A compaction test strip is specified for each soil type or for a certain number of lots. The type of compaction equipment and the number of passes required to achieve specified density is recorded. Acceptance of the work is determined by completion of the number of passes with the right equipment.

The inspector should ensure that the compactor applies at least the minimum number of passes at or below a maximum speed. The initial passes increase the density of the soil significantly. Therefore it is important that the minimum number of passes are applied so the bottom of the lift will be adequately compacted and there will be no significant settlement of that layer.

Corrective methods should be taken when the lift being compacted shows significant weaving, pumping, or rutting under the force of the compactor. The compactor's merely leaving a tire print should not be considered significant rutting; if the equipment displaces the soil laterally out of the wheelpath and leaves a visible rut. then the compaction procedure must be evaluated and modified.

The main purpose of inspection of the compaction operations is to validate that the soil is uniform and dense (Transportation Research Board, 1990). Once the inspector is aware of how the compaction and hauling equipment affects a correctly compacted layer, the contractor should be allowed to proceed with additional fill when the minimum specification requirements have been met.

It may be necessary to perform density tests more frequently at the beginning of the project or when a new soil type is encountered.

It is becoming very common to use some type of tensile reinforcement in backfills behind retaining walls, earth embankments, and slope stabilization projects (Transportation Research Board, 1990). A number of reinforcing materials, such as steel strips, sheets of geotextiles or geogrids, welded wire mats, metal grids or bars, and various anchor systems, have all been used successfully for reinforcement. Soils are relatively strong in compression and shear but are weak in tension. When a material of high tensile strength is placed in the soil, the soil mass possesses greater strength and is able to experience larger movements without distress. The mechanically stabilized embankment (MSE) structure must satisfy both external and

internal stability requirements and be durable enough to last the entire design life. For additional material concerning design and use of MSE's, see section 6F.

Earthwork construction control for MSE structures is basically the same as for conventional retaining structures and embankments, but some additional details require special attention. These usually include specifications for backfill material gradation, connection details, construction sequence, etc. (Transportation Research Board, 1990). Several of the proprietary firms have published quality control procedures and manuals. The Forest Service also has specifications for MSE structures. The contractor should obtain a copy of the proprietary manuals from the company and a copy of the Forest Service specifications for the project MSE structure. The recommendations for these structures should be followed as closely as possible. Field substitution of backfill materials or changes in the construction sequence, procedures, or details should be permitted only with the approval of the COR (preferably after consultation with the geotechnical engineer).

6J.8.1 Inspection Before construction begins, the inspector should discuss with the geotechnical Elements engineer-and become thoroughly familiar with-the following items: plans and

specifications, site conditions relevant to construction requirements, material requirements; and construction sequences for the specific reinforcement system (Christopher et al., 1990).

Proper inspection of MSE's, retaining walls, and embankments involves three principal areas (Boiling, 1986):

(1) Thorough review of plans and specifications and resolution of field conditions, especially regarding foundation conditions;

(2) Thorough inspection of retaining wall and embankment system components (footings, backfill, reinforcing elements, connections, and facia elements) to determine quality and workmanship prior to and during construction; and

(3) Careful inspection of wall and embankment erection with respect to vertical and horizontal alignment, proper layout of reinforcing elements, and backfilling operations.

65.8.2 Plans and The Forest Service inspector should become very familiar with the specification Specifications requirements for the specific type of system to be constructed. Particular attention

should be given to material requirements, construction procedures, soil compaction procedures, alignment tolerances, and acceptanceJrejection criteria. Plans should be reviewed, and unique and complex project details identified and reviewed with the geotechnical engineer and contractor. Particular attention should be given to the construction sequence, corrosion protection systems for metallic reinforcement, special placement requirements to reduce construction damage for geosynthetic reinforcement, soil compaction restrictions, and details for drainage requirements (Christopher et al., 1990). The contractor's document package should be checked to make sure that he or she has the complete set of construction plans, specifications, and contract documents.

65.8.3 Review of The site conditions should be checked to see whether there are any special Site Conditions construction procedures required for preparation of the foundation, site accessibility, and Foundation excavation for obtaining the required reinforcement length, and construction Requirements dewatering and other drainage features (Christopher et al., 1990).

The preparation of the foundation involves the removal of unsuitable materials from the area to be occupied by the retaining structure or fill. This is critical in the facing area because it will reduce facing system movements and will help maintain facing alignment along the length of the structure (Christopher et al., 1990).

Site accessibility will be required for all the construction equipment. A storage site will be required for reinforcement facing panels, reinforcement materials, and, most likely, backfill materials.

If construction of a reinforced fill will require a side slope cut, a temporary earth support system may be needed to maintain stability (Christopher et al., 1990). The contractor's method should be reviewed by the inspector and geotechnical engineer for safety and performance.

Slopes into which a cut is to be made should be carefully observed, especially following intense periods of precipitation, for any signs of seepage. Construction dewatering may be required for any excavations performed below the water table to prevent a reduction in shear strength due to hydrostatic water pressure (Christopher et al., 1990). A geotechnical analysis of slope stability is recommended for large, unbraced construction excavations.

65.8.4 Material Requirements

Material acceptance should be based on a combination of material testing. certification, and visual observations (Christopher et al., 1990). The inspector should thoroughly inspect all material when it is delivered to the site. Once the material is on-site it should be carefully stored and handled to avoid damage. The material supplier's construction manual and Forest Service specification handbook contain additional information on this matter.

65.8.4.1. Reinforcing Elements

Reinforcing materials (strips, mesh, and sheets) should arrive at the project site securely packaged or covered to avoid damage. They come in a variety of material types, configurations, and sizes, and even a basic structure may have different reinforcement elements at different locations (Christopher et al.. 1990). The inspector should verify that the material is properly identified and should check the specified designation (AASHTO, ASTM, etc.). Material verification is especially important for geotextiles and geogrids, because many product styles look similar but have different properties (Christopher et al., 1990). Mesh reinforcement should be checked for gross area and length, width, and spacing of transverse members. The length and thickness of strip reinforcements should be checked.

Protective coatings, such as galvanization or epoxy, should be checked by certification and inspected for defects.

65.8.4.2. Precast Elements

6J.8.5 Backfill

Facing elements delivered to the project site should be examined before erection for the following deficiencies or defects: insufficient compressive strength (minimum recommended, 4,000 psi); imperfect molding; honey-combing; cracking, chipping, or spalling; color or finish variation on front face; out-of-tolerance dimensions; and misalignment of connections. Repair to the damaged facing elements should be done to the satisfaction of the COR and the geotechnical engineer (Christopher et al., 1990).

65.8.4.3. Facing Joint Materials

Bearing pads (cork, neoprene, and SBR rubber), joint filler (synthetic foam), and joint cover (geotextile) should be properly packaged to minimize damage in unloading and handling (Christopher et al., 1990). For example, some geotextiles must be protected from sunlight during storage and thus should be covered.

Although these joint materials are considered to be miscellany, use of the wrong material or incorrect placement can result in significant wall distress (Christopher et al., 1990).

65.8.4.4. Placement of Reinforcing Material

Before geosynthetics, especially geotextiles, are placed, they should be protected from sunlight and extreme temperatures. If they are packaged properly, this should be no problem.

After the reinforcement is in place, it should be examined carefully, and, if any damaged or tom materials are found, they should be removed or repaired according to the specifications or the geotechnical engineer's recommendation. Construction equipment should never be allowed to operate directly on any reinforcement before backfill is placed. Geosynthetic reinforcement should be unrolled transverse to the alignment of the embankment or wall, and wrinkles and folds should not be present (Transportation Research Board, 1990). Procedures for seams and overlaps shown in the plans and specifications should be strictly followed.

Reinforced backfill structures usually require at least 95 percent of the maximum dry density, as determined by AASHTO T-99, in order to mobilize adequate frictional properties along the reinforcement (US. Department of Agriculture, 1975).

65.8.5.1 Placement

On the majority of Forest Service projects, the backfill for MSE walls and embankments is gathered from the project site and placed with reinforcement. A facing is added if it is to be a wall. The material in the fill is normally tested for the standard engineering properties during the design stages, and it is the results of these tests that are used to design the MSE wall or embankment.

Consistent compaction is critical. The lift thickness of wall fill must be controlled according to specification requirements and vertical location of reinforcement

elements. The uniform loose lift thickness of the reinforced backfill should not exceed 12 inches (Christopher et al., 1990).

Moisture and density control is critical. Wall fill material should be placed and compacted within 2 percent on the dry side of the optimum moisture content. Placement moisture content can have a substantial effect on the reinforcement-soil interaction. Moisture content higher than optimum makes it increasingly difficult to maintain an acceptable facing alignment, especially if the fines content is high (Christopher et al., 1990). Moisture content that is too low could result in significant settlement during periods of precipitation.

Reinforced backfill should be deposited onto or parallel to the rear or middle of the reinforcements and bladed toward the front face. Construction equipment should not come into direct contact with the reinforcements. Soil layers should be compacted up to or even slightly above the elevation of each level of the next reinforcement prior to placing that layer of reinforcing elements (Christopher et al., 1990).

65.8.5.2 Compaction Equipment

Within 3 feet of the wall or slope face, use small single or double drum vibratory rollers or vibratory plate compactors. Placement of the backfill in the front should not lag behind the rest of the structure by more than one lift. Poor fill placement and compaction in this area has sometimes resulted in vertical void immediately behind the facing elements (United States Department of Transportation, 1990). Within this 3-foot zone construction control should be maintained by method specification (such as three passes of a vibratory plate compactor). High quality fill is sometimes used in this zone to achieve the desired properties with less compactive effort. Use of excessive compactive effort or overly heavy equipment near the wall face may result in lateral displacement of the wall face or structural damage and overstressing of the reinforcement elements (Christopher et al., 1990).

Generally, with the exception of the 3-foot zone directly behind the wall face or slope face, large smooth drum vibratory rollers should be used to obtain the desired compaction (Christopher et al., 1990). Sheepsfoot rollers should not be used because of their potential to damage the reinforcements.

Inconsistent compaction, under-compaction due to insufficient compactive effort, and "compaction" of backfill by trucks and dozers will lead to settlement and alignment problems and should not be permitted. Compaction control testing of the reinforced backfill should be done regularly throughout construction.

6J.9 Subsurface water is usually removed with an underdrain, such as a geotextile-lined,

Con~truction gravel-filled trench or a prefabricated geocomposite drain. For a more thorough

Control- discussion of drainage and drains, see section 6D. There are several important points

Underdrains to remember when installing geotextiles and geocomposites to ensure that they work as designed. They should be protected from exposure to sunlight, dirt, and contamination during shipping and storage. Installing and backfilling should be done carefully in order to avoid tearing or puncturing the geotextile. Damage to the core must be avoided, and the drainpipe and its outlets must be properly located and aligned.

6J.10 Excavations and Trenches- Safety, Stability, and Design

6J.10.1 General

65.10.2 Collapse Prevention

In many slope stabilization projects, temporary trenches and excavations are needed. (See sections 6D, 6F, and 6G for examples of slope stabilization projects that use temporary excavations and/or trenches as part of the construction process.) These methods of construction pose a serious risk to all persons involved with the project. The greatest risk is in collapse, and when this occurs there is a higher probability of worker fatalities than with other excavation-related accidents (U.S. Department of Labor, 1991). Therefore, this section will cover the main points in the safety, stability, and design of excavations and trenches. For more in-depth coverage of this subject, refer to the U.S. Department of Labor (1990) and (1991), both of which are excellent.

According to the Occupational Safety and Health Administration (OSHA), "A trench is referred to as a narrow excavation made below the surface of the ground in which the depth is greater than the width-the width not exceeding 15 feet. An excavation is any manmade cut, cavity, trench, or depression in the earth's surface formed by earth removal" (US. Department of Labor, 1991). Excavations can be made for anything from cellars to highways.

OSHA requires that in all excavation where employees are exposed to potential cave-ins, they must be protected by: (1) sloping the sides of the excavation. (2) benching the sides of the excavation, (3) supporting the sides of the excavation, or (4) placing a shield between the side of the excavation and the work area. The contractor is allowed to choose the most practical of these four design approaches. Once a method has been selected, the required performance criteria must be met by that system (US. Department of Labor, 1991).

The OSHA standard does not require the installation and use of a protective system when an excavation is made entirely in stable rock, or is less than 5 feet deep and a competent person has examined the ground and found no indication of potential cave-in (U.S. Department of Labor, 1991).

The standard prohibits excavation below the level of the base or footing of any foundation or retaining wall unless a support system such as an underpinning is provided, the excavation is in stable rock, or a registered professional engineer examines the structure and decides that it is sufficiently far from the excavation to pose no hazard to employees (U.S. Department of Labor, 1991).

The OSHA standard provides several different methods and approaches for designing protective systems used to protect against cave-ins (US. Department of Labor, 1990). Each is described in the following subsections.

65.10.2.1. Sloped Sides

One method is to slope the sides of the excavation to an angle not greater than 1% horizontal to 1 vertical (34" measured from the horizontal). These slopes must be excavated in a configuration in accordance with those for type C soil found in appendix B of the OSHA standard (US. Department of Labor, 1990). A slope of this angle or less is considered safe for any type of soil (U.S. Department of Labor, 1991).

65.10.2.2. Tabulated Criteria

A second method, which is applicable to both sloping and shoring, involves using tabulated data approved by a registered professional engineer. The tables specify required slope angles (on sides of trench) or shoring as a function of soil type and trench depth. The data must be in writing and must include enough explanatory information, including the criteria for determining the selection and the limits on the use of the data, to enable the user to make a decision (US. Department of Labor, 1991).

One copy of the above information, along with the name of the registered professional engineer who approved the data, must be kept at the worksite during construction of the protective system. A copy must be available to OSHA if they request it

65.10.2.3 Trench Box

A trench box or shield may be used by the contractor if it has been designed or approved by, or is based on tabulated data that has been approved by, a registered professional engineer. OSHA standards allow the use of a trench shield as long as the protection it gives is equal to or greater than that which would have been provided by the appropriate shoring system.

65.10.3 Analysis When designing a protective system, some or all of the following factors may be and Design involved in the design:

Depth of cut

Soil conditions (i.e., type, strength, structure, unit weight)

Ground water conditions

Duration over which the excavation will be open

Space available around the excavation and presence of nearby buildings and roads

Economics

The stability of excavation slopes depends, in general, on:

Strength of the soil

Unit weight of the soil

Depth of the excavation

Slope angle

Pore-water pressure

When a geotechnical engineer is assessing the stability of shallow open cuts, his past experience is usually a g w d guide. Deep cut stability should be based on geotechnical investigation and slope stability analysis. If an accurate analysis is impossible due to unusual soil conditions, conservative estimates of soil and ground water parameter values should be used in the stability analysis. Studying the condition of previous cuts or trial cuts in the area is also useful. A11 deep trenches should be monitored closely-preferably instrumented-for signs of collapse.

In attempting to improve the stability of open cuts, the geotechnical engineer should try to decrease the stress within the cut or increase the soil strength. Stress may be decreased by flattening of the slope, partial support, or surface drainage control. Soil strength may be increased by dewatering (drainage), temporary freezing, use of piles, or densification.

6K. Post-Construction Monitoring of the Technical Structures and Projects

Ed Rose, Geotechnical Engineer, Klamath National Forest

6K.1 Acknowledg- ment

6K.2 Introduction

6K.3 Limited Monitoring Program

6K.4 Comprehensive Monitoring Program

Most material in this section has been borrowed from the Federal Highway Administration (US. Department of Transportation, 1985) and Christopher et a1 (1990). Dunnicliff (1988) is also a good source.

Post-construction monitoring is an important part of any slope stabilization project. It does not have to be highly complex but should be consistent in application and methodology. Monitoring performance is important because it allows for verification of the original design assumptions and provides data for improvement of future analysis and design. The long-range objectives are to improve safety and reduce cost.

Limited observations and monitoring will normally include: (1) horizontal movements of structure face, (2) vertical movements of the surface of the overall structure, (3) local movements or deterioration of the facing elements, (4) drainage behavior of the backfill, and (5) performance of any structure supported by soil, such as approach slabs for bridge abutments and footings.

Horizontal and vertical movements can be monitored by surveying methods, using applicable measuring points on the structure. Permanent benchmarks are required for vertical control, and one horizontal control station should be provided at each end of the structure. If surveying is not feasible, then strictly visual observation should be conducted. Visual cues include face movement or cracks in or near the structure.

Drainage can be monitored by visually observing outflow points during storm events or through open standpipe piezometers installed near the base and in the back of the structure.

Comprehensive studies involve monitoring of surficial and internal behavior of the soil in the stmcture.

6K.5 Planning Monitoring Programs

6K.5.1 Purpose ol the Monitoring Program

6K.5.2 Define the Project Conditions

6K.5.3 Predict Mechanisms That Control Behavior

6K.5.4 Select the Parameters to be Monitored

A defined, organized plan should be used in both the limited and comprehensive monitoring programs. This section provides a step-by-step process for developing such a program. Christopher et al. (1990) includes a checklist for use in making sure all the steps in the planning process have been followed.

The first step is to define the purpose of the measurements. Instruments should be selected, used, and located accordingly. If there are no questions, then there should be no monitoring (Christopher et al., 1990). The questions and the purpose should be clearly established.

The person responsible for the monitoring program should be thoroughly familiar with the project. Conditions that could effect the structure's performance or the instrumentation should be considered.

The characteristics of the subsurface, backfill material, and any reinforcement andlor facing material as they relate to the behavior of the overall structure must be recognized before the instrumentation program is begun.

Some parameters to be considered in monitoring are: pore water pressure, joint water pressure, total stress, deformation, load and strain in structural members, and temperature (Christopher et al., 1990).

For structures involved with soil and reinforcement, important parameters that should be considered include (Christopher et al., 1990):

Horizontal movements on the face of the structure

Vertical movements of the surface of the overall structure

Local movements or deterioration of the structures facing elements.

Drainage behavior of the backfill

Performance of any structure supported by soil (and reinforcement), such as approach slabs for bridge abutments or footings.

Horizontal movements within the overall structure.

Vertical movements within the structure.

Lateral pressure at the back of facing elements.

Vertical stress distribution at the base of the structure.

Stresses in the reinforcement, especially magnitude and direction of the maximum stress.

Relationship between settlement and stress-strain distribution.

Stress relaxation in the reinforcement with time.

6K.5.5 Predict Magnitudes of Change

6K.5.6 Devise Remedlai Action

6K.5.7 Assign Monitoring Tasks for Design, Construction, and Operation Phases

6K.5.8 Select Instruments

Total horizontal stress within the backfill and at the back of the reinforced section.

Aging condition of reinforcement.

Pore pressure response below structure.

Rainfall, which causes changes in other parameters.

From the design, the magnitudes for each parameter should be predicted. Predictions should be made to establish the required range and accuracy of each instrument (Christopher et al., 1990).

A maximum level for each parameter may be required to provide a warning. An action plan for remedial action should be established in case any maximum parameter value is exceeded.

When assigning tasks, the person or people with the greatest vested interest in the data should be given direct responsibility for producing it accurately (Christopher et al., 1990). Instrumentation personnel should have a fundamental knowledge of geotechnical engineering and electronics and an attention for detail. Below is a list of tasks that need assigning, usually in the planning stage of the project.

TASK ASSIGNMENT LIST (Christopher et al., 1990)

(1) Plan monitoring program.

(2) Procure instruments.

(3) Calibrate instruments.

(4) Maintain and calibrate instruments on regular schedule.

(5) Establish and update data collection schedule.

(6) Collect data,

(7) Process and present data.

(8) Interpret and report data.

(9) Decide on implementation of results.

Instruments should be selected on the basis of reliability and simplicity (Christopher et al., 1990). Things to consider in deciding which instruments to pick are the

overall cost of procuring, calibration, installation, maintenance, monitoring, and data processing.

Limitations in skill level for installing the instruments should be evaluated. This may change the instrument selection and the monitoring program. Access for instrumentation installation and monitoring should also be included in this evaluation.

Table 6K.I is an instrumentation selection chart with recommendations on the types of instruments for monitoring reinforced soil structures.

Parameters

Horizontal movements of face

Vertical movements of overall structure

Local movements or deterioration of facing elements

Drainage behavior of backfill

Horizontal movements within overall structure

Vertical movements within overall structure

Performance of structure supported by reinforced soil

Lateral earth pressure at the back of facing elements

Stress distribution at base of structure

Stress in reinforcement

Stress distribution in reinforcement due to surcharge loads

Possible Instruments

Visual observation Surveying methods Horizontal control stations

Visual observation Surveying methods Benchmarks

Visual observation Crack gauges

Visual observation at outflow points Open standpipe piezometers

Surveying methods (e.g., transit) Horizontal control stations Probe extensometers Fixed embankment extensometers Inclinometers

Surveying methods Benchmarks Probe extensometers Horizontal inclinometers Liquid level gauges

Numerous possible instruments (depends on details of structure)

Earth pressure cells Strain gauges at connections Load cells at connections

Earth pressure cells

Resistance strain gauges Induction coil gauges Hydraulic strain gauges Vibrating wire strain gauges Multiple telltales

Same instruments as for stress in reinforcement

Parameters Possible Instruments

Relationship between settlement and stress-strain distribution

Same instruments as for: Vertical movements of surface of overall Structure Vertical movements within mass of overall structures Stress in reinforcement.

Earth pressure cells

Stress relaxation in reinforcement Same instruments as for stress in reinforcement

Tohl stress within backfill and at back of reinforced wall section

Earth pressure cells

Pore pressure response below structures

Rainfall

Barometric pressure

Open standpipe piezometers Pneumatic piezometers Vibrating wire piezometers

Ambient temperature record Thermocouples Thermistors Resistance temperature devices Frost gauges

Rainfall gauge

Barometric pressure gauge

6K.5.9 Select The selection of instrument locations is based on predicted behavior. Zones of Instrument concern are identified, such as weak zones, heavily loaded zones, and zones of Locations anticipated high pore pressures, and the instrumentation is appropriately located.

When considering zones, variations in geology and construction procedures must also be included.

When selecting locations, survivability of instruments is important and additional replacement instruments should be on hand in case the original ones become inoperative.

Figures 6K.1 and 6K.2 show examples of instrumentation layouts for a wall and a slope. These are based on actual M A monitoring programs.

- . - - "

- i II : I

STRIPPED GROUND - I " - L

W I, -

Figure 6K.1.-Layout of instrumentation for comprehensive monitoring of reinforced bac&jill walls. (Reprinted with permission of STS Consultants from Christopher et al., 1990)

(ASSUMEO LOCATION r OF MAXIMUM STRESS)

Figure 6K.2.-Layout of instrumentation for comprehensive monitoring of reinforced embankment slopes. (Adapted with ~ermission of STS Consultants from Christooher et 01.. 1990)

6K.5.10 Plan It is important to keep records with emphasis on factors that could affect Recording of instrumentation results. Items that should be included during the construction phase Factors That May are as follows (Christopher et al., 1990): Influence Measured Data An installation record sheet for each instrument.

Construction details, progress, and delays.

Visual observations of unusual behavior of the structure

Activities around instrumentation locations.

6K.5.11 Establish Procedures for Ensuring Reading Correctness

6K.5.12 Prepare Budget

6K.5.13 Write Instrument Procurement Specifications

6K.5.14 Plan installation

Environmental factors such as temperature, rainfall, snow, sun, shade, etc.

Records of the subsurface conditions.

Instruments need to be calibrated before any measurements are taken. Anomalous readings do not always mean that the instrument is not working correctly but may indicate unusual behavior.

In critical situations, backup may be provided through use of a number of instruments or by using more than one type of instrument. Data correctness can also be determined by evaluating consistency; for example, when dealing with consolidation, an increase in pore water pressure should be consistent with added loading. Repeatability is also a way of examining correctness. Repeat readings should be taken when feasible over short periods of time to evaluate consistent response of the instrument and reading method (Christopher et al., 1990). Visual and optical surveys can be used to support readings.

Prepare a budget during the planning stage for the tasks listed in the task assignment list in section 6K.5.7. A common error in budget preparation is to underestimate the duration of the project and the actual data collection and processing costs.

Purchase of other than the most simple geotechnical instruments should not be considered a routine construction procurement, because, if the measurements are to be valid, critical attention must be paid to quality and details. Requirements for factor calibrations should be determined so that acceptance tests may be performed to ensure that the instruments are functioning correctly when first received.

Prepare the installation plan in advance of the installation dates. Detailed guidelines for installation of instruments should be given in the manufacturer's instruction manual for the specific instrument in question. In the installation procedures there should be a list of required tools and materials. An installation record sheet should be designed and include such things as listed in table 6K.2. Installation personnel should be given training in installation. Access needs for installation should be arranged and designs developed for protecting instruments from damage and vandalism. If there is construction in the instrumentation area, consider scheduling installation when on-site construction is inactive.

Table 6K.2.-Possible content of installation record sheet. (Adapted with permission of STS Consultants from Christopher et al., 1990).

( ) Project name.

( ) Instrument type and number, including readout unit.

( ) Planned location in plan and elevation.

( ) Planned orientation lengths, widths, diameters, depths, and volumes of backfill around instrument.

( ) Personnel responsible for installation.

( ) Planned lengths, widths, diameters, depths, and volumes of backfill around instrument.

( ) Plant and equipment used, including diameter and depth of any drill casing used.

( ) Date and time of start and completion.

( ) Spaces for necessary measurements or readings required during installation to ensure that all previous steps have been followed correctly.

( ) A log of appropriate subsurface data.

( ) Type of backfill used around instrument,

( ) As-built location in plan and elevation.

( ) As-built orientation.

( ) As-built lengths, widths, diameters, depths, and volumes of backfill around instrument.

( ) Weather conditions.

( ) A space for notes, including problems encountered, delays, unusual features of the installation, and any events that may have a bearing on instrument behavior.

6K.5.15 Plan Carry out factory calibration and acceptance tests when instruments are first received Regular as described in section 6K.5.13. This section is concerned with calibration and Calibration and maintenance during service life. Maintenance

6K.5.16 Plan Data Collection, Processing, Presentation, Interpretation, Reporting, and Implementation

Portable display units are especially sensitive to changes in calibration, usually as a result of mishandling and lack of regular maintenance (Christopher et al., 1990). Required calibration frequency depends on the specific instrument but usually calibration should be done on a regular schedule rather than being arranged haphazardly.

Changes in calibration have sometimes resulted in doubting the correctness of data since the last calibration date. A sticker on the instrument or on the data sheet should indicate the previous and next scheduled calibration date.

A maintenance and instruction manual should accompany each type of instrument. The manual should include a troubleshooting guide; cleaning, drying, lubricating, and disassembly instructions; and recommended maintenance frequency. Detailed guidelines for calibration and maintenance should be provided by the manufacturer of the instrument.

Many projects have files filled with large quantities of partially processed data because sufficient time or funds were not available for these tasks. The time required to accomplish these tasks should not be underestimated.

The following steps are required (Christopher et al., 1990):

(1) Plan data collection: Prepare preliminary detailed procedures for data collection of initial and subsequent data. Prepare field data sheets. Plan training. Plan data collection schedule and duration with the purpose of the monitoring program in mind. Plan access needs.

(2) Plan data processing and presentation: Determine need for automatic data processing. Prepare preliminary detailed procedure for data processing and presentation. Prepare calculation sheets. Prepare data plot format. Plan training.

(3) Plan data interpretation: Prepare preliminary detailed procedures for data interpretation.

(4) Plan reporting of conclusions: Define reporting requirements, contents, frequency.

(5) Plan implementation.

6K.5.17 Write Contractual Arrangements for Field Instrumentation Services

Field services include instrument installation and data collection, processing, interpretation, and reporting. This assumes, of course, that funds are available to contract field instrumentation and also that the expertise to handle some or all of these tasks is not available on forest or within the agency.

6K.5.18 Update Budget

Assemble final budget, taking into consideration all the tasks listed in section 6K.5.7.

6K.6 Executing Monitoring Program

6K.7 Discussion on the Use and Misuse of Instrumentation

The key steps that should be followed in executing a monitoring program include (Christopher et al., 1990):

Procure instruments.

Install instruments.

Calibrate and maintain instruments on a regular schedule.

Collect data.

Process and present data.

Interpret data.

Report conclusions.

There are numerous reasons for using geotechnical instrumentation after completion of a slope stabilization project. If done properly and for the right reasons, it can make an imponant contribution to existing or future projects (US. Department of Transportation, 1985). Instrumentation should not be used unless there is a proper reason and this reasoning can be defended.

The following are reasons for using geotechnical instrumentation (US. Department of Transportation, 1985):

Safety.

Verifying design adequacy.

Verifying long term satisfactory performance.

Proving adequacy of a new construction technique.

Verifying a contractor's compliance with specifications.

Legal reasons.

Observational method.

Advancing the state-of-the-art.

Diagnosing the specific nature of an adverse event.

The following excerpt is taken from Peck (1970, from United States Department of Transportation, 1985) in which he discussed the use and misuse of instrumentation.

( 1 ) Instrumentation has become a catchword-all too often the sponsors do not get their money's worth.

(2) More than a few projects seem to be provided with elaborate instrumentation at considerable expense on the hopeful premise that surely the results of such a praiseworthy endeavor will be capable of answering some question that will turn out to be of interest. With data in hand, the investigator searches for a question that the findings can answer. If he fails, and the odds favor that he will, he can hardly disguise the barrenness of the investigation. If pressed to state the question to be elucidated, he can suggest only a vague generality, such as improving the present state of knowledge. How, he is not sure.

(3) Measurements, even if dignified by calling them the results of instrumentation, do not in themselves make for improved understanding or better practice. They must be made with a purpose, a purpose that can be expressed in the form of a question related in some way to an hypothesis.

(4) The emphasis should be on observation rather than instrumentation. Observation is much the broader term, and includes instrumentation. Even with the most sophisticated instrumentation, other types of observation are essential--details of construction often have significant or even overwhelming influence on the behavior of the structure and of the surrounding soil. For an understanding of the behavior, these details must be observed and recorded.

(5) An instrument too often overlooked in our technical world is a human eye connected to the brain of an intelligent human being. It can detect most of what we need to know about subsurface construction. Only when the eye cannot directly obtain the necessary data is there a need to supplement it by more specialized instruments. Few are the instances in which measurements by themselves furnish a sufficiently complete picture to warrant useful conclusions.

(6) Field observations provide a powerful and sometimes indispensable tool in applied soil mechanics. It concerns me that the legitimate use of instrumentation may be set back by a rising tide of disillusionment on the part of those who have been persuaded to embark on elaborate programs that promise too much. It concerns me that too many programs are based on the number of instruments to be used rather than on the questions to be answered. It concerns me that sophistication and automation are substituted for patient prooftesting of equipment under field conditions. To the extent that such practices prevail, they must be discouraged so that the observational approach itself will not be discredited. We need to carry out a vast amount of observational work, but what we do should be done for a purpose and done well.

In the second and third paragraphs of the excerpt, Peck states that sometimes the investigator, having done elaborate instrumentation and gathered considerable data, looks for a question that the findings can answer-the opposite of the scientific method. In other words, impo~tant observations need to be tied to the scientific method in the planning of the instrumentation project. Also, certain questions need to be asked: (1) Why are you gathering the data? (2) Of what value is it to your purpose? (3) How will you use it?

In the fourth, fifth, and sixth paragraphs he stresses the importance of observation in general, especially visual observation. Observing details of construction are critical because they may have significant effect on the structure and soil. To understand this effect it must be observed and recorded. This shows the importance of construction control and inspection.

Also stressed is the fact that when the visual method cannot obtain necessary data then there may be a need to supplement it with specialized instruments. Instruments by themselves usually cannot obtain all the data needed to draw useful conclusions.

In conclusion, instrumentation is used to supplement and extend manual observations and not to replace them.

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U.S. Department of Labor. 1991. Excavations. Washington, DC: Department of Labor, Occupational Safety and Health Administration.

US. Department of Labor. 1990. Excavation and trenching standard. Washington, DC: Department of Labor, Occupational Safety and Health Administration.

U.S. Department of the Navy. 1982a. Foundations and earth structures. NAVFAC DM 7.2. Alexandria, VA: Department of the Navy, Naval Facilities Engineering Command.

U.S. Department of the Navy. 1982b. Soil mechanics. NAVFAC DM 7.1. Alexandria, VA: Department of the Navy, Naval Facilities Engineering Command.

U.S. Department of Transportation. 1985. Geotechnical instrumentation training manual. Washington, DC: Federal Highway Administration.

US. Department of Transportation. 1976. Slotted underdrain systems. Implementation Package 769. Washington, DC: Federal Highway Administration. 1-7.

References

UTEXAS2 version 1.211. 1987. Austin, TX: University of Texas, Department of Civil Engineering.

Vandre, B.C. 1992. Dump safety factors. Interoffice memorandum (December 10). Ogden, UT: US. Department of Agriculture Forest Service, Intermountain Research Station.

Vandre, B.C. 1980. Stability of non-water impounding mine waste embankments. Ogden, UT: U.S. Department of Agriculture, Forest Service, Intermountain Research Station.

Vandre, B.C. 1975. A case history of a landslide caused by confined groundwater. Bulletin of the Association of Engineering Geologists 12(4):261-273.

White, J.A.; M.H. Agee; and K.E. Case. 1977. Principles of engineering economic analysis. New York, NY: John Wiley and Sons. 480 p.

Williamson, D.A. 1989. Geotechnical field methods. Engineering Field Notes 21 (MarchIApril). U.S. Department of Agriculture, Forest Service. 31-36.

Wooten, J. 1971. Wooten's third law of steel column design. Modem Steel Construction, 1:48 Chicago, IL: American Institute of Steel Construction.

XSTABL version 4.102. 1992. Moscow, ID: Interactive Software Designs, Inc., 953 N. Cleveland Street.

References

APPENDIX 6.1

6.1 Horizontal Drains Design Example-Camp Five slide

Abstract

Michael T. Long, Engineering Geologist, Willamette National Forest

A major slope failure in the central portion of the Cascade Range in the Willamette National Forest, westem Oregon, was stabilized in December 1983 with a network of horizontal drains which continually intercept and divert a localized aquifer (Long, 1986). The "Camp Five Slide" had a histoly of yearly progressive failures from 1981 to 1983. During the winter of 1982-83, the failure increased in size from 30,000 to 250,000 cubic yards.

Exploration and investigative methods included surveying and photogrammetric mapping of the failure and adjacent areas, 1,500 feet of vertical drilling, in-situ permeability testing, electrical resistivity profiling, and ground water tracing using fluorescein and rhodamine dyes. The ground water was contoured using observation well levels to determine equipotential and flow line directions.

Stability was analyzed using the Janbu method of slices. Soil strength parameters were determined by direct shear and triaxial consolidated undrained tests, and by repeated direct shear and plasticity index core relation to residual shear strength. The drain system was designed using Darcy's Law and Manning's Equation with the data from ground water profiling and permeability testing.

The success of the system was demonstrated by a total discharge from 49 individual drains of over 500 gallons per minute (gpm) during periods of peak rainfall, and averaging 200 gpm since installation. The static water level within the failed mass decreased 14 feet after installation.

A resurvey of the slide area in July 1984 indicated a vertical displacement of only 0.2 feet, which is considered to be the result of soil consolidation due to dewatering. No horizontal displacement of the failed mass was measured. No further movement has been noted to date.

Physiographic The Camp Five Slide is located at 1,900 feet elevation, 13 miles northeast of

Setting Oakridge, Oregon, in the central portion of the Cascade Range within the boundaries of the Willamette National Forest. Yearly progressive failures that occurred from 1981 to 1983 blocked a major timber haul route and threatened the water quality and channel characteristics of the North Fork of the Middle Fork of the Willamette River. The North Fork was designated a state scenic waterway by the Oregon Legislature in 1983. It is a major recreation area for fly fishing and a water supply source for the town of Westfir, located 10 miles downstream from the slide area. Average

Appendix 6.1

discharge from the North Fork during peak flow is approximately 1,300 cubic feet per second (cfs).

Site Geology The project site is on the valley side slope of the North Fork. The drainage headwaters are on the north end of Waldo Lake, which is approximately 15 miles to the east. The valley was formed in Oligocene to early Miocene lapilli tuff of the Little Butte Formation (Peck et al., 1964). During the late Pliocene and early Pleistocene, a sequence of eruptions from high Cascade volcanoes caused olivine-rich basalt lavas to flow down the river channel, completely filling the valley. Subsequent erosion during the last 3 million years has formed a valley of inverted topography, whereby the intracanyon flows are now perched 1,000 feet above the valley floor (Priest and Vogt, 1983).

Till deposits indicate evidence of multiple alpine glacial advances in this and the adjoining valleys. Basaltic andesite flows overlying glacial tills have been dated at 675,000 k12.000 years in the adjoining Salmon Creek Valley (Long and Leverton, 1984). Varved proglacial silt deposits up to 30 feet thick are visible in the river cut from the bridge just east of Kiahanee Campground, 7 miles upstream from the site. These glacial advances, combined with alluvial erosion, eventually breached the intracanyon flow at the project site. At this point, the river turns 90" from an east-west course, along the south side of Christy Flats, to a north-south course, along the west side of High Prairie. This multiple-process valley erosion has produced several events of side slope collapse of the intracanyon basalt, mainly by undercutting of the lower-strength lapilli tuffs along the contact and by removal of lateral support after ice recession.

Chronology of Forest Service Road 1926, which crosses the toe of the failure, was constructed in

Events 1935. A single-lane concrete bridge was constructed in 1973 to replace the old bridge. During the summer of 1981, the road was reconstructed to two lanes, which required an alignment shift of 15 feet into the previous cut slopes, resulting in an increase in cut height from 15 to 25 feet, while retaining a 1:l cut slope ratio. Natural side slopes along the cuts range from 20 to 30 percent. Water was observed seeping from the old cut slopes during the winter and was considered a common occurrence in the vicinity. Several small cut failures, less than 10 cubic yards each, occurred during the end of construction. At the end of November 1981, 6,000 cubic yards of material failed, resulting in a head scarp 15 feet high and 200 feet from the ditch line. The failure blocked the road and pushed material into the river channel.

Slide correction, consisting of removing all failed material back to a stable cut slope, began in April 1982. Springs were evident during excavation. Construction was completed by laying the slope back to 1:l and placing a rock blanket over areas where springs were observed. Total project cost was $80,000.

On December 5, 1982, an additional 10,000 cubic yards of material failed over the roadway. The total discharge from a number of springs at the head scarp was meas- ured at 50 gpm. A field reconnaissance and a geotechnical investigation were initiated. During the course of the investigation, on February 19, 1983, a final failure occurred, which increased the failed mass to 250.000 cubic yards with a horizontal displacement of 100 feet. After this failure, the head scarp was 500 feet from the

Appendix 6.1

original ditch line. The scarp height was increased to 30 feet, and spring discharge was measured at over 100 gpm.

Investigation (December 1982 to February 1983)

Field A cross-section was surveyed through the centedine of the slide to a point 1,000 Reconnaissance horizontal feet behind the head scarp to determine the slide geometry and material

relationships (figure 6.1.1). Three additional cross-sections were surveyed across the lateral portions in order to construct a three-dimensional interpretation. Site materials were identified during the reconnaissance and classified into soil and rock units according to the Unified Soil and Unified Rock Classification Systems.

The morphology of the site suggested that an ancient drainage had been filled by previous failure of the upslope intracanyon basalt flows. Ground water concentrations and movement through this material accelerated decomposition of the underlying lapilli tuff to clay minerals. The likelihood of previous movement was high, suggesting residual shear strength values for soil unit C, a plastic sandy silt (MW.

The failure surface was tentatively identified as a contact zone between the colluvial slope collapse material of the intracanyon basalt, soil unit E , and the underlying soil unit C, which was exposed in two areas of the slide. The subsequent drilling investigation was conducted to confirm and refine this initial interpretation.

I

Figure 6.1.1-Centerline profile of Camp Five Slide, February 1983.

Appendix 6.1 937

Drilling Three drill holes were completed behind the head scarp, and I-inch PVC observation wells were installed. Due to the size of subsurface rock fragments, the standard penetration test was not possible. Drilling investigation consisted of 8-inch hollow- stem auger holes and continuous NQ wireline core sampling.

Ground Water The source of the ground water was identified as a waterfall 1.500 feet upslope of Tracing and 500 vertical feet above the slide. The corresponding streamflow entered the

subsurface 100 feet below the plunge pool into the talus slope. To confirm the ground water source, the ground water was traced using fluorescein and rhcdamine WT dyes. Rhodamine WT was placed in the waterfall, and direct samples were taken every 6 hours from a spring in a backhoe trench behind the head scarp with a portable ISCO water sampler. Samples were analyzed using a Turner filter fluorometer (Turner and Associates, 1971). Positive indications of the presence of rhodamine were evident in 3 days.

Fluorescein disodium salt was later placed in the waterfall. Samples were taken indirectly by placing packets of activated charcoal at discharge points in the slide mass and at two culvert outlets west and east of the failure limits. The fluorescein was adsorbed onto the charcoal, then extracted using a 5 percent solution of potassium hydroxide in 70 percent isopropyl alcohol, and then analyzed in the fluorometer. The tracing results indicated that portions of the ground water flow above the slide branched in at least two directions.

Electrical Resistivity Profiling

In order to define the limits of the ground water that was directly affecting the failure, a bore hole (DH-6) was located on a road that runs perpendicular to the slide centerline approximately 1,000 feet north of the head scarp. The hole was placed in what appeared to be a topographic low area, 1,000 feet downslope from the waterfall. The hole was advanced to 60 feet, with no ground water encountered. The contact between the colluvium and the decomposed pyroclastic unit was encountered at 50 feet. At this point, an electrical resistivity profile was obtained east and west of DH-5 in order to determine whether the ground water might be channeled in another location. The profile shown in figure 6.1.2 indicated a rapid drop in apparent resistivity 150 feet west of DH-6. A boring 150 feet west of the DH-6 location encountered water at 15 and 23 feet.

Appendix 6.1

Figure 6.1.2-Apparent electrical resistiviry profile above failed mass.

Investigation (February 1983 to March 1983)

Field ~~~~~~~i~~~~~~ On February 19, 1983, after the final failure, the centerline section was re-surveyed,

and the change from the previous section indicated a horizontal displacement of 96 feet downslope (see figure 6.1.1). One boring (DH-5) was found intact in the middle of the failed mass, enabling ground water measurements to be made in the failed mass during and after subsequent rainfall and drainage installation. Brass hubs were set in concrete on the slide perimeter and within the failure, and an EDM survey was made for future horizontal and vertical control. An aerial photo reconnaissance was made, and the area was photogrametrically mapped.

Drilling Drilling exploration continued along an east-west line above and perpendicular to the failure direction along the road above the slide. The results indicated a concave subsurface contact extending across the lateral portion of the slide area. This information confirmed the belief that the failure was contained within a prehistorically filled drainage, which could contain a localized aquifer. A total of 22 vertical borings was made to confirm this configuration. Inclinometer casing was installed in two of the borings. The construction plan for the site is shown in figure 6.1.3.

Appendix 6.1

Stability Analysis

Sampling and Testing

Method

Soil Units

An undisturbed block sample of soil unit C was obtained from an exposed portion of the failure surface. Laboratoly shear strength testing included consolidated undrained triaxial testing with pore-pressure measurements, undisturbed direct shear tests, and remolded direct shear tests.

The initial stability analysis used Bishop's modified method of slices, and effective stress parameters on the centerline cross-section obtained prior to the final failure. Assuming that excess pore-water pressure was the critical variable in the failure mechanism, back-calculation to achieve a factor of safety of 1.0 was performed by varying the ground water table. By using the ground water level observed in well DH-5, and the elevation of the springs at the scarp face prior to failure, a factor of safety of 0.98 was calculated, giving a high degree of confidence to the soil parameters that were used. Subsequent stability trials were calculated using Janbu's method of slices on a mainframe computer.

Final stability analysis was completed using residual shear strength values obtained from laboratory repeated direct shear test results and plasticity index values. The results showed that in order to re-open the road to two lanes, a 14-foot drawdown with a 2,000-cubic-yard buttress would provide an adequate factor of safety. These results have been repeated and confirmed using XSTABL (XSTABL v. 4.102 1992). These results are demonstrated by the boring logs, input files, and profiles shown in figures 6.1.4 through 6.1.15.

Soil Unit A (2): silty sandy gravel with some clay (GMfSMu); yellowish brown; less than 10 percent is 112- to 3-foot-diameter rock fragments, wet, above the plastic limit; low dry strength and toughness. Origin: failed material.

4 = 28" C = 30 psf

ym, = 115 pcf

Y.w = 125 pcf

Soil Unit B (1): silty sandy gravel with some clay (GMISMu); yellowish brown; less than 10 percent is 112- to 3-foot-diameter rock fragments, damp; low dry strength and toughness. Origin: consolidated glacial valley fill.

4 = 30" C = 250 psf

ymbt = 120 pcf

Y , = 130 pd

Appendix 6.1

1 HOLE ELEV.: ?o3/

Figure 6.1.4.-Log for drill hole DH-2

942 Appendix 6.1

Figure 6.1.5.-Log for drill hole DH-3Z Januaq 1983.

Appendix 6.1 943

Figure 6.1.5.-Log for drill hole DH-3Z, January 1983 (continued).

944 Appendix 6.1

) DRILL RlGJ @ D - ? o ROCK I HOLE DIA.: L/ ' 'FL ' ,< . - ' f i ~ ! ? g E

rc4E//nO /-:c'

Figure 6.1.6.-Log for DH-4Z.

Appendix 6.1 9 45

OLE NO. pH-52 HOLE ELZV.: 2 0 2 5 ,-

DISTFIICT: & ? ~ m ~ b f DEFTH OF HOLE: 5-9 . DATE: /-16-92 DEPTH . TO ROCK: - D R l U R I G C e r 2 ?

, HOLE DIA.: 4 ,--L,':..I 7 P ) ~ ~ Z . C ~ L ROCK OK DE3TH T

HOLE LOC.: r / /XJ y L o ~ / t O , o ~ ~ ? CASING 1

G fl/5fl/ 'A-

Figure 6.1.7.-Log for drill hole DH-52

946 Appendix 6.1

Figure 6.1.8.-Log for drill hole DN-I3Z.

Appendix 6.1 947

Figure 6.1.8.-Log for drill hole DH-13Z (continued).

948 Appendix 6.1

CAMP 5/Back Calc. Residual Shear St . Janbu Factor of Safety for Specified Surface = 1.009

0 I I I I I I 1 I I

0 55 110 165 220 275 330 385 440

X-AXIS (feet)

Figure 6.1.9.-XSTABL profile and outputfile for back-calculation of residual shear sfrength.

( XSTABL File: C51FINA.L 4-01-93 18:Ol

.......................................... XSTABL 4

Slope Stability Analyais using Simplified BISHOP or J M B U methods

Copyright (C) 1992 Interactive Software Dasigns, Inc.

All Rights Reserved

Forest Service US Dept. of Agriculture

MOSCOW, ID 83843

Ver. 4.10 1001 ..........................................

Problem Description : CAMP 5/Back Calc. Residual Shear St.

SEGMENT BOUNDARY COORDINATES

5 SURFACE boundary segments

Segment x-left y-left x-right y-right Soil Unit No. (ft) (it) (ft) (it) Below Segment

5 SUBSURFACE boundary segments

Segment x-left y-left x-right y-right Soil Unit No. (ft) (it) (ft) (ft) Below Segment:

Figure 6.1.9.-XSTABL profile and outputfile for back-calculation of residual shear strength (continued)

950 Appendix 6.1

ISOTROPIC Soil Parameters

3 type(s) of soil

soil Unit Weight Cohesion Friction Pore Pressure Unit Moist Sat. Intercept Angle Parameter Constant No. (pcf) (Psf) (do9 RU (Psi)

1 Water surface(s) have been specified

Unit weight of water = 6 2 . 4 0 (pcf)

Water Surface No. 1 specified by 5 coordinate points

.................................. PHREATIC SURFACE, ..................................

Point x-water y-water No. (ft) (ft)

1 . O O 29 .00 2 59 .00 40.00 3 9 3 . 0 0 5 0 . 0 0 4 1 9 1 . 0 0 8 6 . 0 0 5 374 .00 1 3 9 . 0 0

Trial failure surface specified by 7 coordinate points

Point x-surf y-surf NO. (ft) (ft)

Water Surf act

NO.

1 1 1

Figure 6.1.9.-XSTABL profile and output file for back-calculation of residual shear strength (continued),

Appendix 6.1 951

........................................ SUMMARY OF INDIVIDUAL SLICE INFORMATION : ........................................

slice x-base y-base height width alpha (ft) (ft) ( ft) (ft)

SLICE INFORMATION ... continued : Slice

1 2 3 4 5 6 7 8 9 10 11 12 13 14 15

Sigma (psf)

603.9 1369.6 1415.8 1297.0 1358.2 1246.7 1265.6 2296.7 3615.2 2908.3 1988.2 1222.4 513.5 336.2

-115.0

c-value ( P W

30.0 30.0 30.0 30.0

.o 250.0

.o

.o

.O

.o 250.0 250.0 250.0 250.0 250.0

beta

30.174 30.174 13.276 13.276 13.276 13.276 13.276 78.690 15.878 15.878 15.878 15.878 15.878 15.878 15.878

weight (lb)

26262.7 15773.7

165152.8 730.3 1711.1

12393.0 19343.4 14455.4

556708.9 20714.6 26557.3 24306.4 7977.5 2228.0 2620.0

Delta

.oo

.oo . 00

.oo

.oo

.oo

.oo

.oo -00 .oo .oo .oo .oo .oo .oo

For the single specified surface, Corrected JANBU factor of safety = 1.009 (Po factor -1.058)

Resisting Shear Strength = 250.47E+03 lb Total Driving Shear Force = 262.48E+03 lb

Figure 6.1.9.-XSTABL profile and outputfile for back-calculation of residual shear strength (continued).

952 Appendix 6.1

CAMP 5/Final Profile Before Drainage 10 most critical surfaces, MINIMUM JANBU FOS = 1.017

110 165 220 275 330 385 440

X-AXIS (feet)

Figure 6 1 10.-XSTABL profile and outputfile offinal profile before drainage. \O

iSTABL Fila: C52FINAL 4-02-93 10:04

****t**ttt*t*t**t**tt*t++******t**********t

4 XSTABL * Slope Stability Analysis using *

Simplified BISHOP or JANBU mothod8

t Copyright (C) 1992 Intaractiva Software Designs, Inc.

t All Rights Rosarvad t

Forest Sorvico US Dopt. of Agriculture

Moscow, ID 83843 t t

Var. 4.10 1001 tt*+*t+*t*****+*t*+t**t~*****e****t***t~*t

Problem Dascription : CAW S/Final Profilm Bofore Drainage

SEGMENT BOUNDARY COORDINATES

5 SURFACE boundary sogmonts

Segment x-left y-loft x-right NO. (ft) (ft) (it)

1 .O 29.0 20.0 2 20.0 52.0 191.0 3 191.0 86.0 305.0 4 305.0 100.0 323.0 5 323.0 142.0 420.0

7 SUBSURFACE boundary segmants

Segment x-loft y-left x-right NO. (ft) (it) (ft)

1 . O 29.0 59.0 2 59.0 40.0 175.0 3 175.0 60.0 310.0 4 310.0 85.0 315.0 5 315.0 90.0 400.0 6 .O 24.0 310.0 7 310.0 80.0 400*0

y-right (it)

52.0 86.0 100.0 142.0 170.0

y-right (ft)

40.0 60.0 85.0 90.0 110.0 80.0 105.0

soil Unit Below Segment

Soil Unit Below Segment

Figure 6.1.10.-XSTABL profile and outputfile offinal profile before drainage (continued)

954 Appendix 6.1

ISOTROPIC Soil Parameters

4 type(s) of soil

Soil Unit Weight Cohesion Friction Pore Pressure Water Unit Moist Sat. Intercept Angle Parameter Constant Surface NO. (pcf) (per) ( P W (des) RU (psi) NO.

1 120.0 130.0 250.0 30.00 2 115.0 125.0 30.0 28.00 3 104.0 110.0 .O 17.00 4 120.0 130.0 9999.0 75.00

1 Water surface(s) have been spe'cified

Unit weight of water = 62.40 (pcf)

Water Surface No. 1 specified by 7 coordinate points

.................................. PHREATIC SURFACE, ..................................

Point x-water y-water NO. (ft) (it)

A critical failure surface searchinu method. ueinu a random technique for generating sliding BL~CK surfaces, Gas been specified.

100 trial surfaces will be generated and analyzed.

2 boxes specified for generation of central block base

DEFAULT SEGMENT LENGTH SELECTED BY XSTABL Length of line segments for active and passive portions of slidina block is 43.0 it - ~ - -

Box x-loft y-left x-right y-right width no. (it) (ft) ( ft) (ft) (ft)

Figure 6.1.10.-XSTABL profile and oufputfile offinal profile before drainage (continued).

Appendix 6.1 955

Factors of safety have bean calculated by the :

4 * 4 MODIFIED JANBU METHOD 4 + 4 4

The following is a 8ummary of the TEN most critical surfaces

Problem Description : CAXP S/Final Profi

Modified Correction Initial JANBU FOS Factor x-coord

( ft)

.la Bmfore Drainage

Terminal Driving x-coord Force

( f t) (lb)

+ 4 END OF FILE +

Figure 6.1.10.-XSTABL profile and output file offinal profile before drainage (continued).

956 Appendix 6.1

CAMP 5/Final Profile w/5 F t Drawdown 10 most critical surfaces, MINIMUM JANBU FOS = 1.092

0 I I I I I I I I I

0 55 110 165 220 275 330 385 440

X-AXIS (feet)

Figure 6.1.11.-XSTABL profile and output file of final profile with 5-foot drawdown.

XSTABL File: C53FINAL 4-02-93 10:17

.......................................... XSTABL

Slope Stability Analysis using Simplified BISHOP or JAHBU methodm

Copyright ( C ) 1992 Interactive Software Designs, Inc.

All Rights Reserved

* Forest Service US Dept. of Agriculture

Moscow, ID 83843 *

* Ver. 4.10 1001 ..........................................

Problem Description : CAMP 5/Final Profile w/5 Ft Drawdown

SEGMENT BOUNDARY COORDINATES

5 SVRFACE boundary segments

Segment x-left y-left x-right NO. (ft) (it) (ft)

7 SUBSURFACE boundary segments

Segment x-left y-left x-right NO. (It) (ft) (it)

y-right (ft)

52.0 86.0 100.0 142.0 170.0

y-right (it)

40.0 60.0 85.0 90.0

110.0 80.0

105.0

Soil Unit Below Segment

Soil Unit Below Segment

3 3 3 3 3 4 4

Figure 6.1.11.-XSTABL profile and output fi le offinal profile with 5-foot drawdown (continued).

958 Appendix 6.1

ISOTROPIC Soil Parameters

4 type(s) of soil

Soil Unit Weight Cohosion Friction Pore Pressure Water U ~ i t Moist Sat. Intorcopt Anglo ParamOtOr Constant: Surface No. (pcf) (Pcf) (PSf) (dog Ru (psi) No.

1 Water surface(s) havo been spocifiod

Unit weight of water = 62.40 (pcf)

Water surface No. I spocifiod by 7 coordinate points

.................................. PHREATIC SURFACE, ..................................

Point x-water y-water NO. (ft) (ft)

A critical failure suriaco soarching mothod, using a random tochniquo for gonorating sliding BLOCX suriaco8. ha8 boon spocifiod.

100 trial surfaces will bo goneratad and analyzed.

2 boxes Specified for gonoration of contra1 block bas.

Figure 6.1.II.-XSTABL profile and outputfile offinal profile with 5-foot drawdown (continued)

Appendix 6.1 959

Length of line segments for active and passive portions of sliding block is 43.0 it

Box x-left y-left x-right y-right Width no. (ft) (ft) (ft) (ft) (ft)

Factors of safety have been calculated by the :

* * MODIFIED JANBU METHOD

The following is a summary of the TEN most critical surfaces

Problem Description : CAMP 5/Final Profile w/5 Ft Drawdown

Modified Correction Initial Terminal Driving JANBU FOS Factor x-coord X-coord Force

( ft) (ft) (lb)

* END OF FILE

Figure 6.1.1 I.-XSTABL profile and outputfile offinal profile with 5-foot drawdown (continued).

960 Appendix 6.1

CAMP 5/Final Prof i l e / l4 F t Drawdown 10 most critical surfaces, MINIMUM JANBU FOS =

55 110 165 220 275 330 385 440

X-AXIS (feet)

Figure 6.1.12.-XSTABL profile and output file offinal profile wirh 14-foot drawdown

XSTABL File: C54FINA.L 4-02-93 10:22

..........................................

XSTABL 4

1) slope Stability Analysis using * Simplified BISHOP or JANBU methods

* Copyright (C) 1992 Intaractiv. Softwar. Designs, Inc.

A11 Rights Reserved 4 4

I Forest Service * US Dept. of Agriculture * Moscow, ID 83843 4

4 Ver. 4.10 1001 * * * 4 * * * 4 4 * * * * * * * * * * * * * * * * * * * * * * * * * * * * * * * * *

Problem Description : CAMP 5/Final Profile/14 Ft Drawdown

SEGMENT BOUNDARY COORDINATES

5 SURFACE boundary segments

Segment x-left y-loft x-right NO. (ft) ( ft (it)

7 SUBSURFACE boundary sogmenta

Segment x-left y-loft x-right NO. (ft) (it) (it)

y-right (ft)

52.0 86.0 100.0 142.0 170.0

y-right (ft)

40.0 60.0 85.0 90.0 110.0 80.0 105.0

Soil Unit Below Segment

2 2 2

Soil Unit Below Segment

3

Figure 6.1.12.-XSTABL profile and outputfile offinal profile with 14-foot drawdown (continued).

962 Appendix 6.1

ISOTROPIC Soil Parameters

4 type(s) of soil

Soil Unit Waight Cohasion Friction Pore Pressure Water Unit Moist Sat. Intarcapt Angla Parametar Constant Surface No. (pcf) (pcf) (Psf) (d.9) Ru (Psf) No.

1 Water surface(s) have bean spacifiad

Unit weight of water = 62.40 (pcf)

Water surface No. 1 specifiad by 7 coordinata points

.................................. PHREATIC SURFACE, ..................................

Point x-water y-watar NO. (it) (ft)

A critical failurm surfaca searching mathod, using a random techniqua for gonarating sliding BLOCK surfaces, has bean spocifiod.

100 trial surfacas will be gonoratad and analyzad.

2 boxes specifiad for generation of central block bas@

* DEFAUL+T SEGMENT LENGTH SELECTED BY XSTABL

Figure 6.1.12.-XSTABL profile and output fi le offinal profile with 14lfoot drawdown (continued).

Appendix 6.1 963

Length of line segments for active and passive portions of sliding block is 4 3 . 0 ft

BOX x-left y-left x-right y-right Width no. (ft) (ft) (ft) (ft) (ft)

Factors of safety have been calculated by the :

f MODIFITD JANBU METHOD

The following is a summary of the TEN most critical surfaces

Problem Description : CAMP 5/Final Profile/l4 Ft Drawdown

Modified Correction Initial Terminal Driving JANBU FOS Factor x-coord x-coord Force

(ft) (ft) (lb)

END OF FILE

Figure 6.1.12,-XSTABL profile and output file of final profile with 14-foot drawdown (continued)

964 Appendix 6.1

CAMPS/TO~ Excavated/No Drawdown 10 most critical surfaces, MINIMUM JANBU FOS = ,836

0 I I I I I I I I 1

0 55 110 165 220 275 330 385 440

X-AXIS (feet)

Figure 6.1.13.-XSTABL profile and ourpurfile for toe excavated with no drawdown.

XSTABL File: CSSFINAL 4-02-93 10:26

*******t****t*t***+**t*******t***t*tt*****

t XSTABL * t

t Slope Stability Analysis using t

Simplified BISHOP or JANBU methodm t

t Copyright (C) 1992 Interactive Software Designs, Inc.

t All Rights Reserved t

t

t Forest Service t

t US Dept. of Agriculture t

t Moscow, ID 83843 t

t t

* Ver. 4.10 1001 * **+**+t***t**+*4++****+***********.1.**

Problem Description : CAMP5/Toe Excavated/No Drawdown

SEGMENT BOUNDARY COORDINATEs

8 SURFACE boundary segments

Segment x-left y-left x-right NO. (ft) (ft) (ft)

S SUBSURFACE boundary sopants

Segment x-left y-ldt x-r ight NO. (it) (ft) (it)

y-right (ft)

24.0 34.0 40.0 60.0 86.0 100.0 142.0 170.0

y-right (ft)

60.0 85.0 90.0 110.0 98.0

Soil Unit Below Segment

Soil Unit B o l w Segment

Figure 6.1.13.-XSTABL profile and outputfile for toe excavated with no drawdown (continued).

966 Appendix 6.1

ISOTROPIC Soil Parameters

4 type(s) of soil

soil Unit Weight Unit Moist sat. No. (pcf) (pcf)

Cohesion Friction Poro Pressuro Intorcopt Anglo Parametor Constant

(psi) (d.9) Ru (Psi)

1 Water surface(s) have been spocifie,

Unit weight of water = 62.40 (pcf)

Water Surfaco No. 1 specified by 5 coordinate points

.................................. PHREATIC SURFACE, ..................................

Point x-water y-water No. (ft) (it)

A critical failure surface searching method, using a random technique for generating sliding BLOCK surfaces, has been specified.

100 trial surfaces will bo genuatad and analyzed.

2 boxes epacifiod for goneration of c.ntr.1 block base

DEFAULT SEGUENT LENGTX S-CTED BY XSTABL

Water Surf ace

NO.

1 1 1 1

Figure 6.1.13.-XSTABL profile and output fi le for toe excavated with no drawdown (continued)

Appendix 6.1 967

Length of line segments for active and passive portions of sliding block is 44.0 ft

BOX x-left y-left x-right y-right Width no. (ft) (it) (ft) (ft) (ft)

Factors of safety have been calculated by the :

+ * + + MODIFIED JANBU METHOD + + +

I The following is a summary of the TEN most critical surfaces I I Problem Dascription : CAMP5/Toe Excavated/No Drawdown I

Modified JANBU FOS

Correction Factor

Initial x-coord

( ft)

81.46 85.23 84.02 90.11 81.18 103.94 93.86 92.40 86.71 85.76

Terminal x-coord

(it)

369.13 371.79 385.18 387.98 382.55 381.79 403.60 410.70 356.29 370.28

Driving Force (lb)

2.843E+05 2.64lE+O5 2.821E+05 2.871E+05 2.881E+05 2.8613+05 3.l96E+O5 3.447E+05 2.364E+O5 2. 5OOE+O5

Figure 6.1.13.-XSTABL profile and outputfile for toe excavated with no drawdown (continued).

968 Appendix 6.1

CAMPS/Toe Excavated/ 14 Ft. Drawdown 10 most critical surfaces, MINIMUM JANBU FOS = 1.024

55 110 165 220 275 330 385 440

X-AXIS (feet)

Figure 6.1.14.-XSTABL profile and output file for toe excavated with 14-foot drawdown.

XSTABL File: C56FINAL 4-02-93 11:19

.......................................... XSTABL t

* Slope Stability Analysis using * Simplified BISHOP or JANBU methods

a Copyright (C) 1992 t

Interactive Software Designs, Inc. All Rights Rmserved *

Forest Service t

t US Dept. of Awicultura t Moscow, ID 83843 1,

t

+ Ver. 4.10 1001 ..........................................

Problem Description : CAMPS/Toe Excavated/l4 Ft. Drawdown

SEGMENT BOUNDARY COORDINATES

8 SURFACE boundary segments

Segment x-left y-left x-right NO. (ft) (ft) (ft)

5 SUBSURFACE boundary segments

Segment x-left y-left x-right NO. (ft) (ft) (it)

y-right (ft)

24.0 34.0 40.0 60.0 86.0 100.0 142.0 170.0

y-right (it)

60.0 85.0 90.0 110.0 98.0

Soil Unit Below Segment

Soil Unit Below Sagrent

Figure 6.1.14.-XSTABL profile and outputfile for toe excavated with 14-foot drawdown (continued).

970 Appendix 6.1

rSOTROPIC Soil Parameters

4 type(s) of soil

soil Unit Weight Cohesion Friction Pore Pressure Water Jnit Moist Sat. Intercept Angle Parameter Constant Surface No- (pcf) (pet) ( P W (deg) Ru (Psf NO.

1 Water surface(s) have beon spocifiod

Snit weight of Water - 62.40 (pcf)

iater Surface No. 1 specified by 5 coordinate points

Point x-water y-water NO. (it) (ft)

4 critical tailuro surfaco soarching method, using a random :echniquo for gonorating sliding BLOCX surfaces, has been sp0citi.d.

100 t r i a l surfaces will k goneratod and analyzed.

2 boxes spocifiod for goneration of contra1 block base

Figure 6.1. Id-XSTABL profile and outputfile for toe excavated with 14-foot drawdown (continued).

Appendix 6.1 971

Length of line segments for active and passive portions of sliding block is 44.0 It

BOX X-left y-left x-right y-right Width no. (ft) (ft) (It) (ft) (ft)

Factors of safety have been calculated by the :

+ + + MODIFIED JANBU METHOD + + + *

The following is a summary of the TEN most critical surfaces

Problem Description : CAKPS/Toe Excavated/lQ Ft. Drawdown

Modified Correction Initial Terminal Driving JWBU FOS Factor x-coord x-coord Force

(it) (ft) (lb)

* + END OF FILE +

I I

Figure 6.1.14.-XSTABL profile and output file for toe excavated with 14-foot drawdown (continued).

972 Appendix 6.1

It= 10 most critical surfaces, MINIMUM JANBU FOS = 1.202

0

0 55 110 165 220 275 330 385 440

X-AXIS (feet)

Figure 6.1.15.-XSTABL profile and output file for toe excavated with 14-foot drawdown and bunress. 0 .l W

XSTABL File: C57FINAL 4-05-93 11:59

............................

XSTABL

Slope Stability Analysis using Simplified BISHOP or JANBU methods

* Copyright (C) 1992

Interactive Software Designs, Inc. ~ l l Rights Reserved *

* Forest Service

US Dept. of Agriculture Moscow, ID 83883

* Ver. 4.10 1001 * * * * * * * * * * * * * * * * * * * * * * * * * * * * * * * * *a * * * * * * * *

Problem Description : CAMPS/Toa Exc./l4 Ft.Drvdwn/Buttress

SEGMENT BOUNDARY COORDINATES

I 13 SURFACE boundary segmonts I I Segment x-left y-left x-right y-right Soil Unlt

No. (ft) (ft) (it) (it) Below Segment

1 .O 24.0 40.0 24.0 4 2 40.0 24.0 65.0 60.0 5 3 65.0 60.0 100.0 65.0 5 4 100.0 65.0 75.0 40.0 2 5 75.0 40.0 71.0 34.0 3 6 71.0 34.0 60.0 24.0 4 7 60.0 24.0 40.0 24.0 4 8 40.0 24.0 65.0 60.0 5 9 65.0 60.0 100.0 65.0 5 10 100.0 65.0 191.0 86.0 2 11 191.0 86.0 305.0 100.0 2 12 305.0 100.0 323.0 142.0 1 13 323.0 142.0 420.0 170.0 1 5 SUBSURFACE boundary segments

Segment x-left y-left x-right y-right Soil Unit No. (It) (it) (it) (st) Below Segment

1 75.0 40.0 175.0 60.0 3 2 175.0 60.0 310.0 85.0 3 3 310.0 85.0 315.0 90.0 3 4 315.0 90.0 400.0 110.0 3 S 71.0 34.0 400.0 98.0 4

Figure 6 1.15.-XSTABL profile and output file for toe excavated with 14-foot drawdown and buttress (continued) -- -

974 Appendix 6.1

ISOTROPIC Soil Parameters

5 type(.) of soil

Soil Unit Weight Cohosion Friction Pore Pressure Unit Moist Sat. Intorcapt Anglo Parameter Constant No- (pcf) ( P C ~ ) ( P S ~ ) (dog) RU (Psf)

1 Wator surfaco(s) hava boon spocifiod

Unit weight of water = 62.40 (pcf)

War-r Surfaco No. 1 specified by 5 coordinato points

...............................

PHREATIC SURFACE, * * * * * * * * * * * * * * * * * * * * * * * Z * * * * * * * * * *

Point x-water y-water NO. (ft) (ft)

A critical f8ilute surface searching rathob, using a random technique for generating sliding BLOCX iurfaces, has h e n specified.

loo trial surfaces will k genuatod and uralyzod.

2 boxes spacified for generation of central block base

Water Surf ace

NO.

1 1 1 1 1

Figure 6.1.15.-XSTABL profile and outputfile for toe excavated with 14-fool drawdown and buttress (continued).

Appendix 6.1 975

DEFAULT SEGMENT LENGTH SELECTED BY XSTABL + +

Length of line segments for active and passive portions of sliding block is 44.0 it

Box x-left y-left x-right y-right Width no. (ft) (it) ( ft) (ft) (ft)

Factors of safety have been calculated by the :

MODIFIED JANBU METHOD + + +

The following is a summary of tho TEN most critical surfaces

Problem Description : CAMPS/Toa Exc./l4 Ft.Drvdwn/Buttross

Modif ied JANBU FOS

Correction Factor

Initial x-coord

(ft)

96.63 95.26 112.63 87.21 108.84 96.22 106.86 94.28 107.15 116.83

Terminal x-coord

(ft)

402.44 377.78 382.67 380.92 363.73 361.15 379.60 383.47 352.40 344.91

* END OF FILE * +

Driving Force (lb)

Figure 6.1.15.-XSTABL profile and outputfile for foe excavated with 14-foot drawdown and buttress (continued)

976 Appendix 6.1

Soil Unit C(3): sandy silt (MH); greenish yellow; stiff; moist; medium dry strength; low toughness. Origin: completely decomposed pyroclastic rock.

& =17" C = 0 pcf

y,,,, = 110 pcf y,, = 104 pcf

Liquid Limit = 94 Plastic Index = 45

Stabilization Re-establishing a two lane road width, by providing a long-term stability correction

Alternatives measure, was essential due to the projected high volume of timber harvest from the area accessed by the road system. Several alternatives were considered, and the associated risks and costs were determined. Table 6.1.1 outlines those alternatives. The lower cost of a horizontal drainage design supplemented with a small buttress4ompared to a 17,000 cubic yard buttress-ffset a slightly higher risk.

Horizontal Drains

Table 6, I . I-Summary of stabilization alternatives.

r Alternative Relative Risk Factor of Safety Cost

17,000 cubic yard toe buttress I I

I I I

In order to proceed with a horizontal drain design, a test drain was drilled and installed to ensure that no major problems would be encountered in the material because of oversize rock fragments. Location of the test drain was crucial to a successful demonstration. The ground water table was contoured using water levels in the observation wells to determine equipotential and flow line direction.

Cutoff drain above slide I Low to Medium

Horizontal drain system with 2,000 cubic yard buttress

On June 3, 1983, a drain 175 feet long, inclined at 2 percent, was installed. A continuous 2-gpm discharge was measured after the installation. Total construction time, including mobilization at the site, was 4 hours, with no problems encountered. Because of the success of the test drain and the demonstrated problems which would be encountered with the deep trench construction, the horizontal drain system with a 2,000-cubic-yard buttress was chosen as the favored alternative.

1.35 1 $190,000

Appendix 6.1

Medium 1.20

System Design To determine the volume of water entering the slide area and to design an effective drainage system using Darcy's Law and Manning's Equation, the following informa- tion had to be obtained:

cross-sectional area of the aquifer

hydraulic gradient

hydraulic conductivity

Section 4E.1.4 gives an explanation of Darcy's Law.

Cross-Sectional The cross-sectional area of the aquifer, A, was determined by drilling exploration, Area of the Aquifer resistivity profiling, and ground water mapping. The aquifer was determined to have

a concave lower surface consisting of an impermeable clayey silt (MH) of decomposed lapilli tuff (rock unit 12). Two soil units overlying the lower surface were defined, which represented at least two episodes of slope collapse into the ancient drainage. Soil unit E, consists of a talus deposit with basalt fragments ranging in size from 1 cubic foot to 1 cubic yard, containing up to 45 percent open voids between the fragments. The void spaces were evident in the drilling operation as a sudden drop of the drill casing and a corresponding loss of water return. Core loss zones were observed and recorded in these areas. The void spaces may be result of piping. Soil unit E, surrounds soil unit E, and consists of a colluvial slope deposit of basalt fragments of 1 cubic foot to 1 cubic yard in a matrix of dense silty sand (see figure 6.1.16). The cross-sectional areas of these units were measured with a polar planimeter.

Figure 6.1.16,-lateral cross-section along Forest Service Road 1926639 showing aquifer configuration and material boundaries.

Hydraulic Gradient Hydraulic gradient, i, was computed from the static water level in observation wells DH-10, DH-11, and DH-12, and at the spring at the head scarp, as sin-' (MU) = 13" or 23 percent This is comparable to the average gradient of the impermeable layer (rock unit 12).

Appendix 6.1

Hydraulic The hydraulic conductivity, K, was determined by in-situ maintained and falling head Conductivity tests (US. Department of the Navy, 1982b).

Discharge Rate These values were used in Darcy's Law (Q = KiA) to determine discharge rate as summarized in table 6.1.2.

Table 61.2.--Summary of determination of discharge

Soil Unit

The total discharge of 149 gpm was approximately the 130 gpm measured from all scarp and toe springs.

E ,

Number of Drains Manning's Equation was used to calculate the capacity of a 1-112-inch ID slotted Needed PVC pipe flowing full in various configurations as

Q = VA

Cross- Sectional Area (ft2)

where:

11,888

Q is discharge rate V is velocity in fthec A is end area of pipe (0.012 ft2) R is hydraulic radius (0.031 ft) S is percent slope, and N is roughness coefficient (0.013).

Hydraulic Gradient

Q Total

Table 6.1.3 outlines these results, disregarding inlet control as with a solid pipe.

0.23

Appendix 6.1

Hydraulic Conductivity

(fvday)

0.33

Discharge Rate

(cfs) (gpm)

5

149

0.16 71.1

Table 6.1.3.-Capocity of 1%-in. ID slotted pipe flowing full at various configurations.

v I Q = VA Velocity Discharge at 100 percent capacity I

Drain gradients and number of drains required were then determined based upon cross-section geometry and at 50 percent flow capacity to compensate for low perfor- mance. This configuration had to accomodate approximately 150 gpm based upon the predicted aquifer flows calculated above using Darcy's Law. Table 6.1.4 outlines these results.

Table 6.1.4.-Drain gradient and number of drains required (assuming 50 percentflow capacity).

Discharge at 50 percent Number of capacity per drain I drains I Total discharge I

With this configuration, and all drains at 50 percent capacity, 23 drains would have been 100 percent effective. Based on further literature search (Smith, 1980), it was decided that 25 percent capacity would be used for design purposes, which increased the number of drains to 46 and which could have a maximum 50-percent-capacity of 250 to 350 gpm, depending upon configuration.

(percent)

2

Drain locations were planned according to a horizontal flow net to intercept equipotential lines. Areas which indicated a piezometric "valley" were identified to contain concentrations of drains. A total of 26 drains was planned as interceptor drains above the failed mass, and 16 drains were provided within the failed mass as relief drains. Drain lengths and slopes were based on cross-section geometry and were planned to be embedded no more than 10 feet into the impermeable contact to ensure maximum length into soil units E, and E, and minimum length in rock unit 12. This was done to prevent migration of silt and clay sized particles into the drains.

Appendix 6.1

(cfs) (gpm)

,010 4.5 10

(cfs) (gpm)

0.10 45

Slot Width Slot width configuration was based upon soil unit E, (soil type SM silty sand with rock fragments), and a slot width of 0.05 inches was chosen to satisfy Cedergren (1977):

< 1 . 2 Slot width

A collector system using 12-inch corrugated metal pipe and 8-inch cormgated polyethylene pipe, and anchored above the ground with steel posts and V4-inch wire rope was designed. An above-ground location was chosen so that in the event of further movement separations could be located quickly. Forces at bends were calcu- lated to determine weather thrust blocks would be necessary. The maximum force calculated was 7 pounds, which only required double anchoring.

Effective End Spacing and Drawdown

Effective end spacing and drawdown were calculated using Prellwitz's (1979) method and HP41 program " G W (Prellwitz, 1990). The final profile developed for stability analysis before drainage (figure 6.1.10) was used to generate values for various drain configurations. The X and Y axes were shifted to place the origin at the toe of the slope. The following output files (figure 6.1.17) show the calculated minimum and maximum spacing and the various Y-axis coordinates of the phreatic surfaces undrained (Y,), drained in the same section as the drain (YJ, and drained in a section midway between drains (Y,). The drain slope for this trial was set at 7'. The program calculated a drainage barrier intercept length of 168.1 feet with a minimum spacing of 8.1 feet and a maximum spacing of 65.1 feet. The program was developed for use with parallel drains. When " G W is used with a fan array, the following procedure is useful:

(1) Plot the drains in section and plan view.

(2) Use "GW" to calculate the various Y-values for Y,,, in the minimum and maximum drain spacing intervals.

(3) In plan view, measure the X-value along a set of drains for the various spacing intervals that were calculated.

(4) For each of these X-values, locate the Y-value in the corresponding output column that matches the spacing at that point.

(5) Plot the X-Y coordinate on the cross-section for each X-interval spacing position measured in plan. This will develop a curve of the estimated drained phreatic surface midway between drains for a fan array. The drawdown can be measured on the section at any point or can be calculated as the difference between Y, and Y, in the output file (figure 6.1.17).

Appendix 6.1

SL : D R ~ N SLOPL q:

I.D.:CRIIP 5 200 220 3 00 I du= l4.811. I I I

aloha= 17.61.: 18.8deq. PHR. SURF. D

PFIRFILLEL DRN

PHR. WRF. U h ~ = 1 e . 0 1 ~

b d a = 1.19:1= 48.8der. XB Y8 XDI

-1.8 2.8 216.5 XBD YBD XlD

DRAIN M SL. Hd 1.5 7.8 5.8

Xdb Ldd Ldc 168.1 169.1 162.3

DRAIN AT: X-168.1. Yz25.6

SPRCIHC HIH S.8.1 HAX 3.65.1

Figure 6.1.17.-Profjle used for program GW.

Appendix 6.1

Y, Slopes For Longer Drains

X = 198.8 YI=43.5

w m n 43.3 4 . 2 3 . 8

m = m 5 X = 188.8

Y1=41.7 YU m n

41.4 37.6 34.7 Wi27.7

X = 178.8 Yl=48,8

YU YM n 39.6 34.3 30.8

W=26.8

47.i 45.1 43.4 W=D.8

X = 150.8 YI-43.5

w m YD 43.3 48.8 D . 8

w=F1.5 X * 188.8 X = 1n.n

YI=4.9 w va rm

40.5 16.8 32.6 m=26.9

X = 160.1 X 163.1

Y1=38.8 w m n

D.3 32.8 25.8 W=24.8

Figure 6.1.17.-Profile used for program GW (continued).

Appendix 6.1

C X = 60 S = 18.0 X - 424.9

YI=84.9 w YII n

81.9 84.9 84.9 W170.9

Y : 3W.8 YIi62.9

YU YII YD 62.9 61.8 61.8

m=18.9 X = 248.8

y1=52.3 w Y1I YD

52.3 50.4 58.4 *18. 3

X = 218.8 YI47.8

YU m YD 47.8 43.5 43.4

YB*n.a x = I%.@

Y1.4l.S w m n

43.3 30.1 18.8 18=29.s

Y = 188.8 Y141.7

YU m va 41.4 35.8 34.7

18=27.7 X = 170.8

Yl=48.8 YU YII YO

39.6 30.5 3 . 8 W-26.8

X = 168.1 YI=39.6

w 111 YD 39.2 29.6 28.4

W 2 5 . 6 X = 163.1

rl=38.e YU Y YD

18.3 26.8 25.8 W=24.8

X = 153.1 Yl=37.8

YU YII YO 36.4 23.8 23.8

Y8=23.8 Y = 138.1

Y1=34.3

S - 1 0 O N L Y L O C A L ( N F L U E N C E

X D I S T A N C E - 112 W A Y TO X - 1 9 0 - E N D O F S - 20 I N F L U E N C E

Yl=36.i YU YII YD

36.1 25.1 23.6 YB=22. i

X = 134.3 Y1=33.i

W YII YD IL.9 21.7 21.5

YB=19.7 X = 112.4

YI.29.8 W m YD . 7 18.8

're:::.

F O R X < 1 7 0 Y M IS B E L O W D R A I N

Figure 6.1.17.-Profile used for program GW (continued).

Appendix 6.1

Con~truction A contract was awarded in November 1983 to install approximately 8,000 linear feet of 1-112-inch ID schedule 80, type I1 PVC slotted horizontal drain pipe. Construction began by installing the surface-mounted 12-inch and 8-inch collector pipes, which would be connected to the individual drains by a 2-inch ID wire- reinforced extension hose. A ditch was excavated around the perimeter of the failed mass to channel surface drainage, and a wood fiber excelsior erosion control mat was placed on the surface prior to seeding. Drilling pads were excavated; however, a notice to proceed with installation was not given until December 1, 1983, to ensure that enough rainfall had occurred to recharge the aquifer to allow field judgments for modifications on advantageous drain locations, concentrations, and direction, based upon observed discharge from completed drains.

Installation was accomplished by advancing the drill casing to the desired length with a knockoff, tri-cone roller bit. At that point, the casing was rotated in the reverse direction and the slotted PVC pipe was inserted, thereby knocking off the bit. As the casing was removed, the PVC pipe and roller bit remained in the drill hole. For protection, and to prevent root growth in the drain, a 3-inch galvanized metal pipe was inserted over the discharge end to at least 5 feet and then grouted in place, and the extension hose was connected to the collector pipe. The elevation was measured prior to casing removal by filling the casing with water and elevating a hose connected to the "kelly rod," at the end of the drill casing, until static equilibrium was reached. This height above the discharge end was equal to the head in the drain, and the end elevation and final slope could be calculated and recorded quickly. The system was installed within 30 days at a contract cost of $107,000. The 2,000-cubic- yard toe buttress cost an additional $43,000.

Post- After installation, discharge from the entire system was measured at 250 gpm.

Con~truction During the second week of February 1984, after 4 112 inches of rainfall in 24 hours.

Monitoring total system discharge reached 500 gpm. Figure 6.1.18 displays the relationships among rainfall, ground water levels, previous failures, and system discharge. The interceptor drain system above the failure discharged over 300 gpm, with many drains flowing over 25 gpm. Static water level decreased in the failed mass from a level of 14 feet below surface to a level of 28 feet below surface.

Subsequent monitoring of the failure included installation of pressure transducers in observation wells, which transmit static water levels to a surface voltage recorder (Prellwitz and Babbitt, 1984). A Parshall flume with pressure transducer and voltage recorder was also installed in the 12-inch main collector to record system discharge. Total system discharge averaged 200 gpm during the winters of 1984-85 and 1985-86, with a yearly average of 100 million gallons. A re-survey of the slide area in July 1984 indicated a vertical displacement of only 0.2 feet, which is thought to be the result of settlement due to dewatering. No horizontal displacement of the failed mass was measured. Monitoring has continued through 1993, with no changes occurring.

Appendix 6.1

Conclusions

Figure 6.1.18.4ystern performance showing rainfall (lines) in 198143 and 198344, and total system discharge (dots) in 198445.

Empirical design methods used in the course of this project were based upon theory, experience, and engineering judgment. High precision was not expected, and any achieved may have been coincidental; however, satisfactory results were obtained. Darcy's Law was found to be effective, even in soils of an anisotropic, heterogeneous nature. Based on maximum discharge from the interceptor drains above the failed mass (300 gpm) and the toe relief drains (200 gpm), a back- calculation indicated an average hydraulic conductivity (K) of approximately 28 ft/day, which is slightly higher than the maximum average K of 21 ft/day predicted. This may indicate that the aquifer end area is larger than predicted, or soil unit E, is more extensive than is shown on the cross-sections. However, by using a design flow of 25 percent capacity (150 gpm), the system was able to accomodate the increase. Observations of the discharge rate of individual drains at their final slope revealed that a roughness coefficient of 0.009 is more accurate than 0.013 for PVC pipe slotted on two 120' centers and that Manning's Equation is effective for predicting drain capacity.

Appendix 6.1

References Cedergren, H.R. 1977. Seepage, drainage andflow nets, 2nd ed. New York, NY: John Wiley and Sons. 324-329.

Long, M.T. 1986. Camp five slide--exploration, design and construction of a horizontal drain solution. In: Proc. of the 22nd Annual Engineering Geology and Soils Engineering Symposium: Boise State Univ, Idaho State Univ., and University of Idaho. 24-26 February 1986: Boise, ID. Boise ID: Idaho Department of Transportation. 246264.

Long, M.T. and M.A. Leverton. 1984. Pleistocene interglacial volcanism: upper Salmon Creek drainage, Lane County. Oregon Geology, 46(11). 130-139.

Peck, D.L.; A.B. Griggs; H.G. Schickler; F.G. Wells; and H.M. Dole. 1964. Geology of the central and northern parts of the western Cascades range in Oregon. U.S. Geological Survey Professional Paper 449. 56 p.

Prellwitz, R.W. 1990. "CW"4rained phreatic suflace analysis with the HP41 programmable calculator. Unpublished report. Moscow, ID: USDA Forest Service, Intermountain Research Station. 226 p.

Prellwitz, R.W. and R.E. Babbitt. 1984. Long-term groundwater monitoring in mountainous terrain. Trans. Res. Rec. 965. Washington, DC: Transportation Research Board. 265-281

Priest, G.R. and B.F. Vogt, eds. 1983. Geology and geothermal resources of the central Oregon Cascade range. Special paper 15. Portland, OR: Oregon Dept. of Geology and Mineral Industries. 185 p.

Smith, D.D. 1980. The effectiveness of horizontal drains. Report No. FHWA/CA.TL-8Cb16. Sacramento, CA: California Department of Transportation. 79 p.

Turner, G.K. & Associates. 1971. Fluorometry in studies of pollution and movement of Fluids. Fluorometry Reviews. Palo Alto CA. 1-1 1.

U.S. Department of the Navy. 1982. Soil mechanics. NAVFAC DM 7.1. Washington, DC: Department of the Navy, Naval Facilities Engineering Command, Alexandria, VA.

XSTABL Ver. 4.102. 1992. Moscow, ID: Interactive Software Designs, Inc., 953 N. Cleveland Street.

Appendix 6.1

APPENDIX 6.2

6.2 Shear Trench Sample Problem

Cliff Denning, Geotechnical Engineer, Mt. Hood National Forest

Background A sidehill embankment on Highway 47 has been active for at least 30 years. Initially, the roadway was maintained by placing AC overlays on the sunken grade (head scarp area). In recent years, the roadway has been shifted away from the slide head scarp, and a temporary guardrail has been installed at the new fog line. The original roadway included two travel lanes, one westbound climbing lane, and shoulders. The purpose of the project is to restore the original highway alignment and lane shoulder widths.

The formation exposed in the cutslopes consists predominantly of tuffaceous siltstonelmudstone. The soil unit within the slide area is of residual origin and is classified as silt and clayey silt. Springs and ground movement are evident at the toe of the slide, which is located about 20 feet upslope of a gravel-surfaced county road. Subsurface conditions were explored with three borings and two test pits. Steel inclinometer tubing was installed in each boring to measure the depth to the failure zone. Location of the borings and test pits are shown on the site plan, figure 6.2.1.

The surface soil consists of 5 to 27 feet of fill material that was placed during original construction. The fill material consists of blocky siltstone fragments in a light brown sandy silt matrix. Beneath the fill material is 3 to 7 feet of stiff tan-to-brown clayey silt with some siltstone fragments, wood fibers, and roots. A tan siltstone rock unit underlies the soil units. The siltstone rock is typically soft to medium-hard and is generally rippable. Testing results on samples obtained during the subsurface exploration are listed in table 6.2.1.

Appendix 6.2

L * Figure 6.2.1.-Site plan showing location of the borings and test pits (from ODOT, 1988). a

E s? c

Table 6.2.1.-Test results from subsurface exploration.

Test Pit 1

The basic geologic cross-section and material properties used in the analysis are shown in figure 6.2.2. Enough movement had occurred to anticipate that the friction angle along the failure plane was at its residual value. An XSTABL analysis was run to back-calculate the friction angle. The analysis cross-section consisted of the slide mass and high density fill overlying about 4 feet of the silt and clayey silt soil unit. Below all the soil units is a siltstone unit. A 5-foot-high water table is perched above the clayey silt soil unit. Inclinometer measurements indicate that the failure zone is located within the residual clayey-silt soil unit. Several XSTABL runs, using the specific noncircular option, were made varying the slide zone friction angle. Cohesion was set to zero. Factor of safety versus friction angle was plotted as shown in figure 6.2.3 and the residual angle of friction was estimated to be 23.5". which agrees with charts relating plasticity index to residual friction angles (section 4C.3.4). Figure 6.2.4 shows an XSTABL run with a friction angle of 23.5". resulting in a factor of safety of 1.007. This back-calculated condition was used to calculate the stability of a shear trench.

Source

Test Pit 2

TB-100

Appendix 6.2

USCS

Bag No.1

Sample No.

Bag No. 1

N-5

Natural Moisture Content

15ft

Atterberg Limits

L.L. P.L. Depth Desc.

18 ft

27.5-29 ft

Soil

Silt

Clayey Silt

Clayey Silt

ML

MH

MH

38%

60%

62%

38% 28%

73%

75%

38%

55%

~b ~oa- t h inis J 5w IWI,,~~. rs-loz

84' AHEAD

Figure 6.2.3.-Factor of safety versus friction angle. Residual friction angle was estimated to be 23.5".

Appendix 6.2

SHEAR TRENCH EX. 6 G - 1 PHIz23.5 DEG Janbu Factor of Safety for Specified Surface = 1.007

180 225 270 315 360 405 450 495 540 X-AXIS (feet)

Figure 6.2.4.-An XS7ABL run with a friction angle of XY', resulting in a furor of safety of 1.007. Ris back-analyzed condition war used to calculae the stability of a shear trench

PROFIL SHEAR TRENCH EX. 6G-1 PHIz23.5 DEG

204.00 231.00 328.00 388.00 408.00 472.00

SOIL 5 . 130.00 141.00 50.0 105.00 110.00 .O 140.00 145.00 3000.0 125.00 130.00 .O WATER

1 62.40

SURFACE 7

Figure 6.2.4.-An XSTABL run with a friction angle of 2 3 9 (continued).

Appendix 6.2

Shear Trench Figure 6.2.5 shows the profile of a portion of the slide with a shear trench.

Width XSTABL stability analyses were run with trench widths along the failure plane varying from 30 to 50 feet. The failure surface used in the analysis was the same as the surface used in finding the residual friction angle. Shear trench material was anticipated to be clean, durable rock with a maximum size of about 15 inches. The shear trench was anticipated to intercept the majority but not all of the ground water. A small ground water table perched above the failure plane along the lower part of the slope was used in the stability analysis. A minimum factor of safety of 1.2 was chosen for the site. A trench width of 40 feet resulted in a factor of safety greater than the 1.2 minimum required and was chosen for further analysis. Figure 6.2.6 shows the results of an XSTABL run with a trench width of 40 feet.

Figure 6.2.5.4 portion of the slide with a shear trench. XSTABL stabiliry analyses were run with trench widths, along the failure plane, varying from 30 to 50 feet.

998 Appendix 6.2

EX. 6G- 1 WITH TRENCH WIDTH = 40 FEET Janbu Factor of Safety for Specified Surface = 1.210

180 225 270 315 360 405 450 495 540

X-AXIS (feet)

Figure 6.2.6.-The results of an XSTABL run with a trench width of 40 feet. - 19 19 19

PKOIIL PASSIVE WEDGE FAILURE THRU TRENCH 42 13 180.00 284.00 204.00 284.00 216.00 296.00 231.00 296.00 231.10 296.50 250.00 324.00 322.00 361.00 332.00 362.00 413.00 388.00 449.00 415.00 483.00 415.00 518.00 415.00 520.00 414.00 332.00 362.00 410.00 384.00 410.00 384.00 464.00 412.00 474.00 412.00 407.00 380.00 407.00 380.00 438.00 388.00 472.00 388.50 389.00 352.50 408.00 360.50 231.00 296.50 328.00 328.50 389.00 352.50 408.00 360.50 472.00 388.50 354.00 334.00 388.20 351.00 388.10 347.00 204.00 284.00 231.00 292.00 328.00 324.00 388.10 347.00 408.00 356.00 472.00 384.00 354.00 334.00 368.00 334.00 373.00 341.00 388.00 341.00

SOIL

WATER 1 62.40

472.00 393.00 520.00 414.00 540.00 418.00

SURFACE .,

Figure 6.2.6.-The results of an XSTABL run with a trench width of 40 feet. (continued)

Appendix 6.2

Shear Trench The shear trench will be embedded at least 5 feet into the underlying siltstone. A

Depth stability analysis along the bottom of the trench within the siltstone results in a high factor of safety. During construction it is important that both the design embedment and the siltstone unit be field-verified.

Analysis with Failure with the passive wedge exiting through the shear trench was investigated with

Passive Wedge XSTABL using the block surface search analysis option. The Rankine method was

Exiting Through used in the analysis to define the active and passive portions of the failure surface. Two boxes were defined for the analysis. Box 1 was located to vary the location of

Shear Trench the passive portion of the failure surface as shown in figure 6.2.7. Box 2 was located to allow the active wedge to exit at various points along the road width. Figure 6.2.8 shows the results of the XSTABL analysis. As expected, the most critical surfaces (those with the lowest calculated factor of safety) have the shortest lengths within the high strength shear trench material.

Figure 6.2.7.-Two boxes were defined for the Rankine analysis. Box I was located to vary the location ofthe passive portion ofthe failure surface as shown. Box 2 was located to allow the active wedge to exit at various points along the road width.

Appendix 6.2 1001

PASSIVE WEDGE FAILURE THRU TRENCH 10 most critical surfaces, MINIMUM JANBU FOS = 1.789

200 1, 180 225 270 315 360 405 450 495 540

X-AXIS (feet)

Figure 6.2.8.-The results of the XSTABL analysis. As expecred, the most critical surfaces have the shonest lengths within the high strength shear wench material.

PROFIL PASSIVE WEDGE FAILURE THRU TRENCH

4 2 1 3

3 7 3 . 0 0 3 8 8 . 0 0

SOIL e

1 2 5 . 0 0 1 3 0 . 0 0 . O 4 0 . 0 0 , 0 0 0 WATER

1 6 2 . 4 0

Figure 6.2.8.-The results of the XSTABL analysis (continued).

Appendix 6.2 1003

Analysis with Failure with the passive wedge exiting behind the shear trench was not anticipated to

Passive Wedge be a problem because of the depth to the failure plane. However, analysis for failure

Exiting Behind behind the shear trench is similar to the analysis of failure through the shear trench. Two boxes were defined for the block surface search analysis option of XSTABL as

Shear Trench shown in figure 6.2.9. Box I was located to vary the point where the passive wedge exits, from just behind the shear trench to about 8 feet upslope from the shear trench. Box 2 was not changed from the previous analysis. Figure 6.2.10 shows the ten most critical surfaces. All ten exit points cluster around the back point of the shear trench and intercept box 1 near the downslope box boundary. This essentially defines the shortest path for the passive portion of the failure surface to exit the slide mass.

Figure 6.2.9.-Two boxes were defined for the block surface search analysis. Box I was located to vary the location of the passive wedge exits, from just behind the shear trench to about 8 feet upslope from the shear trench. Box 2 was not changed from the previous analysis.

Appendix 6.2

PASSIVE WEDGE FAILURE BEHIND TRENCH 10 most critical surfaces, MINIMUM JANBU FOS = 2.318

180 225 270 315 360 405 450 495 540

X-AXIS (feet)

Figure 6.2.10.-7he ten most critical surfaces. All ten exit points cluster around the back point of the shear trench and intercept box I near the downslope box oC boundary z

PASSIVE WEDGE FAILURE THRU TRENCH

4 1 0 . 0 0 3 8 4 . 0 0 4 6 4 . 0 0 4 1 2 . 0 0 4 7 4 . 0 0 4 1 2 . 0 0 4 0 7 . 0 0 3 8 0 . 0 0 4 0 7 . 0 0 3 8 0 . 0 0 438 .00 3 8 8 . 0 0 472 .00 3 8 8 . 5 0 3 8 9 . 0 0 3 5 2 . 5 0 408 .00 3 6 0 . 5 0 231 .00 2 9 6 . 5 0 328 .00 3 2 8 . 5 0 389 .00 3 5 2 . 5 0 4 0 8 . 0 0 3 6 0 . 5 0 4 7 2 . 0 0 3 8 8 . 5 0 3 5 4 . 0 0 3 3 4 . 0 0 3 8 8 . 2 0 3 5 1 . 0 0 3 8 8 . 1 0 3 4 7 . 0 0 2 0 4 . 0 0 2 8 4 . 0 0 2 3 1 . 0 0 2 9 2 . 0 0 3 2 8 . 0 0 3 2 4 . 0 0 3 8 8 . 1 0 347 .00 4 0 8 . 0 0 3 5 6 . 0 0 4 7 2 . 0 0 384 .00 3 5 4 . 0 0 334 .00 3 6 8 . 0 0 334 .00 3 7 3 . 0 0 3 4 1 . 0 0 3 8 8 . 0 0 3 4 1 . 0 0

S O I L 5 1 3 0 . 0 0 1 4 1 . 0 0 1 0 5 . 0 0 1 1 0 . 0 0 1 4 0 . 0 0 1 4 5 . 0 0 1 2 5 . 0 0 1 3 0 . 0 0 1 2 5 . 0 0 130 .00 WATER

1 6 2 . 4 0 1 4 1 8 0 . 0 0 2 7 9 . 0 0

Figure 6.2.10.-The ten most critical surfaces (continued).

Appendix 6.2

Analysis with Active Wedge Exiting Shear Trench

The block surface search option of XSTABL was used to analyze this condition. Three boxes were defined for the analysis. Box I and box 2 were point boxes located on the existing failure plane below the shear trench. Box 3 was located to vary the active portion of the failure surface within the shear trench as shown in figure 6.2.1 1. Figure 6.2.12 shows the ten most critical surfaces from the XSTABL analysis. All ten active wedges toe out near the downslope boundary of the shear trench. As expected, the failure surfaces with the least length within the high strength shear trench material generate the lowest factors of safety. As can be seen from the results, the factor of safety is 1.17, which is close to but below the specified minimum of 1.2. Judgment by the designer is required whether 1.17 is an acceptable or an unacceptable factor of safety.

ACTIVE WEDGE FAILURE IN SHEAR TRENCH

180 225 270 315 360 405 450 495 540

X-AXIS ( f ee t )

Figure 6.2.11.-Three boxes were defined for the block surface search analysis. Box I and box 2 were point boxes located on the existing failure plane below the shear trench. Box 3 was located to vary the active portion of the failure surface within the shear trench.

Appendix 6.2

ACTIVE WEDGE FAILURE IN SHEAR TRENCH 10 most critical surfaces, MINIMUM JANBU FOS = 1.173

180 225 270 315 360 405 450 495 540

X-AXIS (feet)

4 0 7 . 0 0 3 8 0 . 0 0 4 3 8 . 0 0 3 8 8 . 0 0 4 7 2 . 0 0 3 8 8 . 5 0 3 8 9 . 0 0 3 5 2 . 5 0 4 0 8 . 0 0 3 6 0 . 5 0 2 3 1 . 0 0 2 9 6 . 5 0 3 2 8 . 0 0 3 2 8 . 5 0 3 8 9 . 0 0 3 5 2 . 5 0 4 0 8 . 0 0 3 6 0 . 5 0 4 7 2 . 0 0 3 8 8 . 5 0 3 5 4 . 0 0 3 3 4 . 0 0 3 8 8 . 2 0 3 5 1 . 0 0 3 8 8 . 1 0 3 4 7 . 0 0 2 0 4 . 0 0 2 8 4 . 0 0 2 3 1 . 0 0 2 9 2 . 0 0 3 2 8 . 0 0 3 2 4 . 0 0 3 8 8 . 1 0 3 4 7 . 0 0 4 0 8 . 0 0 3 5 6 . 0 0 4 7 2 . 0 0 3 8 4 . 0 0 3 5 4 . 0 0 3 3 4 . 0 0 3 6 8 . 0 0 3 3 4 . 0 0 3 7 3 . 0 0 3 4 1 . 0 0 3 8 8 . 0 0 3 4 1 . 0 0

SOIL 5

WATER 1 6 2 . 4 0 14

Figure 6.2.12.-The ten most critical surfaces from the XSTABL analysis (continued).

Appendix 6.2

Stability Below the Shear Trench

General Comments

Reference

A shear trench's full stabilizing effect is on the failure mass located upslope from the trench. Marginal slopes located below the shear trench can be affected positively by the interception or lowering of the ground water. In this example, with tension cracks located within the lower part of the slope, the stability below the shear trench is in question. Results of an XSTABL analysis on the slope below the shear trench indicate that even with a general lowering of the ground water, a high potential for continuing slow movement of the lower slope exists. In many cases the main objective of the use of a shear trench is to stabilize a slope containing a road or a structure. Slow movement of a slope below the shear trench may have little or no consequence and be acceptable.

In this example a small road is located at the toe of the slope. Continuing movement of the lower part of the slope may affect maintenance costs of the lower road or possibly close the road for short times due to small backslope failures. The risk to the lower road may dictate an alternative action, such as a shear trench with buttress at the slide's toe. Comparing and evaluating alternatives by risk analysis is covered in section 61.

Figure 6.2.13 shows a recommended sequence for shear trench stage construction. The excavation should be backfilled with shear trench material by the end of each working day.

At a minimum, riprap geotextile should be placed along the backcut soil slope of the shear trench. The shear trench should be drained. In this case, a 6-inch diameter perforated pipe was used to collect the ground water. The collected water was directed away from the slide area through a nonperforated pipe.

Oregon Department of Transportation. 1988. Geotechnical Report, Staley Slide, Unit 1, Sunset Highway 47, MP 44.6, Washington Counry. Salem, OR: Oregon Department of Transportation.

Appendix 6.2

3. - BENCH CONSTRUGTION

Excavate to bench elevation

II- EXCAVATION FOLLOWED BY BACKFILLING

. E m v a t e 20' maxlrnum sectlo" ond bockfill with rock

- EXCAVATION FOLLOWED BY BACKFILLING IN STAfjES,TO CDMPLETQN IP. - COMPLETION OF UPPER PORTION OF TREdeH

. Cont~nue ercavatmq 20' scctions and bockfillin9 with rock Place rock from bench clewtion t o top of Wr T a W C *

EXCAVATION AND BACKFILL. SEQUENCE NOT TO SCALE

Figure 6.2.13.-A recommended sequence for shear trench stage corrrtruction (from ODOT, 1988). r

E -

References Oregon Department of Transportation. 1988. Geotechnical Report, Staley Slide, Unit I , Sunset Highway 47, MP 44.6, Washington County. Salem, OR: Oregon Department of Transportation.

Appendix 6.2

APPENDIX 6.3

6.3 Rock Slope Stabilization- System specifications Examples

Michael T. Long, Engineering Geologist, Willamette National Forest

This appendix contains example system specifications from the Willamette National Forest for scaling, rock bolts, rock dowels, slope containment mat, and shotcrete application.

DIVISION 2 SITE WORK SCALING 00020

General Description

Site work includes mechanical scaling of the rock slope; installation and testing of tensioned and grouted rock bolts; installing untensioned rock dowels; installing free hanging and slope reinforcing mats; installing an upslope drainage containment and diversion structure and anchoring the structure to the rock face; applying shotcrete; road grading and resurfacing; and removal of all debris generated during slope stabilization. Additional work items required, such as providing emergency access, protecting existing structures, etc., shall be considered incidental to the following items:

Scaling

Rock bolts, types 1 and 2

Rock dowels, types 1, 2, and 3

Rappel anchors

Soil anchors

Free hanging mat

Slope reinforcing mat

Shotcrete application

Upslope drainage and diversion structure

Road grading and surfacing

Concrete

Appendix 6.3

Execution

Work starts by scaling the entire project area. Upon completion of scaling, the contractor will cease work on the portion of the project requiring rock bolts and the exact location of the rock bolts will be located by the Forest Service. This will be followed by installation of rock bolts, tensioning, and shotcrete; construction of drainage and diversion structure; placement of rock dowels and slope containment mats; and reconditioning and resurfacing of the roadway. Specific details and incidental work are given in the specifications andlor shown on the drawings.

DIVISION 2 SITE WORK SCALING 02101

General Description

Rock slope scaling shall be conducted on all rock slopes as directed by the engineer (as shown on the drawings) and in accordance with these specifications and the contractor's work plan and shall include the removal of loose blocks of debris and vegetation from the slope by mechanical methods. Blasting or the use of ground surface power equipment, such as backhoes, etc., shall be prohibited except in construction zone 3 as shown on the drawings.

Crew Complement

The contractor shall provide a minimum of one qualified rock slope scaling crew that consists of a working foreperson and two scalers. The crew size shall be maintained at all times. Any crew member who must leave for any reason shall be replaced immediately by a qualified replacement. If the scaling activities have the potential of endangering adjacent structures (e.g., guardrail, retaining wall, signs) the contractor shall provide appropriate protective devices, as per the contractor's work plan, prior to commencing the scaling work.

Execution

Material Included

All slopes to be scaled are indicated on the drawings. Scaling of these slopes shall include the removal of all loose rock, whether in original position or not; all vegetation over 118 inch in diameter; and any existing wire mesh installed on the slope. Existing wire mesh and associated hardware shall become the property of the contractor.

Methods

Rock shall be scaled manually using a suitable 5-foot standard steel mine scaling rod supplemented when necessary by pneumatic splitters. Vegetation shall be pruned or sawed off even with the rock face. In construction zone 3 only. alternative methods such as blasting or use of ground surface power equipment may be. allowed at the discretion of the Forest Service. If blasting methods are to be used, this portion of the scaling work must precede any other scaling proceeds.

Rock slope scaling shall start at the top of the slope and work shall proceed downward towards road grade, removing all loose rock blocks as the work progresses. Rock blocks or debris which hang up on the slopes during the scaling operations shall be removed upon completion of the first rock slope scaling pass. The new face shall be. inspected by the engineer to determine whether or not the rock slope scaling has been completed.

1016 Appendix 6.3

The contractor shall continue to scale the slope until the scaling has been completed to the satisfaction of the engineer.

All rock and debris produced during the rock slope scaling operation shall be removed by the contractor and disposed of at the waste area shown on the drawings.

If necessary, the contractor shall provide emergency access through the project within 24 hours of notice. A temporary stockpile for scaled material may be located in the ditch and cutslope area between stations - and - . No material shall be stored within 10 feet of the edge of the retaining wall.

Dis~osal of Material

All earth and rock materials removed shall be placed in the designated waste disposal site. All vegetation removed shall be bumed in the designated bum area. The contractor shall obtain an approved burning permit from

Measurement and Payment

Method of Measurement

Rock slope scaling will be measured on a crew-hour basis. A crew is defined as a qualified working foreperson and two qualified scalers. Time will be measured while the crew is working on the slope. Stockpiling, hauling, and disposal of scaled material shall be considered incidental to scaling.

Basis of Payment

Payment for rock slope scaling will be made at the unit price per crew hour for the "rock slope scaling." The unit price shall include the cost of furnishing all the materials, equipment, labor, and incidentals necessary to complete the work as specified. The disposal of all material shall be considered as incidental to rock slope scaling.

DIVISION 2 SITE WORK ROCK BOLTS 02800

General Description

This work shall consist of furnishing and installing rock bolts, complete with component parts and all other materials, in accordance with these specifications and in reasonably close conformity to the lines and dimensions, and at locations to be specified by the engineer.

Single bolts or small groups of bolts shall be used to support potentially unstable isolated blocks of rock. Large groups of bolts shall be installed in a regular pattern to reinforce larger areas in the rock mass. Tensioned rock bolts shall be grouted with Portland cement.

Products

Rock Bolts and Anchor Head Assembly

All rock bolts including anchorages, bearing plates, corrosion protection, and other appurtenances shall be products of a manufacturer regularly engaged in manufacture of rock bolts. Bolts shall be fabricated from deformed bars.

Appendix 6.3 1017

Type 1 Rock bolts shall be 1-inch diameter Williams hollow core prestressable rock and concrete anchor bolts, number RIH-08-C14, or approved equivalent and will require a 1-314-inch drill hole. The head assembly shall be number C14. The design load is 30,000 pounds. An equivalent rock bolt would be a hollow-core deformed rod ASTM A-615 grade 60 steel with a minimum ultimate tensile strength of 50,000 lbs. A mechanical anchor assembly shall be at the end of the bolt rod and shall be capable of exceeding the ultimate tensile strength of the bolt rod. The anchor assembly shall be designed to prestress the bolt rod prior to grouting. The bolt shall be designed to ensure that grout will completely fill the annular space around the bolt rod.

Type 2 rock bolts shall be 1-318-inch diameter Williams high grade hollow core prestressable rock and concrete anchor bolts, number RlHG-11-C18, or equivalent, and will require a 2-114-inch drill hole. The required head assembly is C18. The design load is 80,000 pounds. An equivalent rock bolt would have the same requirements as an equivalent type 1, but would be ASTM A-615 Grade 70 steel, with a minimum ultimate tensile strength of 138,000 pounds.

The rock bolts shall meet applicable specifications of ASTM A-615, ASTM A-331, and ASTM A-722.

All couplers and nuts shall be high strength and the ultimate strengths and maximum working load to yield shall meet or exceed that of the rock bolts.

Where the required length of the rock bolt exceeds 16 feet, the contractor will be permitted to couple lengths of bolts together to produce such lengths. All bolts shall be free of any coatings except at the threaded end. The bolts shall be completely fabricated at the point of manufacture under controlled shop conditions. Couplings for connecting sections of rock bolts shall have a center stop so that each section is connected by an equal length of thread, shall be as strong as the bolt, and shall be so fabricated as not to interfere in any way with the flow of grout.

The surface of the steel rock bolts shall have a continuous pattern of rolled-in raised deformations except for any necessary threads at the ends of the rods.

The rock bolts shall be handled and stored in such a manner as to avoid damage and corrosion. Damage to the rock bolt as a result of abrasions, cuts, nicks, welds, and weld splatter will be cause for rejection. The rock bolts shall be protected from dirt, rust, and harmful substances. A light coating of rust on the steel is acceptable. If heavy corrosion or pitting is noted, the engineer will reject the affected rock bolt.

Prior to installation all mill flaking, rust, and grease shall be removed from the steel. The rock bolt shall be corrosion-protected with grout over the entire surface.

Bearing Plates. Washers, and Nuts

The hearing plates, washers, and nuts shall conform to ASTM F-432 and ASTM A-36. The bearing plates shall be Williams S l K keyhole plates, or approved equivalents. The plates for the type 1 rock bolts shall be 318 by 6 by 6 inches and the plates for the type 2 rock bolts shall be 1 by 8 by 8 inches. Hardware shall be as recommended by the manufacturer.

One or more beveled washers and hardened flat washers for the bolt rod, of a diameter selected to leave the smaller portion of the keyhole shaped opening exposed to a sufficient degree to pass the vent tube without pinching. Anchor nuts shall develop an ultimate strength of not less than 100 percent of the manufacturer's guaranteed ultimate strength of the rock bolt.

Appendix 6.3

Rock Bolt Bearine Pads

A bearing plate, a hardened flat washer, and a nut shall be used to transfer the tension in the bolt to the rock. The plate shall be in uniform contact with the rock surface. If the rock face is not perpendicular to the axis of the bolt, or the rock under the bearing plate is not sound, a bearing pad approved by the engineer shall be constructed so that the bolt is not bent when the tension is applied. Beveled washers may be used to level a bearing plate. Where the rock surface is generally weak or weathered, extra large bearing plates shall be used to distribute the load over a larger surface area to reduce the potential of failure of the bearing zone.

A high-strength, quick-setting cement shall be used for sealing the top of the hole and seating the bearing plate. It shall be of a type recommended by the rock bolt manufacturer and shall be applied as per the manufacturer's recommendations.

Cement Grout

Gmut shall be Wil-X-Cement non-shrink grout mixed with 2-112 gallons of water per 55 lb. pail, or approved equivalent. Alternate cement grouts may be considered that have a fineness as in high early strength cements, such as Portland type I11 (4500 cm2/gm), as measured by the Blaine method. If Portland cement is used, an expansive additive such as "Wil-Grout" powder additive in the proportion 0.005% of the weight of cement (2 grams-1 level teaspoonful-for each 94 lb. sack of cement) shall be required. Hopper working capacity shall be at least 10 gallons.

Comvressive Strength PSI (Modified ASTM C-109)

3 days in moist air = 2800 psi 4 days in water = 2800 psi

Grouting Accessories

Grout tubes, grout sealers, and other grouting accessories for grouting rock bolts shall be types as recommended by the manufacturer and as approved by the engineer.

Lubricant

Lubricant for threads shall be molybdenum disulfide grease

Threads of Bolts and Nuts

The threads of bolts and nuts shall be protected by a plastic tape or molded protector which can easily be stripped off just prior to installation. All rock bolts shall have rolled threads with minimum length of 6 inches and bolts shall be capable of developing the ultimate tensile strength of the bolt.

Corrosion Protection

The exposed rock bolt end, bearing plate nut, washers, and connections for the slope reinforcing mat shall be corrosion-protected. One acceptable method is the use of a bituminous coating such as a single application of TC Mastic Brush-Applied Coating, or an epoxy resin such as Scotch Kote 213, or approved equivalent, applied to a minimum thickness of 5 mm.

Appendix 6.3

Execution

At least 10 days prior to the installation of the rock bolts, the contractor shall supply the engineer the manufacturer's certificate on the rock bolts, accessories, grout, and corrosion protection. At this time, the contractor shall also provide a detailed description of the proposed construction procedure and scheduling to the engineer. The engineer shall be present during the installation of the rock bolts. Rock bolts shall be installed in accordance with and at the locations shown on the drawings unless otherwise directed by the engineer.

The equipment used to drill the holes, to install rock bolts, to effectively seat and tighten the rock bolt andlor establish the anchorage in the hole, and to tighten the bolt to the required tension, shall be in accordance with the instructions of the manufacturer and shall be subject to the approval of the engineer.

The contractor shall provide and maintain in good working condition necessary torque wrenches, hydraulic jacks, and related equipment for installing rock bolts and checking torque.

The contractor shall provide a standard torque wrench for calibrating impact and torque wrenches used in the work. This wrench shall not be used for any other purpose. Prwf of calibration of all impact, torque wrenches, and hydraulic jacks shall be provided at the start of this project.

Location

The tentatively designated bolt locations, numbers, lengths, and spacings are shown on drawings - and -. The final locations and lengths will be determined in the field after completion of scaling. The contractor shall be responsible for temporary bolts required for safety and performance of work.

Drill Holes

Holes for rock bolts may be rotary- or percussion-drilled and shall deviate by not more than 2 percent over their length. The holes shall be declined downward 5 degrees below the horizontal as shown on the drawings, unless otherwise specified by the engineer. Holes where rock bolts cannot be inserted to the intended depth will be rejected. Holes rejected shall be redrilled at no additional cost to the Government.

Holes for rock bolts shall be accurately drilled to the diameter recommended by the manufacturer of rock bolts to be installed or as otherwise approved by the engineer. All threads of the rock bolts shall be free from rust, burrs, and foreign matter immediately prior to the installation of the bolts. Tensioned rock bolts shall have holes drilled 1 foot deeper than the downhole end of the anchor.

Installation

Drilled holes shall be cleaned of all drill cuttings, sludge and debris before the rock bolt is inserted into the hole. The contractor shall make certain that the hollow core of a rock bolt is clear of all obstructions. Rock bolts are to be inserted in the hole with the anchor assembly positioned approximately 12 inches from the bottom of the hole in rock, and the threaded outer end projecting beyond the mortar pad, plate, and nut at least 2 inches. The anchor is to be expanded by rotation of the bolt clockwise with a Williams rock bolt setting tool (or approved equivalent) in accordance with the manufacturer's recommendations. The use of pipe wrenches for applying torque to the bolt rod is expressly prohibited. Torque for setting the anchor assembly shall be in accordance with the recommendations of the manufacturer.

Appendix 6.3

Beveled or hemispherical washers, where needed, shall be placed betkeen the bearing plate and the nut to ensure uniform bearing on the bearing plate. A machine washer shall be placed between the nut and the beveled washer. The washer shall be coated on both sides with molybdenum disulfide grease.

Each rock bolt shall be tensioned to the specified design load prior to grouting. Tensioning shall be done with approved calibrated hydraulic jacks. Tension may be done with a calibrated torque or impact wrench if the relationship between torque and load can be demonstrated to the engineer.

Any rock bolt which is damaged to the extent that it no longer can be retensioned or cannot be grouted completely, shall be replaced at the contractor's expense.

Once a rock bolt has been tensioned, the tension shall not be relaxed for any purpose unless authorized by the engineer.

Type 1 rock bolts shall be tensioned to 32,000 pounds and type 2 rock bolts to 80,000 pounds. Loadextension measurements shall be made during tensioning. The load shall be held for 10 minutes with no loss of load and the extension of the bolt shall not exceed the elastic strain of the bolt by more than 20 percent. For each rock bolt type, a minimum of the first three rock bolts and 10 percent of the remaining bolts will be tested in such a manner. More bolts may be tested if bolts fail or if a valid load-torque relationship cannot be established.

A quality control program of proof-testing of up to 10 percent of all bolts installed shall be carried out on a routine basis by the contractor. Any bolt which fails to satisfy the proof testing shall be replaced at the contractor's cost. A repeat test shall be carried out by the contractor on any new bolt installed to replace a bolt which fails to meet the requirements of the proof-test program by reason of insufficient anchorage capacity.

The engineer may require a repeat of the proof testing at any stage during the execution of the contract to ensure that the method of installation is satisfactory for the conditions encountered. Repeat proof testing shall also be carried out if any changes occur in the method of the installation, material, or personnel.

The proof tests shall be carried out by the contractor using equipment provided by the contractor as follows:

Apply axial load to bolt through a coupling attached to the threaded end, using a hydraulic jack designed specifically for tensioning andor testing rock bolt installations.

Test bolts to 100 percent of design load assuming complete encapsulation

A bolt shall be considered to have failed if any outward movement of the bolt anchorage occurs at a loading below or at the required 100 percent load.

Additional installation quality control will be by means of inspections by the engineer.

Grouting

The entire space between the rock bolt and the sides of the drill hole shall be filled with grout. In order to accomplish this, the grout shall be pressure pumped (using sufficient pressure to overcome the hydrostatic head) with a portable grout pump, providing a minimum of 30 pounds per square inch capacity to the low end

Appendix 6.3 1021

of the hole first and pumping continued until grout is forced out of the de-airing tube at the face of the hole. Failure of grout to be extruded from the face of the hole shall be cause for rejection of the rock bolt installation.

Corrosion Protection

Paint-on corrosion protection shall be applied after all hardware has been connected to the rock bolt in its final configuration.

Measurement and Payment

Method of Measurement

The length of rock bolts will be the sum of the lengths of the rock bolts, measured to the nearest foot, which have been completed and accepted. The length of the head assembly, if any, projecting past the end of the rock bolt shaft will not be included in determining the overall length. The length measurement will begin from the end of the expansion shell. The bolts shall be furnished complete with all accessories including steel bearing plates, nuts, washers, and anchor head assemblies and couplings, if any. This quantity also includes drilling, subdrilling, installing bearing pads, installing the bolt, installing grout or resin, tensioning, testing, locking off, and corrosion protection.

Basis of Pavment

The accepted quantities will be paid in accordance with the contract unit price for each pay item shown in the schedule of items.

DIVISION 2 SITE WORK ROCK DOWELS 02900

General Description

This section specifies the requirements for the installation of type 1. 2, and 3 rock dowels, rappelling anchors, and soil anchors:

Type 1 rock dowel used for cable attachment and anchoring

Type 2 rock dowel used for intermediate anchoring of wire mesh

Type 3 rock dowel used for stabilization of rock blocks

Rappelling anchors

Soil anchors

These rock dowels and anchors will consist of untensioned steel bars, fully encapsulated in polyester resin, epoxy cartridges, cement grout, or another pre-approved method. They are to be installed in drill holes at the locations and to the depths as shown on the drawings or as specified by the engineer. The dowels will be fitted with a face plate, washer and nut, eye bolt, or other hardware as shown on the drawings. An estimate of the number of dowels, and lengths, is shown in drawing -. This estimate is subject to modification based on

1022 Appendix 6.3

actual rock conditions encountered. Payment is to be made on actual quantities used at the unit rates bid on the schedule of items.

The specifications herein refer to the necessary materials, equipment, tools, labor, and workmanship

No borehole anchoring head assemblies shall be used with untensioned rock dowels. Associated hardware shall include standard nuts, washers, face plates, and grout tubes, if required.

Submittals

Not less than 2 weeks prior to commencing the rock doweling, the contractor shall submit in writing to the engineer for approval:

A. Qualifications of the contractor's personnel. The foreman and the drill operator shall have a minimum of 2 years of demonstrated experienced in the installation of rock dowels.

B. A detailed plan for the rock doweling. The plan shall detail:

1. The proposed construction sequence and schedule.

2. The proposed drilling method and equipment,

3. The proposed drill hole diameter.

4. The proposed steel for the rock dowel including certificates.

5. The proposed bearing plate, flat washer, and beveled washer specifications including manufacturer's specifications and catalog cuts.

6. The proposed corrosion protection for the rock dowel system.

7. The proposed epoxy or polyester resin grout specifications including the following:

a. Gel times and final set times, including details of temperature dependency.

b. Resin shelf life and batch numbers.

c. Resin manufacturer's recommendations for mixing times, including temperature dependency

d. Resin manufacturer's recommendations for resin storage

e. Resin manufacturer's recommended cartridge and hole size for the selected bar diameter.

8. That cement grout shall meet the requirement for cement grout in specification 2800 Rock Bolts.

9. The calibration data for each load cell, test jack, pressure gauge, and master pressure gauge to be used in the proof testing. The calibration tests shall have been performed by an independent testing laboratory within sixty calendar days of the date submitted.

Work shall not begin until the appropriate submittals have been approved in writing by the engineer.

Appendix 6.3

Products

Type 1 dowels shall be a 314-inch by 2-foot loop eye rock bolt assembly with a 2-112-inch loop eye. The bar shall be galvanized and deformed, rebar or continuously threaded.

Type 2 dowels shall be ID-inch threaded rebar or 1-inch continuously threaded steel, as shown on the drawings. Dowels and downpipe anchors shall be galvanized.

Type 3 dowels shall be fabricated from no. 6 or no. 10 steel reinforcing bar as shown on the drawings.

Rappelling anchor dowels shall be 318-inch galvanized steel threaded rebar.

Soil anchors shall be galvanized no. 8 rebar with 10 inches of thread, as shown on the drawings.

The minimum hole depth for all dowels is shown on the drawings.

Resin Adhesives

Resin adhesives shall conform to the applicable requirements of AASHTO M-235. If resin cartridges are used, they shall conform to the following:

1. The resin cartridges shall be supplied in cartridge form, and as large in diameter as the hole will allow. The cartridges shall have a casing constructed of a plastic film providing optimum mechanical strength with high frangibility to ensure complete mixing during installation.

2. The cartridge shall contain two distinct fractions of epoxy or unsaturated polyester resin and catalyst. The resin shall be a high-strength polyester or epoxy, highly filled with nonreactive, inorganic filler. The compressive strength of the mixed and cured resin shall be 14,000 psi when tested under laboratory conditions in accordance with ASTM C39-71, "Standard Method of Test for Compressive Strength of Cylindrical Concrete Specimens."

3. Polyester resin or epoxy cartridges shall be readily and individually identified as to their respective gel times.

4. Gel and cure times of anchoring and grouting cartridges shall be as recommended by the manufacturer and approved by the engineer. The resin material shall be thixotropic and of such viscosity that the anchor bar can easily penetrate and mix the material. Cartridge supplies shall be inspected prior to use to see that the resin compounds have not hardened due to improper storage or handling, and meet the above requirements. Cartridges that are older than the stated shelf life shall not be used. Cartridges shall be stored so as to ensure maximum protection until their use. The contractor shall provide facilities to prevent prolonged exposure to sunlight or to high temperatures (above 75 O F ) during storage.

5. Cement grout shall meet the requirements of specification 2800 Rock Bolts

Steel for Dowels

The dowels, soil anchors and rappelling anchors shall consist of ASTM A-615 grade 60 steel and may be in the form of deformed rebar or thread bar.

Appendix 6.3

For type 2 dowels, one end shall be threaded over a length of at least 4 inches. The thread shall correspond to the thread on the face plate nuts.

The dowels shall be free from mill scale, loose flaky rust, and grease.

Type 1 and 2 dowels, rappelling anchors, soil anchors, and pipe anchors shall be hot-dip galvanized.

Corrosion Protection

The exposed end of the dowel and all nuts and washers shall be corrosion-protected. One acceptable method is the use of a bituminous coating such as a single application of TC Mastic Brush Applied Coating. Another acceptable product is a resin such as Scotch Kote 213 applied to a minimum thickness of at least 5 mm. All bearing plates shall be hot-dip galvanized. If the dowel is not installed with cement grout, the entire surface of the dowel, including threads, shall be hot-dip galvanized. Downpipe anchors shall be hot-dip galvanized.

Plate and Fittings

Each type 2 dowel and soil anchor shall be provided with a face plate, except for dowels used to anchor down drain pipe under specification 02703. Soil anchors shall have centering devices placed at 5-foot centers.

The face plate shall be of mild steel, not less than 114 inch thick, and not less than 8 inches square unless otherwise shown on the drawings. Face plates shall be hot dip galvanized. The plate shall have a central hole large enough to fit easily over the dowel while maximizing the available bearing surface for a washer and nut. Spherical seatings will not be required.

For type 2 dowels, the nut shall have a minimum dimension across the plates which results in an adequate bearing surface on the washer and face plate, and shall be to the approval of the engineer. A hardened steel washer shall be placed between the nut and the face plate, or the nut and beveled washers.

Beveled washers shall be used to accommodate non-perpendicular installations.

Execution

Resin Grouted Rock Dowels

Rock dowel hole depth and diameter shall conform to the resin manufacturer's specifications.

Untensioned rock dowels shall be fully grouted either with cement or using a two-part resin cartridge system. Resin shall have gel and set times not exceeding 60 minutes and 4 hours, respectively. Cartridge size and diameter for each borehole shall follow the manufacturer's usage recommendations. All cartridges shall be inspected prior to use to see that the resin compounds have not hardened, are undamaged, and meet the manufacturer's recommendations for the size of the hole to be grouted.

Rock dowelling shall be carried out by operators experienced in rock dowelling, under the direct supervision of an engineer or foreperson who is experienced with the use of these materials and with the installation procedures recommended by the resin manufacturer.

Work shall proceed according to the work plan and schedule submitted by the contractor prior to the commencement of work. The contractor shall notify the engineer, in writing, of any proposed changes to the work plan and schedule, and shall obtain approval of such changes from the engineer prior to proceeding.

Appendix 6.3 1025

Drill Holes

Holes for rock dowels shall be drilled at the angle and to the depth shown on the drawings. The diameter of the holes shall be suitable for the particular diameter of bolt and type of fixing adopted, and shall meet with the approval of the engineer.

Unless otherwise specified, the angle of installation shall be within 5' perpendicular to rock or shotcrete surface.

On completion of drilling, the hole length shall be checked and shall be cleaned to provide a clean, dry, dust- free surface for anchorage.

Installation

Prior to installation, all mill scale, flaking rust, and grease shall be removed from the dowel steel. The rock dowel shall be corrosion-protected over the entire surface. All exposed parts of the rock dowel, bearing plate, and nut on the surface shall be painted with an approved corrosion-protection paint.

Sufficient resin cartridges to bond the entire length of the dowel shall be pushed to the back of the hole and supported in place if necessary.

The dowel shall be inserted into the hole following the manufacturer's recommendations.

After the dowel has been fully inserted, rotation shall be continued at the speed and for the duration recommended by the resin manufacturer. The mixing time shall be adjusted for the ambient temperature of the rock mass or resin storage area according to the manufacturer's recommendations. The dowel shall be maintained in position until the resin has gelled.

When the resin has reached final set, a face plate shall be placed over the dowel and held in place by a nut. The nut shall be torqued to a nominal 100 ft-lbs to ensure proper seating against the rock.

For type 3 dowels, the dowel shall be as close as possible to the rock that it is to support. The dowel shall be grouted full-depth into the drill hole. The dowel and the toe area of the block that it is supporting shall be encased with shotcrete or hand-packed cement for corrosion protection and support. Wood packing shall not be used.

For soil anchors, centering devices shall be installed at 5-foot intervals and the dowel inserted into the drill hole. Grout shall be pumped through a grout tube into the bottom of the hole and the hole completely filled with grout. A temporary form may be used at the face and may consist of the 8 by 8 by 318-inch steel face plate sufficiently greased to prevent bonding of the grout. The face plate may subsequently be removed and used as intended in the installation. Other suitable methods may be approved by the COR.

A quality control program of proof-testing of up to 10 percent of all dowels and rappelling anchors shall be carried out on a routine basis by the contractor. Any bolt or dowel which fails the proof testing shall be replaced at the contractor's expense. A repeat test shall be carried out by the contractor on any new bolt or dowel installed to replace a bolt or dowel which fails the proof test because of insufficient anchorage capacity.

The engineer may require a repeat of the proof testing at any stage during the execution of the contract to ensure that the method of installation is satisfactory for the conditions encountered. Repeat proof testing shall also be carried out if any changes occur in the method of installation, material or personnel.

1026 Appendix 6.3

The proof tests shall be carried out by the contractor using equipment provided by the contractor as follows:

Apply axial load to bolt through a coupling attached to the threaded end, using a hydraulic jack designed specifically for tensioning or testing rock bolt or dowel installations.

Test bolts to 100 percent of design load assuming complete encapsulation. A bolt shall be considered to have failed if any outward movement of the bolt anchorage occurs at a loading at or below the design load (100 percent).

Design loads are. as follows:

Dowel Tvne Design Load

Type 1 18,400 Type 2, No. 4 7,800 Type 2, No. 8 20,000 Type 3, No. 6 18,000 Type 3, No. 10 - Rappelling Anchors 2,700

Additional installation quality control will be by means of inspections by the engineer.

Measurement and Payment

Method of Measurement

Rock dowels and rappelling anchors will be measured for payment on a per-each basis. Soil anchors will be measured for payment per foot of soil anchor installed and accepted, to the nearest foot, not including any measurement protruding from the drill hole.

Basis of Pavment

Payment will be made for each of the following bid items:

PAY ITEM DESCRIPTION

02900(01) Type 1 Rock Dowel Each 02900(02) Type 2 Rock Dowel Each 02900(03) Type 3 Rock Dowel Each 02900(04) Rappelling Anchor Each 02900(04) Soil Anchor Linear Foot

The unit contract price for the above listed bid item shall be full pay for furnishing all labor, tools, materials, and equipment necessary for the completion of the work as specified.

Appendix 6.3

DIVISION 2 SITE WORK SLOPE CONTAINMENT MAT 04000

General Description

Free Haneine Mat

Fabric mat shall be hung on slopes as shown on drawings to direct dislodged rock to a road- elevation ditch for removal.

All matting shall be constructed at the locations shown on the drawings or elsewhere as directed by the engineer and shall conform to the design shown on the drawings and these specifications.

Slow Reinforcine Mat

Fabric installed over rock bolt, soil anchor, and rock dowel reinforced slopes shall be secured to each rock bolt, soil anchor (after testing), or dowel behind a steel plate as shown on drawings . Fabric shall be installed after each bolt or dowel has been installed fully. Material shall be triple twist wire mesh fabric as specified below.

Products

Wire Mesh Fabric

Use galvanized steel wire for wire mesh fabric, meeting the requirements of ASTM A-641, with a nominal diameter of 0.120 inch, class 1 coating, and a minimum tensile strength of 60,000 psi. Maximum mesh size shall be approximately 4-314 inches with triple twist and hexagonal shape.

Anchor Shackles

Anchor shackles shall be 318-inch diameter forged steel with 7116-inch diameter alloy steel screw pins. The minimum working load limit for the shackles shall be 2000 pounds. The shackles and pins shall have a galvanized finish.

Cable shall be 318-inch diameter, 6 by 19 classification, galvanized wire rope with independent wire rope core made from extra-improved plow steel. It shall have a zinc coating of at least 0.20 oz.1sq. ft. on all wires, and a breaking strength of at least 13,000 pounds.

The contractor shall submit certification that the cable meets these standards at least 10 days prior to installation.

Hardware

All rings shall be drop-forged steel heat-treated after forging. Wire rope thimbles weighing approximately 13.8 pounds per hundred shall be used with the 318-inch diameter cable. All rings, thimbles, wire rope clips, and U-bolts shall be galvanized according to AASHTO M-232 (ASTM A-153), class C, except that casting shall be class A, and forgings shall be class B.

Appendix 6.3

Hoe Ring Fasteners

Hog ring fasteners shall be fabricated from no. 9 gauge, zinc-coated steel wire conforming to ASTM A-116, class 1.

Reoair of Damaeed Coating

In addition to the methods given in AASHTO M-36, the contractor may repair damaged zinc or aluminum coating by removing all loose or cracked coating, and welding flux, wire brushing the damaged area, and applying two coats (at least 2 mm total thickness) of a high zinc-dust-content paint conforming to the general requirements of ASTM A 780. The paint manufacturer shall certify in writing that the paint furnished complies with the above requirements.

Execution

Slooe Containment Wire Mesh Fabric

Slope containment wire mesh fabric shall be placed as shown on the drawings, with the fabric secured to the tension cable horizontally and vertically and at overlapping horizontal and vertical joints with galvanized steel hog rings as shown on drawing -. In addition, the fabric shall be secured horizontally to the tension cable with shackles at 5-foot centers. For free hanging mat, the fabric shall not be tensioned in any direction, but must remain loose to increase its dampening effect on rolling rocks. The bottom of the fabric shall rest on the slope as shown, such that material dislodged under the fabric can drain freely from the bottom, yet will not flow or bounce onto the roadway.

No horizontal butt joints are allowed; see the drawings for horizontal and vertical seam details,

Tensioning

For slope reinforcing mat, the wire mesh fabric shall be tensioned to the rock bolts and type 2 rock dowels with a torque wrench (see drawings ). Tension on these nuts shall be at least 50 ft-lbs minimum. Once a rock bolt has been tensioned, do not relax the tension for any purpose unless authorized.

Shackles

Only properly fitting pins shall be used. Shackles should not be pulled at an angle; fittings and cable should be centralized on the pin with suitable washers or spacers.

Measurement and Payment

Method of Measurement

The quantities of free hanging mat and slope reinforcing mat shall be measured by the square foot, to the nearest square foot, of surface area of wire mesh fabric in the mat.

Basis of Pavment

Payment at the contract price per square foot for the free hanging and slope reinforcing mat will be payment in full for furnishing and placing all materials and performing all earthwork, including tools, labor, and incidentals necessary to complete the work.

Appendix 6.3

The accepted measured quantities will be paid for at the contract unit price per unit of measurement for the following items:

PAY ITEM DESCRIPTION UNIT OF MEASUREMENT

04000(01) Free Hanging Mat Sq.Ft. 04000(02) Slope Reinforcing Mat Sq.Ft.

DIVISION 2 SITE WORK SHOTCRETE APPLICATION 03800

General Description

This work shall consist of constructing a pneumatically-applied steel- or polypropylene fiber-reinforced shotcrete blanket and the installation of the bar reinforcement onto rocWsoil surfaces at locations shown on the drawings or as directed by the engineer.

These specifications refer to premixed cement and aggregate pneumatically-applied by suitable equipment and competent operators.

The shotcrete shall be composed of Portland cement, fine and coarse aggregate, and water. Either wet-mix or dry-mix shotcrete may be used.

The shotcrete shall be. applied according to these specifications and applicable sections of the American Concrete Institute's "Guide to Shotcrete" (ACI 506R-85).

Qualifications of Contractors' Personnel

At least 30 days prior to beginning shotcrete work, the contractor shall provide written evidence that the foreperson, nozzleman, and delivery equipment operator have performed satisfactory work in similar capacities elsewhere for a sufficient length of time to be fully qualified to perform their duties.

The foreman shall not have less than 2 years' experience as a shotcrete nozzleman. The nozzleman and delivery equipment operator shall have served at least 1 year of apprenticeship on similar applications with the same type of equipment. Prior to the start of shotcreting for this job, the nozzlemen shall, in the presence of the engineer, demonstrate their ability to apply shotcrete of the required quality on a test panel. One satisfactory test panel shot in a vertical position for each mix design used during the course of the work shall be the minimum qualification test for nozzlemen before they will be permitted to place shotcrete in permanent construction.

Products

Materials shall conform to the requirements of the included specifications supplemented and modified as follows:

Portland cement (11)

Curing materials and admixtures

Water

Appendix 6.3

Fine aggregate

Coarse aggregate

Bar reinforcement

Preuackaeed Product

If approved by the engineer, premixed and prepackaged concrete product with fibers specifically manufactured as a shotcrete product may be provided for shotcrete mixed on-site. The packages shall contain cement, aggregate, and (if appropriate) fibers conforming to the materials portion of this specification.

Admixtures

Admixtures shall not be used without permission of the engineer. If admixtures are used to entrain air, reduce water-cement ratio, retard or accelerate setting time or accelerate the development of strength, they shall be used at the rate specified by the manufacturer and must be compatible with the cement used. Use of calcium chloride accelerating agent will not be permitted. When used, admixtures shall be dissolved in water before they are introduced into the mixture.

Water - The water used in the shotcrete mix shall be suitable for domestic consumption and shall also be free of elements which would cause staining.

Aeereeates

The combined gradation of fine and coarse aggregate used in the shotcrete shall meet the following grading requirements:

Sieve Size Percent Passing bv Weieht

112" 318" No. 4 No. 8 No. 16 No. 30 No. 50 No. 100

Anchor Bars

Unless shown otherwise on the plans, anchor bars shall consist of no. 5 reinforcement bar bent into an L-shape. One leg of the L-shaped bar shall be approximately 6 inches long and the other 2 feet long.

Steel Fiber Reinforcement

Steel fiber reinforcement shall meet the following requirements. Steel fibers shall have a length between 112 and 1-318 inches, have blunt or hooked ends, have a length-to-diameter ratio of less than 80, and shall be cold- drawn carbon steel with a tensile strength of at least 160,000 psi. Only steel fibers manufactured specifically

Appendix 6.3 1031

for use in shotcrete applications will be allowed. The steel fiber content shall not be less than 100 pounds per cubic yard of shotcrete.

Synthetic Fiber Reinforcement

The fibers shall be made from polypropylene, polyester, or appropriate co-polymers, specifically for reinforcing concrete. Fibers shall be at least 314 inch long. Acceptable products include Fibermesh, Forta CR. HiTech Fibers, and Nurlon (normally about 2 lbslcubic yard).

Execution

Pump system. The pump system used to convey premixed shotcrete ingredients shall deliver a uniform and uninterrupted flow of material, without segregation or loss of ingredients. The mixing equipment shall be capable of thoroughly mixing the specified materials in sufficient quantity to maintain continuous placing.

Air compressor. The air compressor shall be capable of maintaining a supply of clean air adequate for maintaining sufficient nozzle velocity for all parts of the work and for the simultaneous operation of a blow pipe for clearing away rebound. The compressor shall be capable of providing at least 250 cfm per operating nozzle.

Batching and mixing equipment. The mixing equipment shall be capable of thoroughly mixing a sufficient quantity of the materials to maintain continuous application.

Dry-mix process delivery equipment. The equipment shall be capable of discharging the aggregate-cement mixture into the delivery hose and delivering a continuous stream of uniformly mixed material to the discharge nozzle. The discharge nozzle shall be equipped with a manually operated water injection system (water ring) for directing an even distribution of water through the aggregate-cement mixture. The water valve shall be capable of ready adjustment to vary the quantity of water and shall be convenient to the nozzleman. The water pressure at the discharge nozzle shall be sufficiently greater than the operating air pressure to assure that the water is thoroughly mixed with the other material. The water pressure shall be steady (nonpulsating). Equipment parts, especially the nozzle liner and water ring, shall be inspected regularly and replaced as required.

Wet-mix process delivery equipment. The equipment shall be capable of discharging the premixed materials into the delivery hose and delivering a continuous stream of uniformly mixed material to the discharge nozzle. Recommendations of the equipment manufacturer shall be followed on the type and size of nozzle to be used and on cleaning, inspection, and maintenance of the equipment.

Surface preparation. Immediately prior to shotcrete application, rock surfaces of the areas to be shotcreted shall be scaled of all loose material and shall be thoroughly cleaned by use of air or water jets, or other means approved by the engineer, in order to provide a good bonding surface. Soil surfaces shall loose material by means of an air jet.

Shotcrete shall not be placed on any surface which is frozen, spongy, or where there is free water. The surface shall be dampened before applying shotcrete.

Appendix 6.3

Shotcrete blanket thickness control. The thickness of the shotcrete blanket shall be controlled by installing noncorrosive pins, nails, or other gauging devices normal to the face, such that they protrude the required shotcrete thickness outside the face. These pins shall be placed on a maximum 5-foot square pattern.

Anchor bars. Unless otherwise shown on the plans, anchor bars shall be placed at approximately 10-foot centers maximum, both horizontal and vertical, in I-114-inch holes drilled approximately 24 inches deep into the rock or soil. The drilled hole shall be blown clear prior to installation of the anchor bar. The drilled hole shall be filled completely with neat cement grout using a grout tube extending to the bottom of the hole. The anchor bar shall be pushed into the grout-filled hole and centered such that the short leg of the L-shaped bar points upward and is located about 1-112 inches from the rock or soil surface.

Weep holes. Unless otherwise shown on the plans, weep holes shall be provided throughout the shotcrete mat at 10-foot centers maximum, horizontal and vertical. The weep holes shall consist of 2-foot-long, 2-inch- diameter Schedule 40 PVC slotted drain pipe placed within predrilled holes and sloped 5 percent to drain. Predrilled holes shall not be larger than 3 inches in diameter. The slotted drain pipe shall be installed prior to placement of shotcrete. The daylighted end shall extend 1 to 3 inches outside the slope. During placement of shotcrete, the weep holes and drain pipes shall be protected against contamination to ensure that they function properly upon completion.

Batchine and Mixing Shotcrete

Dry-mix process. The cement and aggregate shall be batched by weight. Predampening shall be. carried out prior to flow into the main hopper and immediately after flow out of the packaging in order to ensure that the premix will flow at a uniform rate (without slugs) through the main hopper, delivery hose, and nozzle to form uniform shotcrete, free of dry pockets. No predampened cement-aggregate mix shall be used if it is allowed to stand for more than 90 minutes.

Wet-mix process. Batching and mixing shall be done according to the applicable provisions of ASTM C 94.

Batchine and Mixine Fibers

The procedure used for adding steel fibers to the shotcrete shall be determined by the contractor and shall be subject to approval by the engineer. The procedure to be used shall be demonstrated in the field before approval for production operations is granted. If fiber addition takes place at the nozzle, fibers shall be uniformly distributed throughout the mortar matrix without isolated concentrations. If fibers are added to the dry or wet mix during the batching and mixing process, a screen having a mesh of 1-112 to 2-112 inches shall be used to prevent any fiber balls from entering the shotcrete line. Batching through a screen will not be required if it is demonstrated that fiber balls are not being formed. Fibers shall not be added to the dry or wet mix at a rate faster than they can be blended with the other ingredients without forming balls or clumps. Bulk fibers that have a tendency to tangle together shall pass through a vibrating screen or be carefully sifted into the mix so that they enter it as individual elements and not as clumps.

Shotcrete Ao~lication

The minimum thickness of shotcrete shall be 3 inches. The shotcrete shall be applied from the lower portion of the area upwards so that rebound does not accumulate on the portion of the surface that still has to be covered. Rebound material shall not be worked into the finished product. Rebound is defined as the shotcrete constituents which fail to adhere to the surface to which shotcrete is being applied. It shall not be salvaged and included in later batches. Shotcrete shall emerge from the nozzle in a steady unintempted flow. When, for any reason, the flow becomes intermittent, the nozzle shall be diverted from the work until steady flow resumes. A nozzleman's helper, equipped with an air blowout jet, shall attend the nozzleman at all times during the placement of shotcrete to keep the working area free from rebound.

Appendix 6.3 1033

Shooting shall be suspended if:

High winds prevent the nozzleman from properly applying the material,

The temperature drops below 40 O F , or

External factors, such as rain or seepage, wash cement out of the freshly-placed material or cause sloughs in the work.

Construction joints shall be tapered over a minimum distance of 12 inches to a thin edge and the surface of such joints shall be thoroughly wetted before any adjacent section of m o m is placed. Square construction joints shall not be permitted.

The surface shall be sounded with a hammer for unsound areas resulting from rebound pockets or lack of bond. These areas, sags, or other defects shall be cut out carefully and replaced with a succeeding layer at the contractor's expense. When fabric reinforcement is used and is damaged or destroyed by such repairs, the damaged area shall be replaced by additional properly lapped and tied wire fabric.

Where a layer of shotcrete is to be covered by a succeeding layer, it shall first be allowed to take its initial set. The initial layer shall be cleaned of all loose material prior to placing succeeding layers.

Finishing. The shotcrete surface shall be left in the natural gun finish,

Curing. Air-placed shotcrete shall be cured by applying a white-pigmented, liquid membrane-forming curing compound as specified in subsection 9-23.8 of the standard specifications. The curing compound shall be applied immediately after gunning. The air in contact with shotcrete surfaces shall be maintained at temperatures above freezing for a minimum of 7 days. Curing compounds shall not be used on any surfaces against which additional shotcrete or other cementatious finishing materials are to be bonded unless positive measures, such as sandblasting, are taken to remove curing compounds completely prior to the application of such additional materials.

Acceptance Sampling and Testing

Shotcrete Com~ressive Strength

The shotcrete shall be capable of attaining 2500 psi compressive strength at 7 days (1800 psi at 3 days) and 4000 psi at 28 days as determined by AASHTO T-22 (ASTM C39-84) testing of compression test cylinders.

Failure of Shotcrete

Should any shotcrete section be deficient in any of the specified criteria, that section shall be remedied to the engineer's satisfaction at the contractor's expense. Such remedies may include, but are not limited to, removal and replacement of the substandard section.

Measurement and Payment

Method of Measurement

The area of shotcrete blanket to be paid for will be the number of square feet constructed according to the plans or as directed by the engineer.

Appendix 6.3

Basis of Pavment

Payment for shotcrete blanket will be made at the unit price per square foot for the item "shotcrete rock slope stabilization." The unit price shall include the cost of furnishing all materials, labor, equipment, and incidentals necessary to complete the work described in this section. The anchor bars and their installation shall be considered as incidental.

Appendix 6.3

APPENDIX 6.4

6.4 Powder Creek SSI Demonstration Problem

Rend Renteria, Geotechnical Engineer, Intermountain Regional Office

Introduction

History

Powder Creek slide is a large translational slope movement located on the Willamette National Forest near Oakridge, Oregon. The site is located on Road 2120 at milepost 4.8 in the Buck Creek Drainage at the 3,500 foot elevation in section 13, T. 23 S., R. 3 E., W.M.

Displacement of the roadway (then Road 2321) was noted in the early 1960's. Wooden piles were driven into the slide mass in an attempt to stop movement. The treatment was unsuccessful, as indicated by 1984 sightings of pile tips above the failure surface.

A slope movement southeast of Powder Creek Quarry circa 1974 required a road realignment. The realignment shifted the road centerline upslope. The counterline subsequently became the head of the present failure of Area A. The realignment created a through-cut east of Area A. It is thought that separate movements occurred south of the through-cut and south of the road crossing at Powder Creek.

By 1976, the slope movements had caused numerous tension cracks along the road in Area A. The recommendations at that time were to relocate the Powder Creek channel to the east, drain the marsh upslope of Area A, and drain the marsh near the road at Powder Creek. The marshes were drained by excavating a trench to Powder Creek. The trench and stream relocation successfully drained the marsh near the road and eliminated the tension cracks in the roadway. However, the marsh above Area A remained.

A large clearcut unit at the headwater of Powder Creek was harvested circa 1979. The logging operation required that a one- to two-acre spring-fed pond be drained; a trench was created to drain the pond into Powder Creek.

At this time the road grade began a vertical displacement of 1 to 3 feet yearly through Area A. The maintenance crew has bladed and reshaped the road annually with base rock at an estimated average cost of between $3,000 and $5,000 per year.

In 1984, a comprehensive slope stability investigation involving a site survey and subsurface investigation was begun by the South Zone Geotechnical Group. Subsurface soil samples were classified and tested. Piezometer tubes and a slope inclinometer casing were installed in the drill holes to monitor subsurface water and to locate the failure surface. Comparison of survey points in the middle of Area A from 1976 to 1984 indicated an average horizontal displacement of 10 feet per year,

Appendix 6.4

with a corresponding estimated average yearly loss of 2,500 cubic yards of soil into Buck Creek.

During the winter of 1984-85 a slope movement sheared the piezometers and inclinometer casing. With the failure surface defined, analysis indicated that water pressure was the driving mechanism. During 1985, three measures were attempted to control the water in the failure area. In March, three horizontal drains were installed from below the road, and four vertical siphon wells were installed. A subsequent movement sheared and blocked the drains. The vertical siphon wells were also a disappointment due to clogging by soil fines and low permeability limiting the siphon flow. In September, an attempt was made to drain the marsh above Area A. The installed trench reduced the subsurface water levels near the marsh, but no effect was noted downslope of the marsh. The water source of the marsh is still undetermined.

In 1986, an additional subsurface exploration was conducted. Samples were taken at the failure surface to be classified and tested for strength and piezometers were installed to monitor the subsurface water levels. The yearly horizontal movement was measured to be 15 feet, with 2,000 cubic yards of soil being moved.

By March 1987, an additional 12 feet of horizontal movement at the toe was noted. The road grade had displaced 1 to 2 feet vertically and horizontally. The estimated volume of soil displaced was less than 1,500 cubic yards.

Subsurface Subsurface exploration was conducted in 1984 and 1986 using an Acker Mark IV

~ ~ ~ l ~ ~ ~ t i ~ ~ and track-mounted drill. The objectives of the drilling were to confirm or modify the

Testing subsurface interpretation of material type and extent, the location of the failure zone, and the subsurface water conditions, and to obtain and test samples of subsurface materials for engineering properties relevant to the design of alternatives.

Drilling was performed using hollow stem augers and the standard penetration test (SPT) method andlor wireline core drilling. Shelby tube samplers were pushed to obtain undisturbed samples for strength testing. Two slope inclinometer casings were installed to monitor slope movement and provide data for identifying the failure surface. One-inch-diameter PVC pipe was installed in the drill holes to monitor the seasonal subsurface water.

Subsurface testing consisted of SPT, soil permeability, and strength testing in the shear zone. Permeability tests were run using constant-head and falling-head methods. In-situ strength testing was accomplished with the Iowa borehole shear strength device.

Laboratory testing was used to confirm the field soil classification and to perform strength tests on undisturbed samples. Soils were classified according to the Unified Soil Classification System (USCS). Direct shear strength tests were run on samples taken from a saturated soil layer above the shear zone. The samples were tested under field moisture and saturated moisture conditions. Triaxial shear strength tests were run on saturated shear zone samples using the consolidated undrained method with pore pressure measurements.

Appendix 6.4

Site Conditions The Powder Creek drainage is located in a former glacial cirque basin which drains transversely into Buck Creek. The soils found in the Powder Creek area are the result of glacial, fluvial, and mass movement processes. As Buck Creek enters the basin, the channel has been deeply incised into the soils at the lower edge of the basin. Subsurface investigation has shown soil depths greater than 100 feet within the basin area, excluding the Buck Creek channel.

On average, the slide progresses 10 feet horizontally per year by slow movement, with one to three major movements between early winter and early spring. The body of the failure translates as a block slide that displaces material off the slope and into Buck Creek. At the failure head, the response affecting the roadway has been limited to 1 to 3 feet per year of vertical displacement and minor horizontal movement. This observation indicates that movement is not initiated by the weight of material at the head. The saturated conditions at the toe cause a liquid condition of the soil material which will flow the 400 feet downslope to Buck Creek.

For engineering and discussion purposes four soil units were identified. A brief description of the units follows:

Soil Unit A1 (SU-A1)Sil tv sand, damp to wet, plastic, USCS: SM (material of failure masses).

Soil Unit A2 (SU-A2)Sil tv sand with rock fragments, damp to moist, plastic, USCS: SM-GM.

Soil Unit B ( S U - B t R o c k fraements with siltv sand, plastic, USCS: GM-GW.

Soil Unit F (SU-F)-Fill material: silty sand with rock fragments, damp to moist, plastic, includes aggregate surfacing.

Interpretation of the data obtained by the subsurface exploration is believed to confirm the failure mass as translational with water pressure at the failure surface. The SPT results indicate a failure zone of 1 foot or less. The initial drilling in October 1984 indicated no water in the failure zone of Area A. The shearing of the water monitoring pipes in December 1984 provided the location of the failure surface. However, the actual water pressure at the failure surface could not be determined reliably from the sheared pipes (because of leakage and blockage). Followup drilling in March 1986 confirmed water at the failure surface within a zone of anesian pressure. The water pressure head measured at the failure surface (sealed ADH-14A) was similar to the head measured in piezometers sealed near the top of the water zone (ADH-14B, 14C). This indicates a common water source for the horizontal soil layers noted in the drilling investigation.

Water pressure heads inside Area A vary less than measurements taken in the outside areas. In Area A the water pressure heads measured in the upper section of the failure mass increase 1 to 2 feet from summer to winter. Water flow at the toe area of the failure falls off sharply in the summer (soil is damp; no functioning water observation pipes). The maximum rise in water level in Area A was monitored using ground-up cork placed in an observation pipe. Observations of water pressure head outside Area A show increases of 2-8 feet from summer to winter.

Appendix 6.4

Analysis

Stabilization Alternatives

The conclusion is that subsurface water from outside of Area A, increased by winter rains, is forced into permeable zones of Area A from all sides, and possibly from below. Water flow in Area A may occur as parallel artesian flow separated by a less permeable zone. Over the wet season, the less permeable zone will become saturated, and the subsurface water will become more like a "ground water table." Increased subsurface water has lowered the effective soil strength, which allows failure to occur. This interpretation is based on observations of the varying subsurface water conditions during drilling and over two to four winter seasons. Supporting evidence is provided by the observation during the summer of relatively unchanged subsurface water conditions at the head of the failure and a drained toe of the failure mass. Therefore, the rise in water pressure head outside of Area A must contribute to the water found in the toe during the winter by restricting the summertime drainage paths.

The failure was analyzed as a translational block, with a rotational toe and head, and residual shear strength mobilized to resist the movement. The location of the failure surface was established from an analysis of field data and observations. Residual strength theory was used as the basis for the strength parameters obtained through a back-calculation analysis at the moment of failure (factor of safety equal to 1.00). The hypothesized mechanism of failure begins when increased subsurface water from outside of Area A restricts available drainage paths of the toe area. The water pressure then increases to a level sufficient to initiate failure.

The shear strength in the failure zone of Area A has been decreasing over time as a result of large horizontal displacements. A factor of safety was obtained using the friction angle and cohesion value based on laboratory and field strength tests, and values for similar local soils. Residual strength values were obtained by assuming no cohesion and a friction angle within the range given by figure 4C.12 (PI=18), to obtain a back-calculated factor of safety of 1.

Five potentially feasible alternatives were considered to stop the movement and provide a stable roadway. Two alternatives which proved to be infeasible were to shift the alignment and remove the fill (analysis showed no gain in stability) and to realign the road through the marsh area (environmentally unacceptable to damage wetlands). A description of each feasible alternative is presented here, and a summary of the alternatives is presented in table 6.4.1. Probabilities of success of stabilization were subjectively assigned to each of the alternative outcomes based on the use of the conservative separate ground water model. This model considers that ground water flow may not be continuous, an artesian condition may exist, and the toe area may be influenced during the winter months, creating a separate water source from that under the road.

Alternative A-continue with maintenance C'do nothine"). Repair consists of blading and shaping. Minimal aggregate should be added to restore a smooth grade. Continued failure will produce deepening of existing vertical sag curve. An unusual storm event (such as rapid snowmelt or a 50-year event) increases the probability of a complete debris flow failure involving up to 45,000 cubic yards of soil.

z y . Replace the excavated material using select borrow material from Powder Creek Quarry. There is

Appendix 6.4

a moderate probability that water may re-enter the area from the sides or from below. A failure at the toe area would resume the resource damage and eventually affect the roadway.

Alternative C-rock buttress at the toe. Excavate a foundation at the toe and place a rock buttress. This alternative is not influenced by direction of water flow entering the failure area, but placement of a buttress below the road is infeasible due to the depth to the failure surface.

Alternative D-dr i l l and install horizontal drains at the toe. A drill pad and access road would have to be constructed. Access would be limited to May through October (FOS=l.lO) due to saturated conditions at the toe. A dual-tier drain design increases the probability of draining water at the failure surface regardless of the direction from which water enters. There is a moderate probability that a circular failure at the toe could shear the drains and block the drainage.

Alternative E-drill and install oiles at the toe. Placement of piles at the toe protects Area A from further resource damage. Placement from below the road would increase the depth of drilling. This alternative is not influenced by direction of water flow.

The recommended preferred alternative is to drill and install horizontal drains at the toe (alternative D). This alternative is the most cost-effective based on an expected monetary value (EMV) analysis that considers probabilities of failurelsuccess and anticipated resource damage. If movement at the toe cannot be halted with the drains, then alternatives B and C would also likely not succeed. If the drains are not fully successful in stopping the movement, a much smaller buttress than that required in alternative C may be considered. However, a high degree of control would be needed not to damage the installed horizontal drains.

Table 6 .4 .14ummary of alternatives.

Appendix 6.4

Alternative

A - Maintenance

B - Shear Trench

2 - Riprap Buttress

D - Horizontal Drains

3 - Drilled Piles

FOS

- < 1.00

1.25

1.25

1.19

1 SO

Design Life 25 Yr Possible Outcomes

"status quo"

non-maint. failure

success

failure

success

failure

success

failure

success

failure

Outcome Probability (%)

50

50

90

10

90

10

80

20

95

5

Outcome PW Cost ($)

54,700

85.000

45,000

82,200

249,700

249,700

37,200

9 1,900

300,000

345,000

C O ~ C ~ U S ~ O ~ S The Powder Creek Basin is an area of past instability reactivated by major

and disturbances in the past 25 years. The location of Road 2120 at the toe of this basin

Recornrnend- makes the alignment vulnerable to mass movements of reactivated ancient failures. The material in the failure zone has shown a reduction in strength with increasing

ations displacement. It is believed that this strength reduction has reached its lowest value, which is not a stabilizing effect by itself. Failures driven by the subsurface water will continue and may be accelerated by activities which increase the subsurface flow further up the drainage. It is felt that the failing grade of Road 2120 through Area A can be treated with a high probability of success.

The recommendation to repair the failing grade on Road 2120 and to minimize resource damage at the site is to install horizontal drains (alternative D). This alternative is the most cost-effective and has a high probability of success. In addition to installing the drains, adequate ditch drainage must be improved through Area A. This can be accomplished by use of a trenchlpipe drain or a vertical curve improvement.

It is further recommended that the area be monitored and all activities in the area be examined for their effects on the subsurface water flow. No live vegetation should be removed from the failure areas. The geotechnical group should be consulted for technical input for future sale activity and road construction in the upper reaches of the Powder Creek Basin.

Appendix 6.4

APPENDIX 6.5

6.5 Probability and Statistics

Rend Renteria, Geotechnical Engineer, Intermountain Regional Office

The content of this appendix is borrowed liberally from Miller et al. (1987)

Probability Theory: An internally consistent branch of mathematical logic, consisting of a systematic statement and formulation of principles that necessarily follow from a limited set of fundamental axioms.

Probability concepts are related closely to similar ideas found in set theory.

Set Theory

Space S, largest set containing all elements of all sets under consideration

Elements a , b, ...

Sets A, B, ... (subsets of S)

Probability Concepts

Sample space (sure event)

Sample points (possible outcomes or occurrences)

Events (collections of one or more sample points)

Impossible event

Event B occurs

Event B does not occur

Event A occurs, or event B occurs, or both occur

Both events A and B occur

Events A and B are mutually exclusive (they cannot occur simultaneouslv

In probability theory it is assumed that a random experiment, or sampling exercise, will have outcomes (or sample points) that depend on chance. A collection of one or more outcomes is known as an event.

Appendix 6.5

Define p(B) as the probability of event B.

The value p(B) is a finite number assigned to event B, and thus can be thought of as a probability function whose value is a function of the event B.

The assignment of probability values to events usually is based on:

Relative frequency of observed occurrences ("hard" facts or information based on sampling history and past experiences).

Relative subjective likelihood ("soft" information based on available information, professional judgment, common sense).

Axioms of Probability

0 5 p(B) 5 1 (for any event B in the sample space)

p(S) = 1 (S is the sample space)

If A,. A,, ... A, are mutually exclusive events in S, then p(A, A, LJ ... u A,) =p(A,) +p(AJ + ... +P&)

If event A is contained in event C (A is a subevent of C), then p(A) < p(C).

p(A u B) = p(A) + p(B) - p(AB) for any events A and B in S

Two events A and B are said to be independent (one event has no relation to the other event) if and only if

PVIB) = p(A)p(B).

Conditional Probability

The conditional probability of event A, given that event B has occurred, is defined as

If events A and B are independent, then A does not depend on B and p(AIB) = p(A); B does not depend on A and p(BL4) = p(B); Each event is independent of the other's complement, and the two complements are independent.

Appendix 6.5

Theorem of Total Probability:

For events B,, B,, ... , B, that are mutually exclusive and exhaustive (i.e., B, u B, u ... u B, = S), the probability of any arbitrary event A in S is given by

Bayes Theorem:

For any two arbitrary events A and B with p(A) # 0 and p(B) # 0,

For a set of mutually exclusive, exhaustive events B,, the above relation can be expressed as

This theorem allows the posterior probability p(BIA) to be evaluated in terms of the prior information given by p(B) and p(AIB).

Appendix 6.5

Reference Miller, S.; Prellwitz, R.; and Hammond, C. 1987. Applications of probabilistic methods and decision analysis to geotechnical engineering and resource management: Siskiyou National Forest. Course notes. 1 6 1 8 June 1987: Gold Beach, OR. Moscow, ID: U.S. Department of Agriculture Forest Service, Intermountain Research Station.

Appendix 6.5

APPENDIX 6.6

6.6 Field and Laboratory Evaluation of Geocomposite Drain Systems for Use on Low-Volume Roads

Edward Stuart 111, Geotechnical Engineer, Pacific Southwest Region Kenneth S. Inouye, Geotechnical Engineer, Pacijic Southwest Region James A. McKean, Engineering Geologist, Pacific Southwest Region

The following paper was published in the Transportation Research Record, Fifth International Conference on Low-Volume Roads, May 19-23, 1991, Raleigh, NC, Transportation Research Board, National Research Council, Washington D.C., No. 1291, Vol 2. pp 159-165.

The Forest Service conducted an in-depth laboratory and field investigation to evaluate geocomposite drain systems for use on low-volume roads. The laboratory testing program was designed to determine the flow capacity of various manufactured systems when subjected to varying lateral loads and hydraulic gradients. Because no standard tests were available to evaluate geocomposite drain system performance under field conditions, a test system consisting of a large triaxial chamber and special plumbing was developed. Geocomposite test specimens were placed vertically in a 6- by 12-inch mold, which was then filled with a compacted silty soil. Changes in flow rates through the specimens were measured as both the gradient and lateral pressure were varied. A wide range in performance was noted among the different systems. At the lower confining pressures, the flow capacities were fairly similar. As the confining pressures were increased, the flow capacity of some drain systems dropped markedly. However, all products tested, except one, had a minimum flow rate of about 1 gaVmin per foot of drain width when subjected to a confining pressure of 30 psi and a hydraulic gradient of 1.0. In conjunction with the laboratory testing, three field installations of prefabricated drain systems were instrumented. All three sites had piezometers in front of and behind the drain systems, and one site had an outflow recording device. Results of these installations validated the laboratory test results.

Appendix 6.6

Field and Laboratory Evaluation of Geocomvosite rain Svstems for Use on Low-volume ~ b a d s

I he I-{irert S r ' n ~ c c~,ndui.ted .In in~ i lcp th hhor;nor! ,id iield ~nvcm.;!tlon 10 c'i.iluati. pcocomporlte dr:iln \)itern\ fin urr. on Ihu-\olurns n,.rd\ I he Iahur.ttw! redn? propram w.i\ derlpned to durr.rminc thc ilm c;ip;!c~ty of \;m,u\ rmimufactured \)\tern\ whro \uhjr.ctt.d ~ $ 1 \ ;cr)mp l a teu l I h d r and hydr;iulr gr;idit.nt\ Ur.c.ue no ,t.gnd.grd rcrtr uerc i l \ in l , th/~ 10 e \ i h i i t ~ pcocorn~~ pobite d r m \yrti.m pi.rtorm:$nc~. under iield condirun*. ;i tert \?\tern ckrnr~ \ r ino t a hrpe t c i a x ~ l ch:irnhcr and \peci:d [,lumhtn~ uilr dcvrlopcd ( i r onmpmi fe trrt rpcoirnrm were placed vcr- t~call) In a h - h! I?- ln. mold. which w.!\ thrn filled w ~ t h :I com- pouted \Or) wxl. Change, In flou rate, through the rprcirnrm ue i s me.iwrcd '1, hoth the grd lent and lateral p r rsurc sere barled A \\I& ianpr in pcrfurm:mcc ua \ n n r d : m u q the dl l - k rcnr \)\term At rhr Ihwer ionfming prcwurc,. rhe tlow c.tpac- (tie\ wcrc tmrl! \imll.tr hr the wntmnp pic\rurL.\ nere incji.n*ed. thc I lou c;ip:iclt! (11 v m u di.ifn \?rturn\ dropped rn;!rLcdl!. H w - e\er. ,311 prt,duct\ tu\fcd. except ,mu. Ih;d .b mmrnum l lou i J t r of ;!hour I e;il mln per l w t a i dram w d t h u h r n \ u h p t ~ . d to ;I

cmtmlnp pr rnure of 311 prl m d J hydraulx prid>cnt ,,I I 11. In con junctm u i l h the idhoraror) tertmg, three field in\r;tIl;monr of prrt:lhric;itcd dr:iin \)\tern\ were inmumentud. A l l t h i r r vte\ had plerinrnetrrr in fnmt of and hchind the drain r)*tern\, and m e *it< h i d .m i r u l i l w recordmg dt.\lcr. Rs\ult\ t , i thr \ r imral- I h t ~ m s v:ql~Iated the laboratory tv\t r ~ ~ l t ~

GEOCOMPOSITE DRAIN SYSTEMS

( ieocumpmite drain rystrms. originally c:illed rm d r a m <,i prefabr~cated drain wrtemh. consist o f a grotcxt i l r covering

USDA Fwcrl Srrwce. 2245 Morello A b e . Pleasant HIII. C.1111 44523.

Appendix 6.6

TEST PROGRAM

A teht program was debeloped to i leterni~t ic rhc t h ~ i.;ip;~ctt) o f the geocompmite drain hv\tcmr whcn w h l c c t d 1 ~ 1 \;ir\lni:

the core material by the lateral loads.

LABORATORY TESTING

FIGURE I Laboratory lest syslem.

FIELD 1NSTALI.ATIOSS

In conjunction n l th the I r lh~r i~ tor ! Ic\ting. lhrcc ticl i l lrl\l.lL lation\ of peocompmitu d w n \?\tcnl\ uc lc 11~rrrtimi.ntci1 tc,r

4 8 17 16 20 24 28

CONFINING PRESSURE. PSI

FIGURE 2 Equilibrium flow versus conlining pressure: 1.0 gradienl.

Appendix 6.6

LSOI

. m i e ( nun ' lk ( l Je5Ua.L :()()()t u!wpeq( ; :.lain[ aua 'u!eJpexuy :u!wpJam\ :il!.)ndm ulal\i* :i~ua!pala 0.1 ' a ~ n * c u d Xa!uguw !sd-~c '*imp 91-01 aluy mwar wou w n ! ~ q ! l ! n N F :4H.>:)I.4

FIGCIKE S Equilibrium now versus time ill-51Hl &a?*. 311-psi runlining prewwr. 1.0 pradirnll: r>rtcm rap;lcit(: Enkadmin, two lay,: Wradrain 61WO: Tmsar D V I . I r e lays': Coslrrh Slripdrain 125 pil.

The ~ L . S s~.lected (or tuturc monlt~mnp u r r c Site I i. uhwh involved a retaming wall ;ippronim~itely 1511 tt lkme and 21 i t high at i t \ high~.\t pomt. and S~te 17. mhich required ;i w.di ill0 ft long and 21.5 i t h~gh. Imtiiilatton {lf the drain \)rtcnl proceeded without dtfficulty. The roll\ E l p k i l n wme In lengths specified 10 fit the proleut cond~t~on\ . A \ \hewn in

Figure 6 . thu drains ucre ptnncd ;st thc top ;and unnllled ~ n t o the collcctinn trench .it the hare ot the d~a in \\rtcrn K i t h the dram \ystcm in place. the wall could he huilt e;i\tI!. The cwnpicted wall at Sits 17 1s shoun in Flgurc. 7.

The rnomtormg cqulpment c.onslsted of hand-f.ihrir;ited pl- ezometer sensorr inst:!llrd in front of and hehind the dr;un system. The Omn~data Datapod Model DP212 ha\ urcd to

E q u i l i b r i u m flow, gprn/ft Year T e s t e d P r o d u c t G r a d i e n t : 0 . 3 1 0 2 . 0

1986 System C a p a c i t y 6 . 8 1 1 . 7 1 6 . 9 1987 Amerdra in 3 . 0 5 . 2 7 . 5 1986 C o n t e c h S t r i p d r a i n * 0 . 6 1 . 2 1 . 8 1986 E n k a d r a i n 9 1 2 0 , 1 L a y e r 0 . 4 0 . 9 1.2 1986 E n k a d r a i n 9 1 2 0 , 2 L a y e r 1 . 5 2 . 8 4 . 1 1987 M i r a d r a i n 4000 0 . 4 0 . 9 1 . 2 1987 M i r a d r a i n 6000 4 . 6 8 . 1 1 1 . 8 1987 Tensar D N 1 , I Layer 0 . 8 1 . 5 2 . 2 1987 Tensar D N 1 , 2 L a y e r 1 . 6 2 . 8 4 . 0

* 25 p s i c o n f i n i n g pressure

Appendix 6.6

FIGURE 6 lnrlalllltlon sf drain s)stcm at Silr 17 on Stump Springs Road.

t en installed op\lope of the drun shuncd iluctuating higher groundwater le\els. Momtoring at Sitc 17 a150 rhnwrd a direct

FIGURE 7 Completed wall al Site 17 on Slump Springs Road.

Appendix 6.6

On the hasis of results ot thc Iahoratory ~ o r k ,lnd held ill\t,il- latians. the following concluwn\ were drawn:

. . study.

0 The equilibrium tlmv of ;i lproduct :at ;! conttn~ni. Iprc\\urc can usually he drtermincd after ;i pcrtcld w\i'r.cl d.i\\. f4owcver. the confining prc\\ure ma? need to hc ~m. !~n t .~n~ t l for longer period\. up to \cver,fil rmmthi.

Reduct~on in flow capacity of thc drmn \ntcn1 nl.t! lcwl t from crushing or compresion af the ccxe rm;itert.tl 01' tnlm elongation of the geotcxtile due to incrc.txd \ r , ~ l prc>rilrc\.

A l l oroducts tested. ereem tor the Contech Str~pd!.icn (which parttally collapsed at 25 ~ I I . had mmtmum q o ~ l t h ~ rium flow rates of about I eal'min ocr toot i,f dlmn wdth when subjected to ;I confining p r u w m crt 311 p\i ;and ;I h\dr.irtl~c gradient of I.O. 'The Contech product trmmuttcd thi, \ ~ ~ l u m i . at a confining presure of ?i pbl.

8 Munitor~ng of sebrral t d d in\tall;munr $ ' ~ump<h ! t c drain systems has shown that thehe *!\tuns funct~on .I\ d c y w d As demonstrated during ;t severe \term. peoc,,rnpcwli, dr.un systems can accommodate tranhicnt uul l .I* \tc.~d>-\~.ttc groundwater condit~ons.

A l l field installations were eaiy to in\taIl. For most Forest Service applicatalns m\<ll\lng Idti.rd

pressures less than 15 psi and low nuu. rate\. an!. c,f thc t c \ t d systems would perform satisfactorily.

FIGURE 8 Site 17 on Stump Springs Road: (top) daily average water table elevation, (middle) daily drain nun. (boaom daily precipitation.

Appendix 6.6

FIGIRE 9 lnslallatiun sf drain \\<tern en \lesquito Kidgc Road.

The duthor\ w r h 1,) ektend their appreciatmn to Kodncy W. Prsllwitr of the Forest S e n k c ' \ Intermr,unt;iin Station lor developing the recording goundwa le r moniturinp equipment

Appendix 6.6

FIGURE I0 Completed till on Mosquilo Ridge Road.

and methudology and t o A l m W c a ~ e r and P.tuhne ( ' ; ! l ~ t I i . formerly of the Slerra Nat~onai Forest. for thctr ;i.v.t:lnce in mon~tor inp the Stump Springs Road \ ~ t c s . Thank\ ;(re ;ilw extended to L.enn! Lethaby and Phil Feidman 01 the Forc.\t Service's Pleasant H ~ l l Lahor;~tory for the11 aw\t;mcu in con- ducting the laboratory testing p rug r . , n .

APPENDIX 6.7

Soil Nailing

Introduction

John E. Steward, Chief Geotechnical and Dams Engineer, Washington Office

Soil nailing is a method of reinforcing soils and slopes in situ by inserting reinforced elements through drilling and, more recently, ballistic methods. Conventional soil nailing typically involves installing and grouting reinforcement rods into drilled holes. Unlike tiebacks, the nails are passive elements that are not post tensioned.

Soil nailing has been used in Europe during the last two decades. The U.S. Department of Transportation is leading the effort in the United States to develop guidelines for soil nailing in transportation works (U.S. Department of Transportation 1991). Readers are referred to US. Department of Transportation and other publications on soil nailing.

A new type of soil nailing called launched or ballistic soil nails shows promise for rapid and economical repair of unstable road shoulders. This method, described in the following paper, uses high pressure compressed air to rapidly insert 20-foot-long, 1.5-inch-diameter, galvanized steel rods (nails) and perforated drain pipes into the soil.

The following paper was prepared for the 1994 International Roads Federation Conference in Calgary, Alberta, Canada.

Appendix 6.7

LAUNCHED SOIL NAILS A NEW TECHNOLOGY FOR STABILIZING

FAILING ROAD SHOULDERS

by John E. Steward, Chief Geotechnical and Dams Engineer

USDA Forest Service

Paper prepared for presentation

at the Innovative Design and Construction Session

of the 1994 IRF Conference

Calgary, Alberta

Appendix 6.7

LAUNCHED SOIL NAILS A NEW TECHNOLOGY FOR STABILIZING

FAILING ROAD SHOULDERS

by John E. Steward, Chief Geotechnical and Dams Engineer

USDA Forest Service

ABSTRACT

A variety of methods have been used during the last 20 years to reinforce soils. One of these methods is soil nailing. Most often, soil nails are installed by inserting steel rods in drilled holes then grouting in-place. Sometimes the nails are inserted using percussion methods. These methods generally require excavation of a working bench in order for the construction equipment to work below the slope being nailed. These methods are not applicable to repair of small slips of road fills and embankments where access is limited.

Launched soil nailing, a new technique developed in the United Kmgdom, by Soil Nailing, Ltd., allows nails to be inserted into the slope using a launcher attached to the end of an excavator boom. Using this method, the nails can be installed into slopes up to 25 to 35 feet above or below the road surface without excavation or ground disturbance. The launcher utilizes high pressure compressed air to install the nail. The depth of penetration depends both on the com- pressed air pressure and the in situ material.

In July and August of 1992, the USDA Forest Service sponsored a launched soil nailing demon- stration project in the western United States. The five week demonstration involved installation of launched soil nails at eight sites in four States and three Forest Service Regions. The demon- stration included soil nailing of road shoulders, a retaining wall, a cut bank, and a sand bank. Financial and technical assistance was provided by the Federal Highway Administration Coordi- nated Technology Implementation Program, the States of Washington and Colorado, and the seven National Forests that participated in the demonstrations.

The project successfully demonstrated the feasibility of using launched soil nails for the stabili- zation of failing road slopes. Small slope failures (no deeper than about 15 feet) can be stabi- lized for about $14 per square foot of slope face. Low height retaining walls and excavate and replace methods typically cost $15 to $60+ per square foot of face area. Equipment mobility, minimum site disturbance, and low costs indicate a strong future for launched soil nails for the repair of the road infrastructure.

Appendix 6.7

INTRODUCTION

In July and August of 1992, the USDA Forest Service and Soil Nailing Limited from the United Kingdom sponsored a launched soil nail demonstration project in the western U.S. The demon- stration involved installation of launched soil nails at eight sites in four States and three Forest Service Regions. Demonstrations included soil nailing of road shoulders, retaining walls, a cut bank, and a sand bank. Financial assistance was provided by the Federal Highway Admini- stration Coordinated Technology Implementation Program (CTIP). Technical assistance was provided by the Washington and Colorado Departments of Transportation, and seven National Forests that participated in the demonstration project.

The project was developed to demonstrate the use of launched soil nails for repair and reinforce- ment of unstable cut bank and embankment slopes. The demonstrations provided an opportunity for engineers, maintenance personnel, and contractors to view and explore the potential for using launched soil nails.

A video, an Application Guide (1) and the Project Report For Launched Soil Nails-1992 Dem- onstration Project, (2) are products of the demonstration project. Demonstration site experi- ences, participant interviews and questionnaires, and the simplified wedge analysis for soil nailing provide the basis for this paper.

DEMONSTRATION PROJECT RESULTS

The soil nail launcher successfully installed 1.5 inch (38 mm)* (3) diameter, 18 foot (5.4 m) long galvanized steel nails into a wide variety of materials. Launcher air pressures of 600 to 2500 psi (4.1 to 17.2 KPa) resulted in nail tip penetrations of 5 to 18 plus feet (1.5 to 5.4 m).

A production rate of 15 nails per hour was achieved by a 3-person work crew (launcher operator and two helpers).

Participants in the demonstrations indicated:

High potential for slope reinforcement, especially road shoulder and backslope rein- forcement. Medium potential for retaining wall reinforcement, horizontal drain, and anchor inser- tion. Either tracked or rubber tired excavators are suitable; tracked excavators may be more versatile. Use of a self propelled rubber tired excavator for road shoulder repairs could eliminate the excavator hauling unit. Minimum ground disturbance and mobility are important features of the technology. Potential limitations: - Length of nails; limited to smaller slides. - Penetration in cobbly soil. - Controlling depth of nail penetration.

* This paper uses soft metrics: English units followed by Systems International (SI) units in parenthesis. The conversion in the reference section, from ASCE Civil Engineering magazine will aid quick mental calculations.

Appendix 6.7 1069

Design concerns: - Nail pullout resistance. - Practical design guidance. - Need more experience with the technology; case histories. - Corrosion for permanent installations. Facing systems would be appropriate for temporary walls and for very shallow slides or erodible soils.

Several actions have been initiated as a result of these demonstrations:

Design charts for stabilizing road shoulders with launched soil nails are included in this paper (Figures 7, 8.9). A longer term (I to 2 years) demonstration project is being planned to gain experience with completed projects and to develop case histories.

SOIL NAILING USING LAUNCHED NAILS

Soil nailing is a soil reinforcement technique that inserts long steel rods into an unstable or potentially unstable existing soil mass. Soil nails installed into the soil act to reinforce the soil mass by transferring tensile and shear resistance of the nail to the soil. The nails maintain the restraint force because they are anchored beyond the slip surface. Figure 1 shows how these forces act to retain a small soil slip.

During the past 20 years, a variety of methods have been used to install soil nails. Most often they are inserted into drilled holes and then grouted in place. Sometimes they are driven into the soil using percussion methods. These methods generally require the excavation of a working bench (Figure 2) for the equipment.

Launched soil nailing, also called ballistic soil nailing, is a new technique developed in the United Kingdom. Soil nails are installed by means of a launcher mounted on a hydraulic exca- vator (Photos 1 and 2). The launcher utilizes high pressure air acting upon a collet (plastic collar) attached at the tip (front end) of the nail (Photo 3). Compressed air suddenly released against the collet forces the collet and nail through the launcher barrel, much like a dart through a blow gun (Figure 3).

The nails are launched at speeds of over 200 miles per hour (320 km/'), at pressures approach- ing 2500 psi (17.2 M Pa). The collet breaks away as the nail enters the soil. As the launched nail passes into the soil, the ground around the nail is displaced by compression at the tip. This forms an annulus of compression (Figure 3f), reducing soil-drag on the nail and damage to the galvanized coating. Depth of nail penetration is normally controlled by air pressure and ground resistance. Optionally, the nail penetration can be arrested by fitting the end of the nail with a tapered screw-on coupling (Figure 3e). During launching, the force (air pressure) acts upon the tip of the nail, thus placing the nail temporarily in tension, preventing the nail from buckling.

The launcher typically launches plain or galvanized steel nails up to 1.5 inches (38 mm) in diameter and up to 20 feet (6 m) in length. (NOTE: 18 foot (5.5 rn) long nails were used for the 1992 demonstration due to the length of the shipping container.) The nail should be oriented normal to the potential slip plane to act primarily in shear and bending, with the tension being induced by movement.

Appendix 6.7

Depending upon the length of the boom, the launcher can be positioned 5 to 35 feet (1.5 to 11 m) above or below the excavator's platform (Figure 4). The attachment to the boom is by an articu- lated knuckle (Photo 2) that allows tilting the launcher to almost any desired angle. Excavation for a working bench is usually not needed for road repairs using the launcher.

The soil nail launcher has been used in the United Kingdom to successfully install nails in a variety of soil and slope conditions, primarily for reinforcement of road and railroad embank- ments and to strengthen retaining walls. Prior to this demonstration, the equipment had not been used in the United States.

NEED FOR LAUNCHED SOIL NAILS

Roads constructed on steep slopes are susceptible to sliding and shoulder cracking (Photo 4). These cracks allow water from rain and snow melt to enter the cracks, adding excess moisture and water pressure directly to the slide mass. These areas are periodically filled and patched to smooth the road, adding weight to the sliding mass and further decreasing stability.

These fill failures are costly to repair, impair safe travel, and can cause extensive damage to the surrounding land and streams. Obviously, permanent repair methods are preferred over the annual crack filling and patching of these unstable areas.

Launched soil nails offer a rapid economical alternative to reoccurring maintenance or other reconstruction solutions. Often several small fill failures can be fixed in one day without excava- tion. The launcher can be moved with ease between trees and shrubs, resulting in little or no vegetation removal and little need for environmental or visual mitigation (Photo 5).

The soil nail launcher, which weighs about 1000 pounds (453 kg) mounted on a standard hydrau- lic excavator, is highly mobile and can be mobilized rapidly for quick response. Small slides can be quickly stabilized before they progress into larger slides. This quick response prevents more expensive repairs and further environmental damage.

DESIGNING WITH LAUNCHED SOIL NAILS

A number of methods can be used to account for the reinforcement benefit to the slope using launched soil nails. Soil Nailing Limited developed a design method using the simplified wedge analysis (Figure 5) (1). The soil nails impart both tensile and shear resistance from the nail to the soil as illustrated in Figures 6a and 6b.

The I992 demonstration project provided a qualitive demonstration of the equipment capability. Test sites were not designed for stabilizing moving slopes. However, areas of known movement were selected for most demonstration sites to judge potential performance.

As a result of the field demonstrations and work with technical advisors, a simplified wedge design methodology was developed to aid in selecting nail spacing to stabilize small road shoul- der slides on low volume roads. Typically these slides may require 15 to 50 nails to stabilize, at a cost of $2,000 to $6,000 per site. Geotechnical drilling can cost $3,000 to $10,000 and is usually not warranted for these slides. The design method assumes a site evaluation has been performed by experienced geotechnical personnel, usually without exploratory drilling.

Appendix 6.7

Design charts developed for 26" (1:2) slopes, 33" (1: 1.5) slopes, and 45" (1: 1) are shown in Figures 7.8, and 9 respectively. The design charts are based on several assumptions.

Material factor of safety, fm, of 2.0 is applied to nail pullout (tension) and shear resis- tance. Site factor of safety, fn, of 1.1 is applied to the overall stability analysis. The number of nails, N, is for each 3.3 feet (1 m) along the road shoulder. Soil unit weight of 127 pounds per cubic foot (3034 kg/m3). The existing field condition is at limit equilibrium (i.e., existing site factor of safety, fn, = 1.0). The slope has been in place for a number of years and can be represented by the con- solidated - undrained condition during slope movement. The soil strength can be represented by an effective cohesion of zero (C'=O) and an apparent angle of internal friction of 0 (phi) estimated from site failure geometry, soil classification, and seepage conditions. Ground water and seepage pressures are either minimal or controlled by installed drainage. Nails are installed nearly normal to the slide plane. The depth through the active zone into the resistant zone and in the active zone is at least 3.3 feet (1 m) to develop nail resistance. The top row of nails is placed about 3.3 feet (1 m) from the road shoulder, the bottom row of nails is no closer than 3.3 feet (1 m) above the toe of the slide, and the remain- ing nails are evenly distributed throughout the slide mass.

The full design method is contained in an Application Guide (1).

The design process requires an assessment of the cause of failure and collection of field data. The required design information can be gathered from a visual assessment of the site and prepa- ration of a field developed cross-section. Figures 10 and 11 are sample forms that can be used in the field to gather the required site information.

To ensure full penetration by the soil nails, the soil should not contain a high percentage of cobbles or boulders. Launching nails in ordinary sands, gravel, silts, and clays or mixtures of these are no problem. Penetration will be reduced in dense gravels and stiff clay. A few cobbles and boulders will not be a problem since penetration can still be achieved even if the nail is deflected into another portion of the soil. Nail locations can be adjusted around obstacles to install the correct number of nails. The launcher can easily be repositioned and a replacement nail installed for the nails blocked by subsurface objects.

A "best estimate" of subsurface conditions at the site is necessary to evaluate stability and con- duct a preliminary design of nail spacing. The Field Data Form (Figure 10) should be used to note the general soil, rock, vegetation, drainage, grade, etc., at the site. An estimate of the subsurface moisture condition at the time when slope movements occurred is essential in the overall evaluation of stability. Engineering geologists or geotechnical engineers should perform the field evaluation and design.

Physical features should be located accurately on the site plan or cross-section. Examples of physical features are areas of seepage or wet soil, large trees or stumps, boulders, and exposed bedrock.

1072 Appendix 6.7

The Site Factor Check List (Figure 1 I) contains site factors to adjust the nail spacing for local site conditions. The site factor evaluation is based on local conditions, the confidence in the site condition assessment (probability of sliding), and the consequence of continued slope movement. Generally the High site condition deserves a more critical design review (higher site factor of safety, fn) than the Low site condition. The Site Factor Check List is suggested for selecting an appropriate site factor of safety. An example of a completed Field Data Form is included in the Design Example below.

A "High" ground water condition indicated by "seeps" (Figure 12) must be controlled by installed drainage. When the site condition is High, or when the consequences of additional sliding are High, it is recommended that more in-depth site investigations andlor mathematical slope stabil- ity analyses be performed before a final repair alternative is selected.

Since seepage pressures can have a major effect on the stability of the slope, it is best to install seepage control measures. Drilled horizontal drains and drainage trenches are commonly used to control ground water and seepage pressures in slopes. Launched horizontal drains can also provide the needed drainage.

A high water table will effect the geometry of the slide, resulting in a larger slide and a lower apparent soil 0. Use of launched horizontal drains (Figure 12) and appropriate apparent 0 may counter the need to increase the number of nails to account for the ground water table. This question will be answered as we complete and monitor full-scale field installations. Until then, it is recommended that either the number of nails be increased or the groundwater be controlled in areas with active seepage.

The number of nails, N, from Figures 7 ,8 , and 9 can be adjusted to fit the condition. Adjust- ments of 0.5 N for Low, N for Medium, and 1.5N for High conditions will yield overall factors of safety (fm + fn) of about 1.1, 1.2, and 1.3, respectively.

Figure 13a shows the preferred diagonal nail pattern.

DESIGN EXAMPLE

Figures 14a and 14b show the completed Field Data Forms and Site Factor Check List for a typical road failure site on an older road in steep, mountainous terrain. The desigr of the launched soil nail stabilization for this site follows:

Design Information x = 10 feet (3 m) H = I8 feet (5.5 m) X --0.6 H - 8 = 32' use @ = 30" 8 = 42" use 0 = 45"

Number of Nails per 3.3 feet (1 m) along road shoulder

From Figure 9 Curve C N=4

Appendix 6.7

Site Condition

Medium; use N = 4 Note: For High condition use 1.5 N = 6

For Low condition use 0.5 N = 2

Nail Spacing on the Slope

Total Number of Nails

Assuming the unstable area is within the limits on the sketch and is rectangular with two rows of nails outside the defined site area, the area to be nailed is:

Area - -- - 62 nails Areahail - 27

Final selection of the number of nails and the nail spacing will depend on the following consid- erations:

The risk and consequence of failure assessed in terms of loss of life, property damage. environmental damage, and traffic disruption. (Low, Medium, or High from Figure I I .)

The existing stability of the slope and the ability to support the weight of the launcher and excavator (approximately 43,000 lbs, 19,500 kg).

The sequence of nail installation to enhance the stability of the working area

The maximum depth to the slip surface, perpendicular to the slope surface not exceed- ing 15 feet (4.5 m) for 20-foot (6-m) nails.

The site factor of safety applied relating to the level of confidence in and certainty of the factors influencing the slope's stability.

Evaluation of the influence of the ground water and surface water at the worst case seasonal condition.

The durability of the nail. Factors that may accelerate corrosion must be appraised. High or low ground water levels, pH conditions, and the presence of external contami- nates such as road salt, organic debris, and waste leaches should be examined. Galva- nized steel nails are expected to last as long as galvanized steel culverts in similar conditions.

Appendix 6.7

COST ESTIMATING

The Design Charts can be used in conjunction with the Field Data Form to estimate the number of nails required. After setup on the site, the launcher is capable of installing 15 nails/hours. A cost range of $80 to $135/nail is appropriate for an initial cost estimate for the launched soil nail repair alternative, including mobilization.

LOGISTICS

Since the excavator normally works from the roadway, little or no site work is usually required for equipment access. On two lane roads, traffic can usually proceed using traffic control, with full traffic stoppage only during actual launching. Single lane roads may require longer delays in traffic. The excavator can be moved out of the way for traffic passage after several launches.

The support equipment needed for the soil nail launcher is minimal. The launcher can be moved to a site, set up, launch nails, and move off the site in one day. The launcher can be removed from the excavator's mounting within 30 minutes. A heavy duty flat bed trailer or truck is needed to transport the launcher (1,000 pounds, 453 kg), rods (120 pounds each, 54 kg), and miscellaneous supplies.

OTHER POTENTIAL APPLICATIONS

Horizontal drains. Landslides are frequently associated with ground water and ground water seeps. Drilled horizontal drains have proven to be effective in reducing or controlling the effect of this ground water. Launched perforated pipes up to 20 feet (6m) in length have been used to drain local areas.

Vertical eas venting. Vertical perforated plastic and metal pipes have been used to vent methane gas from landfills. This application proved fast and safe for the installers.

Wall strengthening. This is a method for rapidly adding reinforcement to the materials behind retaining walls for the purpose of replacing deteriorating tiebacks, supporting increased external loading, supporting excavation at the toe, and compensating for aging components (Photos 13 to 16).

Ground anchors and tiebacks. With a typical pull out resistance of 2,000 to 3,000 pounds (9 to 13.5 kN), direct pull anchor uses may be limited.

Facines and mesh hold&. Support of mesh on rocky slopes and to support erosion control materials on raveling slopes and fills.

Temporay excavation suppofi. To hold an excavated face until a permanent wall is constructed or while work is completed in the area and back filled.

Road widening. To steepen a cut slope or build a small permanent wall at the toe of a cut slope as an alternative to widening the fill, building a retaining structure, or moving into the cut slope.

. . Cut slope stab~hzation. To add reinforcement and stabilize cut slopes.

Appendix 6.7

9. Vertical u. To aid dewatering or consolidating loose materials such as dredge spoils and wet areas under roadways and bridge approaches.

SUMMARY

Launched soil nails have been used to stabilize road and railroad related landslides in Europe and the United Kingdom since 1989. The technology was successfully demonstrated in the western USA in 1992.

Design charts for selecting the number of nails required to stabilize small landslides for road shoulders and embankments have been developed using the simplified wedge analysis method. The chart method of design is appropriate for low volume roads where the cost of geotechnical drilling is generally not warranted. Other methods of design will certainly develop as case history projects are designed and constructed.

Launched soil nails appear to be an effective, rapid, and practical method for stabilizing small road shoulder landslides. Full scale design and construction projects are needed to verify the design method proposed in this report and generate commercial interest in the technology.

REFERENCE

(1) . Draft Applications Guide for Launched Soil Nails, unpublished, USDA Forest Service, Washington, DC, January 1994.

(2) . Draft Project Report for Launched Soil Nails - 1992 Demonstration Project, unpublished, USDA Forest Service, Washington, DC, January 1994.

(3) . "Metric Conversion", ASCE Civil Engineering Magazine, New York, NY, December 1994.

METRIC CONVERSION

1 acre = 0.4 ha I acre-ft = 1.233 m'

I ft = 0.3 m I sq ft = 0.09 m2 1 cu ft = 0.03 mJ

I gal (U.S.) = 0.004 m3 1 in = 25.4 mm

I sq in = 645 mm2 1 lbf = 4.5 N

I Ibm = 0.5 kg

1 psf = 48 Pa I psi = 6.9 kPa

1 mgd = 0.04 m'/s 1 m i = 1.6km

1 sq mi = 2.6 km2 1 short ton = 907 kfg

1 yd = 0.9 m 1 sq yd = 0.8 mJ 1 cu yd = 0.8 mJ

Appendix 6.7

/ DISTURBING FORCES - ~

ACTIVE ZONE SLIP SURFACE

~.~

RESTORING FORCES

RESISTANT ZONE

Figure I . Forces acting on a road slopefailure.

Sprayed concrete face

I-\ a. Excavate veltical face c. Build wall face

b. Drill and groute soil nails d. Drill and groute soil nails in slope

Figure 2. Convenn'onal soil nailing

Appendix 6.7

Breech 7 / Release Valve

a. Nail presented for loading.

Breech lntedock

Breech i 111 I "'I

Nolse and

c. Nail firing by air pressure release.

Breech lnterlodr

Release Valve

Breech

Flared Head 01 Nail Arrested by Washer

(Optional) Nail

. Nail fully installed and arrested. (Flared Head and Arresting Washer are optional; not used in as Demonstration Project.)

Breech Inter(

Release Valve Collel

Breech (1.11

Debrls Shroud

b. Nail loaded in launcher.

d. Nail impact with collet release.

I soil

f. Annulus of soil compression around nail.

Figure 3. Illusrrarion of Soil Nail Launching Sequence

1078 Append'i 6.7

Appendix 6.7

X

X' -1 4 crack

Where: W = Mass of sliding wedge y = Unit mass of the soil X' = Distance to visible crack X = Distance to projected slide plane W = I / Z y x H

Figure 5. Simplified wedge forces.

Pressure on nail below slide piant causes resistance to nail pullout. An 1,. 7,; of 2.0 is used to caicu- late the allowable friction on the nail and resulting tension capacit)

/ r, =normal soil pressure on

f,= material factor of safety

Figure 6a. Tensile resistance of nail.

The ultimate bearing capacity, q,,, of the nail is available lo resist soil movement. An f,, of 2.0 is used to calculate the allowable beanng capaclry in f)\ calculat~ng tne ummate shear resistance. Su, of the nails ?b

Figure 66. Shear resistance of nail.

Appendix 6.7

38 mrn (1.5 in) diameter nails

Figure 7 Number of nails required ro srabilize a road shoulderfor 1:2 (26') slope.

Appendix 6.7

Value of H (meters, feet)

Figure 8. Number of nails required to stabilize a road shoulder for 1 : I S (33') slope.

Appendix 6.7

38 mm (1.5 in) diameter nails /

Value of H (meton, feet)

x/H = 0.8 6 8 x/H =0.4 c .!w-o.s -a-

.........o*~"' phi = 30" ...--- phi - 309

E a ' x / H = ~ . 8 ..2 2 * = 0 . 4 G -2 x/H = 0.2 ..... phi = 30" phl - 35" / phi = 40'

Figure 9. Number of nails required to stabilize a road shoulder for 1:l (45') slope.

Appendix 6.7

Slip

FIELD DATA FORM FOR LAUNCHED SOIL NAILS

Road Name Road No. - Data-

Mile PostIStation Location T. - R.- Sec.

General Site Description:

Repair Priority

(1-10 High)

Completed by:

Projected limit of slide

7 Hinge Point

*{ In* Surface

HW

02 A Original Ground

To toe of embankment if lower HW limit of slide cannot be located B

0

Cross-Section

Inside Shoulder 7 --a3 D4 Cracks

I t X' O u t s l d e l d e r

t (Hinge Point)

Slide Limits

Toe of Slide

Plan

Figure 10.

Appendix 6.7

SlTE FACTOR CHECK LIST FOR LAUNCHED SOIL NAILS

Unified or AASHTO Classificalion " Design charts were developed for medium Risk Level (Site Factor of Safety. 1, = 1.1) I Relates to the probability of failure. 2 Relates to the consequence of failure.

SlTE SPECIFIC PLAN AND (OR) CROSS-SECTION

Figure I I .

Appendix 6.7

Figure 12. Warer table consideration.

Figure 13. Nail pattern.

Appendix 6.7

FIELD DATA FORM FOR LAUNCHED SOlL NAILS

Repair P W l y 7

(1-lOH@l)

Completed by: a

SITE FACTOR CHECKLIST FOR LAUNCHED SOlL NAILS

Figure 14a. Design example field form. Figure 146. Design example field form.

Photo I . The excavator with the launcher artached installing soil nails in Colorado. To prevent complete failure of the slope as seen in the foreground. The nailed area has not moved since the nails were installed in 1992.

Photo 2. A close-up view of the launcher. Numbers on photo show: I ) Nail guide. 2 ) Air chamber and valve, 3) Barrel. 4 ) Noise and debris shroud, 5 ) spring loaded safety switch. 6 ) Aniculated knuckle, and 71 Excavator boom.

Appendix 6.7

Photo 3. Soil m i l ready for insertion loading into the launcher. Compressed air is introduced between rhe locking washer ( I ) and the collet (2) . The collet separates from the nail in the noise and debris shroud (Photo 2).

Photo 4. Failing road shoulder typical of those needing srabilization.

Appendix 6.7

Phoro 5. Nails installed in road shoulder without disrurbing rhe vegetation. The nails were curoff ar rhe ground surjace. Norice rhe pavement displacement in the right foreground area ( I ) .

- Phcro 6. hunching nails into a backslope in Northern California

Appendix 6.7

References US. Department of Transportation. 1991. Soil Nailing for Stabilization of Highway Slopes and Excavations. McLean, VA: Federal Highway Administration.

Appendix 6.7

Slope Stability Index Volume I11

active wedge . . . . . . . . . . . . . . . . 847.848.850. 1001. 1007 angle of internal friction . . . . . . . . . . . . . . . . . . . . . . . . . 758 annual wonh . . . . . . . . . . . . . . . . . . . . . . . . . . 863.866. 867 aquifers . 767.770. 802 . 863.935. 939. 978. 980. 985. 986

perched . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 770

back-calculation analysis . . . . . . . . . . . . . . . . 734. 735. 1042 Bayes Theorem . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 1049 Bishop method of slices . . . . . . . . . . . . . . . . . . . 837.839. 941 barehole instrumentation . . . . . . . . . . . . . . . . . . . . . . . . . 785 breaksven analysis . . . . . . . . . . . . . . . . . . . . . . . . . 868-869 Bmler Slide . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 750-756 brush layering . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 819

Camp Fivc Slide . . . . . . . . . . . . . . . . . . . . . . . . 800.935.987 clearing . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 900. 901 Colorado Rockfall Simulation Program (CRSP) . . . . . . . . 858 compaction . . . . 735. 818.823.825.900.905.907. 908

moisture control . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 902 weaving and pumping . . . . . . . . . . . . . . . . . . . . . . . . . 902

. . . . . . . . . . . . . . . . . . . Conditional Probability Thwrem 888 construction convol . . . . . . . . . . . . . . . . . 823.897.91 1.926 contracting officer (CO) . . . . . . . . . . . . . . . . . . . . . 898 . 1032 contracting ofticer's representative (COR) . . . 898 . 899. 901.

903 . 905 . 907. 1026 . . . . . . . . . . . . . . . . . . . . . . . . . . . . . counterbalance fill 848

crest venical curves . . . . . . . . . . . . . . . . . . . . . . . . . 742-744 cutoff trenches . . . . . . . . . . . . . . . . . . . 768.770. 808. 809. 850

Darcy's Law . . . . . . . . . . . . . . . . 784.788.935.978.980.986 decision analysis . . . . . . . . . . . . . . . . . . . . . . . . . . . 861-895 decision trees . . . . . . . . . . . . . . . . . . . . . . 861. 881. 886-893

. . . . . . . . . . . . . . . . . . . . . . . . . . . . . . dcsignsegments 853 deterministic analysis . . . . . . . . . . . . . . . . . . . . . . . . 872 . 882 direct shear . . . . . . . . . . . . 774.810.823. 842.935.941. 1040 dominance principle . . . . . . . . . . . . . . . . . . . . . . . . . . . . 882 drive probe exploration . . . . . . . . . . . . . . . . . . . . . . . . . 785

factor of safetv (FOS) . . . . . 734.736 . 739 . 740 . 748 . 754.755 .

941 . 977. 993. 995.9%. 998. 1W1 . 1W7 . 1042 . I043 Fairview Sanitary Landfill . . . . . . . . . . . . . . . . . . . . . . . 800

. . . . . Federal Highway Administration (FHWA) 747. 820.913. 918.987. 1068

field-developed cross-sections . . . . 733. 756.757 . 767 . 784 . 809 flattened slopes . . . . . . . . . . . . . . . . . . . . . . . 738.740. 742 flaw lines . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 935. 977 fluorescein . . . . . . . . . . . . . . . . . . . . . . . . . . . 786.935.938 future wonh . . . . . . . . . . . . . . . . . . . . . . . . . . 863.866. 867

. . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . gabion baskets 803 . . . . . . . . . . . . . . . . geocomposite drain 821 824 908 1053-1061 . . . . . . . . . . . . . . . . . . . . . . . . . . geogrid reinforcement 746

gealextile . . . . . . . . . . 746. 768. 770. 805. 817. 819. 822. 824. 850 . 907.908. 1010

. . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . grubbing 900.901 gunite . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 857

. . . . . . . . . . . . . . . . . . . . . . . . . . . . . hazardassessment 872 Highlands Interchange Slide . . . . . . . . . . . . . . . . . . . . . . 802

. . . . . . . . . horizontal drains 754. 768.780. 781.783.785.788. 791. 793.797. 799. 802 . 853. 862. 868. 869. 876. 877 . 880. 883 . 884.935.977. 985. 987 . 1040. 1043. 1044

Horseshoe Bend Hill . . . . . . . . . . . . . . . . . . . . . . . . . . . 747 hydraulic conductivity . . . . . . . . . . . . 786. 788. 978. 979. 986 hydraulic gradient . . . . . . . . . . . . . . . . . . 786 . 788.978. 1053

instrumentation . . . . . . . . . . . 785. 914.916 . 918.921.924.926

Janbu method of slices . . . . . . . . . . . . . . . . 7N). 870.935. 941

E Krassel mck slope failure . . . . . . . . . . . . . . . . . . . . . 756.761

eanh buttresses . . . . . . . . . . . . . . . . . . . . . . . . 758.805.806 L effective end spacing . . . . . . . . . . . . . . . . . 788 . 790. 792.981

. . . . . . . . . . . . . . . . . . electrical resistivity 785. 935. 938. 939 Laplace Principle . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 883 Elk River . . . . . . . . . . . . . . . . . 808.814.816.826. 838-840 Latin Hypercube simulation . . . . . . . . . . . . . . . . . . . 874. 875 equipotential lines . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 980 Level I Stability Analysis (LISA) . . . . . . . . . . . . . . . 874 . 875 expected monetary value (EMV) . . . . . . . . 885.888.890.892. lightweight embankments . . . . . . . . . . . . . . . . . . . . . . . . 740

894.895 . I043 limit equilibrium . . . . . . . . . . . . . . . . . . . . . . . . . . . 820. 821 expected preference value (EPV) . . . . . . . . . . . . . . . . 894. 895 LISA . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 874. 875 expected value . . . . . . . . . . . . . . . . . . . . . . . . . 882 . 883.886

Volume UI I- 1

Manning's Equation . . . . . . . 784 . 788 . 789 . 935 . 978 . 979 . 986 Markland's test . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 853 mchaniurlly stabilized embankment (MSE) . . . . 742 . 803 . 805 .

904905 . 907 Monte Carlo simulation . . . . . . . . . . . . . . . . . . 874.875.887 most probable future principle (MPFP) . . . . . . . . . . . 882 . 883

optimist principle . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 884 optimum moisture content (OMC) . . . . . . . . . . . . . . . 9U2.908 ordinary method of slices (OMS) . . . . . . . . . . . . . . . . . . 759 Oregon Dewinen1 of Transpartation (OWT) . . . . . . 841.842.

passive wedge . . . . . . . . . . . . . . . . . . . 846.847 . 1001 . 1004 PCSTABM . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 820 ~ss imis t urincide . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 884 . . photognunmetry . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 800 phreatic surface . . . 739. 752. 783. 787.788.790.981 . 987 pieromelen . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 785.1039 pipegeomvy . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 770 planning horizon . . . . . . . . . . . . . . . . . . . . . . . 862.863. 866 plasticity index . . . . . . . . . . . . . . . . . . . . . 822 . 935 . 941 . 993 Powder Creek Slide . . . . . . . . . . . . . . . . . . . . . . . . 801 . 1039 prefecence function . . . . . . . . . . . . . . . . . . . . . . . . . 894.895 present worth . . . . . . . . . . . . . . . . . . . . . . 863.864.866.868 probabilistic analysis . . . . . . . . . . . . . . . . . . . . . . . . . . . 883 pullout resistance . . . . . . . . . . . . . . . . . . . . . . . . . . . 819. 822

Quentin Slide . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 801

reinforced fills . . . . . . . 733.734. 817. 820. 822. 826. 828. 832. 834. 838 . 839 . 906

relative density . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 902 residual angle of friction . . . . . . . . . . . . . . . . . . . . . . . . . 993 retaining walls . . . . . . . . . . 734.746.757.760.764.807. 817 .

819.820 . 863 . 905. 909 . 1016. 1017 rhodamine WT . . . . . . . . . . . . . . . . . . . . . . . . . . . . 786. 938 @RISK . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 875-876 risk analysis . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 872-873 road management objectives (RMO's) . . . . . . . . . . . . . . . 737 rock buttresses . . . . . . . . . 749.797. 803.805. 81 1. 862. 1043

riprap mck buttresses . . . . . . . . . . . . . . . . . 803.884. 1043 rulting . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 903

. . . . . . . . . . . . . . . . . . . . . . sag vertical curve 742.744. 755 salvage value . . . . . . . . . . . . . . . . . . . . . . . . . 862.863.868 sensitivity analysis . . . . . . . . . . . . . . . . . . . 869. 870. 872. 874 shear dowels . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 855-856 shear key . . . . . . . . . . . . . . 747.748.778.780. 841.843. 844 shear strength . . . . . . . 767.785.823.841. 842.850.903.906.

935. 937.941.949-952. 1040 . 1042 shear trenches . . . . . . . . . . . . . . 841.850 . 883.884.99L.1012.

1042-1043 shotcrele . . . . . . . . . . 851.857. 1015. 1016. 1026. 1030-1035 slot width . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 789. 981 sodiumchloride . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 786 soil nailing . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 1065-1090 Spencer's method of slices . . . . . . . . . . . . . . . . . . . . . . . 760 STABGM . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 820 standard penemtion test (SPT) . . . . . . . . . . . .785 . 1040 . 1041 static water level (SWL) . . . . . . . . 786 . 800-802 . 935. 978.985 steepened slopes . . . . . . . . . . . . . . . . . . . . . . . 733.739.740

tensioned rock bait . . . . . . . . . . . . . . . . . . . . . . . . . . . . 854 TENSU) . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 820 Total Probability Theorem . . . . . . . . . . . . . . . . . . . . . . . 888 trench geometry . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 770

underdrains . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 789.908 Unified Rock Classificalion System (URCS) . . . . . . . . . . . 937 Unified Soil Classification System (USCS) . . . . . 745. 874 . 993.

I0401041 untensioned mck bale . . . . . . . . . . . . . . . . . . . . . . . . . . 854 UTEXAS2 . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 820

Vann slope stability investigation . . . . . . . . . . . . . 748-750

weaving and pumping . . . . . . . . . . . . . . . . . . . . . . . 902-903 Willow Slide . . . . . . . . . . . . . . . . . . . . . . . . . . . . . . 826-828 Wwten'sThirdLaw . . . . . . . . . . . . . . . . . . . . . . . . . . . 734

X XSTABL . . . . . . . 750. 752.756. 760. 762.763. 771. 776.780.

808 . 81 1. 814-816. 820 . 826. 834. 842. 843 . 870. 871. 941.949.976. 987. 993. 996.1004. 1007-1010

Volume UI

Engineering Management Series

Administrative Distribution

The Series THE ENGINEERING MANAGEMENT SERIES (EM) publications are published periodically as a means of exchanging engineering-related ideas and information. Each EM is designed to serve a specific purpose and, therefore, is distributed to an audience interested in that particular subject.

Submittals Field personnel should send material through their Regional Information Coordinator for review by the Regional Office to ensure inclusion of information that is accurate, timely, and of interest Service-wide.

Regional R-1 Clyde Weller R-4 Ted Wood R-9 Fred Hintsala Information R-2 Lois Bachensky R-5 Rich Farrington R-10 Betsy Walatka Coordinators R-3 Bill Woodward R-6 Carl Wofford WO Bryon Foss

R-8 Bob Bowers

Inquiries Regional Information Coordinators should send material for publication and direct any questions, comments, or recommendations to the following address:

Forest Service--USDA Engineering Stafl- Washington Office AlTN: Mary Jane Senter, Editor

Sonja Turner, Asst. Editor 201 14th Street, SW Washington, DC 20250

Telephone: (202) 205-0820

This publication is an administrative document that was developed for the guidance of employees of the Forest Service--US. Department of Agriculture, its contractors, and its cooperating Federal and State Government Agencies. The text in the publica- tion represents the personal opinions of the respective authors. This information has not been approved for distribution to the public and must not be constnred as recom- mended or approved policy, procedures, or mandatory instructions, except by Forest Service Manual references.

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This information is the sole property of the Government with unlimited rights in the usage thereof and cannot be copyrighted by private parties.

Slope Stability Reference Guide for National Forests in the United States-Volume Ill

August 1994 EM-71 70-13